Frontiers in Offshore Geotechnics II

PROCEEDINGS OF THE 2ND INTERNATIONAL SYMPOSIUM ON FRONTIERS IN OFFSHORE GEOTECHNICS, PERTH, AUSTRALIA, 8–10 NOVEMBER 2010

Frontiers in Offshore Geotechnics II Editors Susan Gourvenec & David White Centre for Offshore Foundation Systems, University of Western Australia

© 2011 by Taylor & Francis Group, LLC

Cover photo credit © Norske Hydro

CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business © 2011 Taylor & Francis Group, London, UK Typeset by MPS Limited, a Macmillan Company, Chennai, India Printed and bound in Great Britain by Antony Rowe (a CPI Group Company), Chippenham, Wiltshire All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Published by:

CRC Press/Balkema P.O. Box 447, 2300 AK Leiden, The Netherlands e-mail: [email protected] www.crcpress.com – www.taylorandfrancis.co.uk – www.balkema.nl

ISBN: 978-0-415-58480-7 (Hbk + CD-ROM) ISBN: 978-0-203-83007-9 (ebook) © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Table of Contents

Preface

XIII

Committees

XV

Reviewers

XVII

1 Keynotes A systematic approach to offshore engineering for multiple-project developments in geohazardous areas T.G. Evans Recommended best practice for geotechnical site characterisation of cohesive offshore sediments D.J. DeGroot, T. Lunne & T.I. Tjelta Gulf of Guinea deepwater sediments: Geotechnical properties, design issues and installation experiences J.-L. Colliat, H. Dendani, A. Puech & J.-F. Nauroy Geotechnics for subsea pipelines D.J. White & D.N. Cathie

3 33

59 87

Axial and lateral pile design in carbonate soils C.T. Erbrich, M.P. O’Neill, P. Clancy & M.F. Randolph

125

New frontiers for centrifuge modelling in offshore geotechnics C. Gaudin, E.C. Clukey, J. Garnier & R. Phillips

155

Risk and reliability on the frontier of offshore geotechnics R.B. Gilbert, J.D. Murff & E.C. Clukey

189

2 Geohazards and gas hydrates Neotectonic deformation of northwestern Australia: Implications for oil and gas development J.V. Hengesh, K. Wyrwoll & B.B. Whitney

203

Deepwater Angola part I: Geohazard mitigation A.J. Hill, J.G. Southgate, P.R. Fish & S. Thomas

209

Deepwater Angola part II: Geotechnical challenges A.J. Hill, T.G. Evans, B. Mackenzie & G. Thompson

215

Shallow gas hazard linked to worldwide delta environments S. Kortekaas, E. Sens & B. Sarata

221

Analysis of submarine flow slides in fine silty sand P.V. Lade & J.A. Yamamuro

227

Hydrate dissociation around oil exploration infrastructure A.K. Sultaniya, J.A. Priest & C.R.I. Clayton

233

An investigation of past mass movement events in the West Nile Delta S. Thomas, L. Bell, K. Ticehurst & P.S. Dimmock

239

Deformation of seabed due to exploitation of methane hydrate reservoir J. Yoneda, M. Hyodo, Y. Nakata, N. Yoshimoto & R. Orense

245

© 2011 by Taylor & Francis Group, LLC

V

3 In situ site characterisation and pore pressure measurement A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils D. Borel, A. Puech & S. Po

253

Enhancement of the ball penetrometer test with pore pressure measurements N. Boylan, M.F. Randolph & H.E. Low

259

Laboratory free falling penetrometer test into clay S.H. Chow & D.W. Airey

265

Offshore sediment overpressures: Overview of mechanisms, measurement and modeling B. Dugan, T.C. Sheahan, J.M. Thibault & T.G. Evans

271

Angolan deepwater soil conditions: GIS technology development for sediment characterization P. Enjaume, M. Hamon, K. Epalanga & B. Mackenzie

277

Strength measurement in very soft upper seabed sediments P. Kelleher, H.E. Low, C. Jones, T. Lunne, S. Strandvik & T.I. Tjelta

283

CPT in polar snow – preliminary observations A.B. McCallum, A. Barwise & R. Santos

289

Parametric study of a free-falling penetrometer in clay-like soils M. Nazem & J.P. Carter

293

The future of deepwater site investigation: Seabed drilling technology? J.J. Osborne, A.G. Yetginer, T. Halliday & T.I. Tjelta

299

Mini T-bar testing at shallow penetration A. Puech, M. Orozco-Calderón & P. Foray

305

Piezometer installation in deepwater Norwegian Sea T.I. Tjelta & J. Strout

311

Luva deepwater site investigation programme and findings T.I. Tjelta & A.G. Yetginer

315

Investigations into novel shallow penetrometers for fine-grained soils Y. Yan, D.J. White & M.F. Randolph

321

Seabed drilling vs surface drilling – a comparison A.G. Yetginer & T.I. Tjelta

327

4 Soil characterisation and modelling Rheological behaviour of soft clays P.E.L. de Santa Maria, I.S.M. Martins & F.C.M. de Santa Maria

335

A three-dimensional finite element study of the direct simple shear test J.P. Doherty & M. Fahey

341

Repeated loading and unloading of the seabed H.J.E. Hu, K.K. Tho, C.T. Gan, A.C. Palmer & C.F. Leung

347

A new interpretation of the simple shear test H.A. Joer, C.T. Erbrich & S.S. Sharma

353

Physical modelling of the crushing behaviour of granular materials H.A. Joer & S.S. Sharma

359

New evidence for the origin and behaviour of deep ocean ‘crusts’ M.Y-H. Kuo, M.D. Bolton, A.J. Hill & M.J. Rattley

365

Soil unit weight estimated from CPTu in offshore soils P.W. Mayne, J. Peuchen & D. Bouwmeester

371

© 2011 by Taylor & Francis Group, LLC

VI

Strain rate dependent simple shear behaviour of deepwater sediments in offshore Angola M.J. Rattley, A.J. Hill, S. Thomas & B. Sampurno Simplified calibration procedure for a high-cycle accumulation model based on cyclic triaxial tests on 22 sands T. Wichtmann, A. Niemunis & Th. Triantafyllidis Understanding cyclic loading behavior of soil for offshore applications J. Yang

377

383 389

5 Shallow foundations Observations of shallow skirted foundations under transient and sustained uplift H.E. Acosta-Martinez, S. Gourvenec & M.F. Randolph

397

Numerical study of grillage foundations on sand under combined VHM loading M. Banimahd, A. Maconochie & J. Oliphant

403

The vertical bearing capacity of grillage foundations in sand M.F. Bransby, P. Hudacsek, J.A. Knappett, M.J. Brown, N. Morgan, D.N. Cathie, R. Egborge, A. Maconochie, G.J. Yun, N. Brown & A. Ripley

409

Behaviour of skirted footings on sand overlying clay C.T. Gan, K.L. Teh, C.F. Leung, Y.K. Chow & S. Swee

415

Numerical study of piping limits for suction installation of offshore skirted foundations and anchors in layered sand L.B. Ibsen & C.L. Thilsted

421

Shallow foundation performance in a calcareous sand B.M. Lehane

427

A numerical study of the vertical bearing capacity of skirted foundations D.S.K. Mana, S. Gourvenec & M.F. Randolph

433

The effect of torsion on the sliding resistance of rectangular foundations J.D. Murff, C.P. Aubeny & M. Yang

439

Foundation design challenges of the MCR-A skirted gravity platform L. Tapper, C. Humpheson & B.M. Lehane

445

Constructing breakwater with prefabricated caissons on soft clay S. Yan, X. Feng & J. Chu

451

6 Piled foundations Simplified analysis of laterally loaded pile groups F.M. Abdrabbo & K.E. Gaaver

459

Behavior of piles under combined lateral and axial loading M. Achmus & K. Thieken

465

Investigations on the behavior of large diameter piles under cyclic lateral loading M. Achmus, J. Albiker & K. Abdel-Rahman

471

BP Clair phase 1 – Pile driveability and capacity in extremely hard till T.R. Aldridge, T.M. Carrington, R.J. Jardine, R. Little, T.G. Evans & I. Finnie

477

Photoelastic investigation into plugging of open ended piles J. Dijkstra, E.A. Alderlieste & W. Broere

483

Soil-pile interaction during extrusion of an initially deformed pile C.T. Erbrich, E. Barbosa-Cruz & R. Barbour

489

BP Clair phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till T.G. Evans, I. Finnie, R. Little, R.J. Jardine & T.R. Aldridge

495

© 2011 by Taylor & Francis Group, LLC

VII

Pile driving experiences in Persian Gulf calcareous sands K. Fakharian & I.H. Attar

501

FLAC3D analysis on soil moving through piles E.H. Ghee & W.D. Guo

507

Cyclic loading of barrettes in soft calcareous rock using Osterberg cells C.M. Haberfield, D.R. Paul, M.C. Ervin & G.A. Chapman

513

Shaft capacity of drilled and grouted piles in calcareous sandstone B.M. Lehane

519

Numerical analysis of mudmat contribution to capacity of piled offshore platforms L.S.D. Lorenti, M.A. Ismail & B.M. Lehane

525

Simplified numerical model for analysis of offshore piles under cyclic lateral loading M.M. Memarpour, M. Kimiaei & M. Shayanfar

531

Centrifuge modelling of rapid load tests with piles in silt and sand C.T. Nguyen, H. van Lottum, P. Hölscher & A.F. van Tol

537

Field measurements on monopile Dolphins A. Sadeghi-Hokmabadi & A. Fakher

543

Behaviour of driven tubular steel piles in calcarenite for a marine jetty in Fujairah, United Arab Emirates J. Thomas, M. van den Berg, F. Chow & N. Maas CPT-Based design method for axial capacity of offshore piles in clays B.F.J. Van Dijk & H.J. Kolk

549 555

7 Foundations for renewable energy Evaluation of pile capacity approaches with respect to piles for wind energy foundations in the North Sea M. Achmus & M. Müller

563

Installation of suction caissons for offshore renewable energy structures O.J. Cotter, B.W. Byrne & G.T. Houlsby

569

Lateral behaviour of large diameter monopiles at Sheringham Shoal Wind Farm L. Hamre, S. Feizi Khankandi, P.J. Strøm & C. Athanasiu

575

Centrifuge modelling of offshore monopile foundation R.T. Klinkvort & O. Hededal

581

Gravity based foundations for the Rødsand 2 offshore wind farm, Denmark L. Krogh, J.H. Lyngs & J.S. Steenfelt

587

Geotechnics for developing offshore renewable energy in the US M. Landon Maynard & J.A. Schneider

593

Engineering issues for fixed offshore wind turbines on Lake Michigan Mid Lake Plateau, USA P.J. Lang, J.A. Schneider, K. Smith & T. McNeilan

599

Centrifuge model tests on piled footings in clay for offshore wind turbines B.M. Lehane, W. Powrie & J.P. Doherty

605

Design of monopile foundations in sand for offshore windfarms M. Saue, T.E. Langford & N. Mortensen

611

Experimental evaluation of backfill in scour holes around offshore monopiles S.P.H. Sørensen, L.B. Ibsen & P. Frigaard

617

An investigation of the use of a bearing plate to enhance the lateral capacity of monopile foundations K.J.L. Stone, T.A. Newson, M. El Marassi, H. El Naggar, R.N. Taylor & R.J. Goodey

623

© 2011 by Taylor & Francis Group, LLC

VIII

Optimizing site investigations and pile design for wind farms using geostatistical methods: A case study B. Stuyts, V. Vissers, D.N. Cathie, C. Jaeck & S. Dörfeldt

629

Towards the FE prediction of permanent deformations of offshore wind power plant foundations using a high-cycle accumulation model T. Wichtmann, A. Niemunis & Th. Triantafyllidis

635

Cyclic accumulation effects at foundations for offshore wind turbines H. Wienbroer, H. Zachert, G. Huber, P. Kudella & Th. Triantafyllidis

641

Study on soil-structure interaction of suction caisson by large-scale model tests B. Zhu, D.Q. Kong, L.G. Kong, R.P. Chen & Y.M. Chen

647

8 Jack-up units Simplified VH equations for foundation punch-through sand into clay J.-C. Ballard, P. Delvosal, P.H. Yonatan, A. Holeyman & S. Kay

655

Characterisation of undrained shear strength using statistical methods B. Bienen, M.J. Cassidy, M.F. Randolph & K.L. Teh

661

Centrifuge modelling of spudcan deep penetration in multi-layered soils M.S. Hossain, M.F. Randolph & Y.N. Saunier

667

A probabilistic approach to the prediction of spudcan penetration of jack-up units G.T. Houlsby

673

An assessment of jackup spudcan extraction O.A. Purwana, H. Krisdani, X.Y. Zheng, M. Quah & K.S. Foo

679

3D FE analysis of the installation process of spudcan foundations G. Qiu, S. Henke & J. Grabe

685

Undrained bearing capacity of deeply embedded foundations under general loading Y. Zhang, B. Bienen, M.J. Cassidy & S. Gourvenec

691

9 Anchoring systems Trajectory prediction for drag embedment anchors under out of plane loading C.P. Aubeny & C.-M. Chi

699

Setup following keying of plate anchors assessed through centrifuge tests in kaolin clay A.P. Blake, C.D. O’Loughlin & C. Gaudin

705

Seismically-induced displacements of a suction caisson in soft clay A.J. Brennan, S.P.G. Madabhushi & P. Cooper

711

SEPLA keying prediction method based on full-scale offshore tests R.P. Brown, P.C. Wong & J.M. Audibert

717

Set-up of suction piles in deepwater Gulf of Guinea clays J.-L. Colliat & D. Colliard

723

Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners and set-up time J.-L. Colliat, H. Dendani, H.P. Jostad, K.H. Andersen, L. Thorel, J. Garnier & G. Rault

729

Numerical FEM and laboratory study of the bearing capacity factor Nc for plate anchors L.N. Equihua-Anguiano, M. Orozco-Calderón, P. Foray & M. Boulon

735

Caisson capacity in clay: VHM resistance envelope – Part 2: VHM envelope equation and design procedures S. Kay & E. Palix Installation and in-place assessment of drag anchors in carbonate soil M.P. O’Neill, S.R. Neubecker & C.T. Erbrich © 2011 by Taylor & Francis Group, LLC

IX

741 747

Caisson capacity in clay: VHM resistance envelope – Part 1: 3D FEM numerical study E. Palix, T. Willems & S. Kay

753

Numerical investigation of the behaviour of suction caissons in structured clays S. Panayides, M. Rouainia & A. Osman

759

Cyclic moment loading of suction caissons in sand B. Zhu, B.W. Byrne & G.T. Houlsby

765

10 Pipelines and risers Multidirectional analysis of pipeline-soil interaction in clay R.G. Borges & J.R.M.S. Oliveira

773

Geotechnical challenges for deepwater pipeline design – SAFEBUCK JIP D.A.S. Bruton, M. Carr & F. Sinclair

779

Large deformation finite element analysis of vertical penetration of pipelines in seabed S. Chatterjee, M.F. Randolph, D.J. White & D. Wang

785

Implementation of geotechnical techniques in the analysis of pipeline response G. Cumming & N. Brown

791

Lateral soil resistance to an untrenched pipeline under the action of ocean currents F.P. Gao, S.M. Yan, E.Y. Zhang, Y.X. Wu & X. Jia

797

Vertical cyclic testing of model steel catenary riser at large scale T.E. Langford & V.M. Meyer

803

Kupe gas project pipeline – optimisation of discrete rock berm design shore approach B.L. Larsson

809

Model test studies on soil restraint to pipelines buried in sand R. Liu, S.W. Yan & J. Chu

815

Pipe-soil interaction on clay with a variable shear strength profile D.R. Morrow & M.F. Bransby

821

Sweeping behaviour of shallowly-embedded pipeline during cyclic lateral movement T. Takatani

827

Advanced nonlinear hysteretic seabed model for dynamic fatigue analysis of steel catenary risers I.H.Y. Ting, M. Kimiaei & M.F. Randolph

833

Mobilization distance in uplift resistance modeling of pipelines J. Wang, S.K. Haigh, N.I. Thusyanthan & S. Mesmar

839

Theoretical, numerical and field studies of offshore pipeline sleeper crossings Z.J. Westgate, M.F. Randolph, D.J. White & P. Brunning

845

Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE® D.J. White, A.J. Hill, Z.J. Westgate & J.-C. Ballard

851

11 Trenching, ploughing, excavation and burial Influence of object geometry on penetration into the seabed A. Ivanovi´c, R.D. Neilson, G. Giuliani & M.F. Bransby

859

Investigation into the effect of forecutters on plough performance K.D. Lauder, M.J. Brown, M.F. Bransby & J. Pyrah

865

State-of-the-art jet trenching analysis in stiff clays J.B. Machin & P.A. Allan

871

Numerical modelling of soil around offshore pipeline plough shares W. Peng & M.F. Bransby

877

© 2011 by Taylor & Francis Group, LLC

X

Anchor–chain–rockfill–soil interaction: Evolution of design methods H. van Lottum & H.J. Luger

883

Development of a jet trenching model in sand J.-F. Vanden Berghe, J. Pyrah, S. Gooding & H. Capart

889

12 Design and risk Structural factors affecting the system capacity of jacket pile foundations J.Y. Chen, R.B. Gilbert, J.D. Murff, A.G. Young & F.J. Puskar

897

The new API Recommended Practice for Geotechnical Engineering: RP 2GEO P. Jeanjean, P.G. Watson, H.J. Kolk & S. Lacasse

903

Comparison of ISO 19901-2 and API RP 2A seismic design criteria for a site in the Caspian Sea, Turkmenistan Z.A. Lubkowski, J.E. Alarcon & Z.A. Razak

909

Offshore geotechnics – safe and sustainable J. Peuchen & J. Haas

915

Author Index

921

© 2011 by Taylor & Francis Group, LLC

XI

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Preface

The International Symposia on Frontiers in Offshore Geotechnics (ISFOG), hosted by the Centre for Offshore Foundation Systems (COFS) at the University of Western Australia (UWA), provide a platform for academics and practitioners to discuss the emerging challenges in offshore geotechnical engineering and present recent practice and research. Offshore design and construction presents unique challenges to geotechnical engineers. Many of the challenges that are routinely encountered today have persisted for decades and continue to be gradually overcome with advances in technology and analysis. Meanwhile, new challenges are faced as the industry moves to harness resources in deeper waters, harsher environmental conditions and new frontier regions with previously-uncharacterized seabed conditions. The inaugural ISFOG was hosted by COFS in September 2005 and arose from discussions with colleagues from around the world which highlighted a consensus that there was a niche for a new specialist offshore geotechnical conference. That symposium was attended by 182 delegates from 22 countries and the proceedings contained 7 keynote papers and 127 peer-reviewed general papers that reflected the state-of-the-art of practice and research (Gourvenec and Cassidy, Eds, Taylor & Francis, ISBN 041539063X). Immediately afterwards and over the years following the symposium, we received extremely positive feedback from many of the participants, who had found the event to be both valuable and enjoyable. As a result, in the latter part of 2007 we canvassed the International Advisory Committee (IAC) of the inaugural ISFOG – a 30-strong team of leading academics and practitioners – for their views on a second ISFOG. They offered their unreserved support for a repeat event and it was agreed that a 5-year gap would be appropriate between symposia. The original Australian Research Council grant that led to the establishment of COFS expired in 2005. In some respects the first ISFOG was a chance to mark the achievements of COFS during its original tenure, at what might have been its zenith. However, COFS has grown since 2005 and is now more active and larger than ever. We therefore felt able to organise a 2nd International Symposium on Frontiers in Offshore Geotechnics, inviting friends and colleagues, old and new, to visit Perth again in November 2010. On advice from the IAC, we retained the single session format of the inaugural ISFOG, which kept participants together throughout the event, allowing continuity of discussions and avoiding clashes between concurrent presentations. Emphasis on the poster sessions and general reports offers exposure to the papers that inevitably do not receive full oral presentation given the constraints of the single session format. Key themes of the inaugural ISFOG are still as relevant 5 years on and therefore remain unchanged for the second ISFOG – these include geohazards, site investigation techniques and foundations for renewable energy, as well as anchoring solutions for deep water and pipeline geotechnics. Greater emphasis has been placed on in situ soil testing techniques and pore pressure measurement, and the characteristics of unusual seabed deposits found in frontier regions. The keynote papers reflect the key stages of an offshore project. The first three keynotes cover the assessment and interpretation of offshore geohazards, in situ site characterisation and pore pressure measurement, and the behaviour of West African clays – West Africa being a critical frontier region with particular challenges; two further keynotes describe state-of-the-art design considerations for piled foundations and pipelines; a further keynote describes recent advances in centrifuge modelling and the final keynote examines risk and reliability in offshore geotechnics. The keynotes alone contain many man-years of accumulated experience. We are sincerely grateful to the keynote authors, particularly those from industry who have generously found time to contribute their expertise to these comprehensive papers. The papers collected in these proceedings include the 7 keynotes and a further 117 peer-reviewed general papers that represent the current state-of-the-art in offshore geotechnics. These provide an invaluable resource to all those working in offshore construction, design and research. Each of the papers has been peer-reviewed by at least 2 reviewers, drawn from COFS and from academic institutions and industry within Australia and around the world. We are indebted to the reviewers for all their efforts, which have ensured that the papers are of a very high standard. We also thank Divya Mana and Santiram Chatterjee of COFS for their editorial assistance whilst assembling these proceedings. Stephanie Boroughs and Monica Mackman of COFS have provided administrative support throughout the organisation of this second

© 2011 by Taylor & Francis Group, LLC

XIII

ISFOG, and we acknowledge in particular Stephanie’s tireless work corresponding with the authors and reviewers to compile the proceedings. The assistance of the Local Organising Committee is also acknowledged, in particular Noel Boylan of COFS who co-ordinated our sponsorship programme. Finally, COFS is grateful for the support for ISFOG provided by the following organisations and individuals: the Australian Department of Innovation, Industry, Science and Research through the International Science Linkages Programme, our industry sponsors, the members of our International Advisory Committee and the International Society for Soil Mechanics and Geotechnical Engineering (ISSMGE) under whose auspices the ISFOG symposia are held. Susan Gourvenec and David White July 2010

© 2011 by Taylor & Francis Group, LLC

XIV

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Committees

LOCAL ORGANIZING COMMITTEE Susan Gourvenec (Chair)

Centre for Offshore Foundation Systems (COFS), UWA Stephanie Boroughs (Sec.) COFS, UWA Monica Mackman (Sec.) COFS, UWA Britta Bienen COFS, UWA Nathalie Boukpeti COFS, UWA Noel Boylan COFS, UWA Mark Cassidy COFS, UWA Liang Cheng School of Civil and Resource Engineering (SCRE), UWA Fiona Chow Chair, Australian Geomechanics Society (AGS), WA Chapter James Doherty SCRE, UWA Sarah Elkhatib Arup Martin Fahey COFS, UWA Ian Finnie Advanced Geomechanics

Andy Fourie SCRE, UWA Christophe Gaudin COFS, UWA Andrew Grime Arup Jim Hengesh COFS, UWA Shazzad Hossain COFS, UWA Yuxia Hu SCRE, UWA Mehrdad Kimiaei COFS, UWA Barry Lehane SCRE, UWA Mark Randolph COFS, UWA Yinghui Tian COFS, UWA Dong Wang COFS, UWA Phil Watson Advanced Geomechanics David White COFS, UWA Long Yu COFS, UWA Hongxia Zhu COFS, UWA

INTERNATIONAL ADVISORY COMMITTEE Susan Gourvenec (Chair) David Airey Peter Allan Marcio Almeida Knut Andersen Charles Aubeny Fraser Bransby Nick Brown Byron Byrne Tim Carrington John Carter David Cathie Johnny Cheuk Chris Clayton Ed Clukey Jean-Louis Colliat Don DeGroot Jason DeJong Earl Doyle Carl Erbrich Trevor Evans Albert Griffith

© 2011 by Taylor & Francis Group, LLC

Australia Australia UK Brazil Norway USA Australia Australia UK UK Australia Belgium Hong Kong UK USA France USA USA USA Australia UK Saudi Arabia

Andy Hill Phil Hogan Jacques Garnier Bob Gilbert Harry Kolk Richard Jardine Andy Lane Colin Leung Jayme Mello Roger Moore Don Murff Julian Osborne Andrew Palmer Alain Puech Matthew Quah Richard Raines Roderick Ruinen Marc Senders Dan Spikula Jørgen Steenfelt Tor Inge Tjelta Shuwang Yan

XV

UK USA France USA Netherlands UK Australia Singapore Brazil UK USA UK Singapore France Singapore USA Netherlands Australia USA Denmark Norway China

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Reviewers

Tony Abbs Martin Achmus Hugo Acosta-Martinez David Airey Tom Aldridge Marcio Almeida Knut Andersen Senthil Arasu Charles Aubeny Jean-Christophe Ballard Edgard Barbosa-Cruz Britta Bienen David Bonjean Nathalie Boukpeti Noel Boylan Fraser Bransby Andrew Brennan David Bruton Byron Byrne Tim Carrington Mark Cassidy Santiram Chatterjee Liang Cheng Johnny Cheuk Fiona Chow Patrick Clancy Ed Clukey Richard Dean Andrew Deeks Jason DeJong James Doherty David Edwards Clarence Ehlers Sarah Elkhatib Ed Ellis Carl Erbrich Martin Fahey Ian Finnie Andy Fourie Christophe Gaudin

© 2011 by Taylor & Francis Group, LLC

Guido Gottardi Susan Gourvenec Jürgen Grabe Andrew Grime Ole Hededal James Hengesh Andy Hill Matt Hodder Phil Hogan Shazzad Hossain Guy Houlsby Yuxia Hu Hans Hugel Mostafa Ismail Ana Ivanovic Christophe Jaeck Richard Jardine Eric Jas Hackmet Joer Steve Kay Lindita Kellezi Feizi Khankandi Yoshiaki Kikuchi Mehrdad Kimiaei Melissa Landon Thomas Langford Kok Kuen Lee Barry Lehane Colin Leung Nina Levy Tom Lunne Jon Machin Chris Martin Jayme Mello Richard Merifield Vaughan Meyer Neil Morgan Damian Morrow Don Murff Tejas Murthy

XVII

Andrzej Niemunis Steve Neubecker Tim Newson Elio Novello Conleth O’Loughlin Michael O’Neill Julian Osborne Ashraf Osman Andrew Palmer Jeff Priest Richard Raines Mark Randolph Mike Rattley Oliver Reul David Richards James Schneider Marc Senders Shambhu Sharma Tom Sheahan Dan Spikula Doug Stewart Kevin Stone Kar Lu Teh Luc Thorel Yinghui Tian Manh Tran Jean-Francois Vanden Berghe Alastair Walker Dong Wang Phil Watson Zack Westgate David White Nobutaka Yamamoto Jun Yang Gulin Yetginer Long Yu George Zhang Hongxia Zhu

1 Keynotes

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

A systematic approach to offshore engineering for multiple-project developments in geohazardous areas T.G. Evans BP Exploration Operating Company, Sunbury-on-Thames, UK

ABSTRACT: The development of economic and safe oil and gas facilities in geotechnically challenging and geohazard-prone areas calls for inter alia early engagement of geospecialists, fit-for-purpose data acquisition, geological understanding, engineering pragmatism and innovation and perhaps above all, multi-disciplinary teamwork. Jeanjean et al. (2005) described some of the strategies and practices developed by BP that embody these basic principles and this paper updates and expands these ideas and concepts with special reference to the investment sharing and planning opportunities afforded by programmes of multiple projects. 1

INTRODUCTION

centralised Geohazard Assessment Teams (GATs) and regional ground modelling. This paper develops the geohazard engineering risk management themes and methods described by previous authors, with particular reference to the opportunities for investment sharing and learning from BP’s multiple-project development programmes in deep water offshore Angola and Egypt. The paper is in two parts: Part 1 (Section 2) is a review of the some of the main engineering challenges faced by operators such as BP in geohazardous environments and Part 2 (Section 3) describes some of the ways that BP is tackling these challenges in Angola and Egypt as part of its asset management process.

In their keynote paper to ISFOG 2005, Jeanjean et al. (2005) described the challenges faced by oil and gas operators when developing offshore facilities in deep water frontier areas and environmentally and geologically difficult settings. The main theme of the paper was geotechnical and geological risks faced by projects operated by BP in geohazard-prone areas in the Caspian Sea, West Nile Delta (WND) and the Gulf of Mexico (GoM), with particular reference to the lessons learnt from the Mad Dog and Atlantis fields that are located along the Sigsbee Escarpment in GoM. The underlying message of the keynote was that geohazards do not necessarily preclude safe offshore developments provided operators are proactive and invest sufficient resources and planning in the geohazard management process. Some of the principles and best practices described in the paper include the use of multidisciplinary geospecialists, integrated and phased field-wide geophysical surveys and geotechnical surveys, development of new fit-for-purpose survey tools and use of project-specific engineering analyses to inform risk assessments. From about 2003 the focus of BP’s new deepwater projects shifted from the GoM towards the Caspian Sea, Angola and Egypt West Nile Delta (WND). Each of these areas has its own specific geotechnical and geohazard challenges that require some form of rigorous management. Significantly, BP’s developments in Angola and Egypt involve rolling programmes of projects that have provided systematic learning and investment leveraging opportunities in geohazard risk management, integrity management and value engineering not necessarily available for single projects. Evans et al. (2007) and Moore et al. (2007) introduced some of the multiple-project investment concepts that have been applied in the WND, including the use of © 2011 by Taylor & Francis Group, LLC

2

OFFSHORE GEOTECHNICAL ENGINEERING CHALLENGES

2.1 The geotechnical maze The ever-increasing geotechnical challenges faced by offshore operators in pursuit of hydrocarbons in frontier areas have been widely reported in the past decade: Power and Clayton (2003), Kvalstad (2007), Jeanjean et al. (2003, 2005), Evans et al. (2007), Evans et al. (2010) and Hadley et al. (2008) and many others. However, as discussed by Evans et al (2007), offshore exploration and development policy is rarely, if ever, dictated by shallow subsurface and geotechnical considerations. Rather, the shallow geo-specialists are expected to work as part of a bigger team to: (1) ensure that shallow-subsurface issues do not compromise project-life safety and function (loss prevention) and (2) offer cost-effective geotechnical solutions. These two goals are particularly onerous in frontier areas where the challenges for geotechnical and foundation engineering are especially labyrinthine, as illustrated on Figure 1. The chaotic nature of this figure shows

3

Figure 1. Geotechnical challenges.

how offshore geotechnical engineering challenges can arise from the interactions of one or more negative factors that can result in increased costs and risks to exploration, development and production. For the purpose of this paper the geotechnical challenges are assumed to have eight root causes:

seabed-founded structures. Offshore developments, in water depths of 500 m to 1500 m, are now relatively commonplace and a number of projects are underway in over 2000 m of water. Within five years it is expected that development will extend beyond 3000 m of water. The costs of geotechnical operations in such water depths are high, especially in remote and harsh conditions. Consequently, despite the obvious technical advantages of early geotechnical data for supporting exploration and development, it is often difficult to justify spending money on such activities until there has been some exploration success. Engineers therefore rarely get the early geotechnical insight that is actually needed for frontier developments where there are no ‘offset’ data.

1 2 3 4 5 6 7 8

Water Depth Harsh environments Remoteness Geological Complexity Development Scale Data Acquisition Constraints Structural Complexity Inexperience and Uncertainty Each of these root causes is discussed in the remaining parts of Section 2.

2.3

In an unpublished review of literature on submarine geologies and environments Brunsden (2010) confirmed the complex morphological, geomorphological and geological legacies which geo-specialists and engineers need to understand and resolve to support offshore exploration and development. The main purpose of this review was to summarise conventional wisdom on landslide controls and processes. The main findings of the review were:

2.2 Water depth, environment and remoteness – (Root Causes 1 to 3) The first three root causes listed relate to access difficulties; geographical, subsea and subsurface. The logistical problems of working in frontier areas away from established centres of industry and support infrastructure are clear and will have a significant affect on the efficiency and cost of even routine offshore operations. Deep water and rough weather can restrict access further and cause additional problems for geophysical and geotechnical data acquisition and for the installation of foundations, anchors and other © 2011 by Taylor & Francis Group, LLC

Complex geology (Root cause 4)

– Offshore depositional centres are largely determined by plate tectonics and they are subject to repeated sedimentation and landslide processes.

4

Table 1.

Offshore geological and geotechnical controls.

Origin

Type

Cause

Effect/Factor

Natural

Inherited quasistatic conditions

Past geological & environmental processes

Preconditioning – structural relief/seabed ruggedness – slope geomorphology & geomorphology – sediment fabric and structure – sediment stress-strain history – sediment mechanical properties

Inherited transient Equilibration of conditions Aggravating processes arising from past geological & environmental processes or events Future processes Ongoing or new slow Aggravating geological and environmental processes Future episodic Random short-duration Triggering events natural phenomena Man-made Future processes or events

Episodic or progressive offshore activities

Triggering

– fluid flow – swelling – strain softening – physiochemical changes – sedimentation at slope crest – slope toe erosion – uplift from diapirism – earthquakes – storm waves and currents – tsunamis Project-life operations: Exploration, e.g. – exploration well drilling – rig anchoring – geotechnical investigation Construction/Installation, e.g. – development well drilling and anchoring – installation of gravity structures – pipe laying – pile driving Hydrocarbon Production, e.g. – operational loadings on facilities – fluid and gas injection – fluid withdrawal and subsidence

complex interactions between inherited characteristics and future processes, events or activities that may precondition, aggravate or trigger geohazard incidents that threaten projects. A good understanding of the geological past and project-life future is important since damaging geohazard events, such as submarine slides, may arise during the lifetime of a project due to ongoing geological processes alone, or in combination with shorter duration triggers. Geological history also has an important influence on the physical properties of marine sediments. Chandler (2000) and Cotecchia and Chandler (2000) describe how the mechanical behaviour of a clay is controlled by the ‘soil structure’ that it develops during deposition solely as a result of one-dimensional consolidation and by natural post-depositional geological processes such as mechanical unloading, creep, diagenesis and tectonic shearing. The following subsections discuss some of the sedimentological and post-sedimentological processes that can affect the physical behaviour of marine clays in areas such as Egypt and Angola. Soil Structure. A challenge for the offshore geotechnical engineer is to understand the physical characteristics of shallow marine sediments and how these may affect their behaviour when disturbed by natural events or by exploration and development activities. Marine sediments of most interest to engineers

– Terrestrial, hemipelagic, contourite and shelf sediment sources provide soils at variable rates and in relatively definable packages. – Structural controls are determined by plate tectonic history, fabric, structural relief, sedimentary and basin architectures, surface morphology and residual accumulated strains are the results of a chequered inheritance. – Mass movements are much larger than onshore but the seabed often exhibits extraordinary freshness that may reflect long periods of inactivity between episodic events and/or our inability to resolve slow sedimentological or geomorphological changes with current geophysical imaging technology. Leroueil (2001) describes how geological and geomorphological inheritances and future aggravating or triggering factors are the major controls for terrestrial and submarine landslides. Locat and Lee (2000) and Locat (2001) present similar ideas in connection with instabilities along ocean margins which can be equally applied to geohazards in general. The main natural and man-made controls on seabed instability and other geohazards in the context of offshore exploration and development are summarised on Table 1. A significant challenge for those responsible for planning and engineering offshore developments is to unravel the © 2011 by Taylor & Francis Group, LLC

Examples

5

in deep water are predominantly clays and Burland (1990), Chandler (2000) and Cotecchia and Chandler (2000) describe how such soils have significant geotechnical inheritances that condition their response under mechanical loading. Burland (1990) used the inherent compressibility and strength characteristics of reconstituted clays, called intrinsic properties, as a reference for interpreting the in situ state and physical characteristics of inorganic natural non-carbonate clays. He performed one dimensional compression tests on clay slurries and by plotting a normalising parameter called the Void Index , Iv, against the vertical effective stress σv
, derived an Intrinsic Compression Line (ICL) as a datum for the in situ stress states for natural clays. Iv is defined as: Figure 2. Normally consolidated clays with sedimentation structures – after Cotecchia and Chandler (2000).

deposited slowly in still water will have more random open structures with Iv values that may lie above the SCL, Point’C’. Highly flocculated clays deposited in seawater with higher than average salt concentrations and siliceous and carbonate oozes would also be expected to plot above the SCL. When natural clays are loaded by foundations in the field, or in one dimensional compression tests (oedometer tests) in the laboratory, the rate of loading is sufficiently high to disrupt the inter-particle bonding and fabric of the clays. The corresponding compression lines are relatively steep compared to the natural SCL and fall towards the ICL, also shown on Figure 2. It is therefore particularly important to detect and characterise clays with natural states that plot above the SCL since they are especially brittle and compressible and therefore more susceptible to progressive failure and possibly liquefaction under external loading. Post-sedimentation Effects. All naturally sedimented clays are altered to some degree by environmental, geological and physio-chemical processes, or other ageing effects. As described previously, many large-scale post-sedimentation changes can occur in submarine environments and these can be expected to alter the stress states and mechanical properties of sediments significantly. However, even in the most benign environments, soils will experience postsedimentation changes due to one or more of the following processes:

In this expression e0 is current in situ voids ratio, e∗100 is the voids ratio of a reconstituted sample of the same soil compressed one dimensionally to a vertical effective stress, σv
, of 100 kPa and C∗c is the compression index of the reconstituted soil measured between σv
of 100 kPa and 1000 kPa. The asterisk denotes an intrinsic property of a reconstituted soil prepared from a slurry mixed at a water contents of between 1.25 and 1.5 times the liquid limit. According to Chandler (2000), Iv is a useful normalising parameter for different soil types because C∗c is defined uniquely by the one-dimensional compression procedure, the soil mineralogy and the pore water chemistry. Burland also established an Iv compression line, for marine clays in their natural state, called the Sedimentation Compression Line (SCL). The distance between the SCL and ICL in Iv-σv
space is a measure of the sensitivity and brittleness of the natural clay. The SCL represents the average condition for natural marine clays with sensitivities of between about 2 and 9 and soils that plot on the idealised SCL have sensitivities of about 5 (Cotecchia and Chandler 2000). Sedimentation Structure. Mitchell (1976) defined soil ‘structure’ as a state that reflects the arrangement of soil particles, or soil fabric, and interparticle bonding. As shown on Figure 2, the ICL and SCL provide a simple and convenient framework in for comparing the in situ stress states or ‘structures’ and the mechanical behaviours of natural and reconstituted clays (Chandler 2000). The structure that clay has inherited during sedimentation depends on its mineralogy and the depositional conditions and it is not easily changed by further burial. The natural stress states of most marine clays lie on or around the SCL, Point ‘A’ in Figure 2. Clays that are deposited relatively rapidly from dense suspensions in high energy environments such as hyperpycnal flows, tend to have more orientated compact structures with Iv’s closer to the ICL, Point ‘B’. Debrites, contourites and turbidites may also have these characteristics. Conversely, hemipelagic clays that are © 2011 by Taylor & Francis Group, LLC

– Ageing – Creep./Secondary Consolidation – Unloading and swelling (Mechanical Overconsolidation) – Diagenesis – Physio-chemical changes including bonding and cementation, – Biological activity (Ehlers et al. 2005, Kuo et al. 2010) Such soils will exhibit gross yield, a threshold stress state beyond which soil stiffness and strength fall significantly (Hight et al. 1992, Cotecchia and Chandler 2000). Mechanical behaviour pre- and post-yield will be fundamentally different, so the identification of

6

Figure 4. Normally consolidated and mechanically over-consolidated clays with diagenetically altered structures – after Cotecchia and Chandler (2000).

Figure 3. ‘Aged’ normally consolidated and mechanically overconsolidated clays with sedimentation structures – after Cotecchia and Chandler (2000).

and slopes and (2) the interpretation of insitu tests that are used to infer these mechanical properties (Randolph et al. 2005). In addition soils with natural stress states close to or above the SCL and/or with significant post-sedimentation structures that do not fit well in to normalised stress history frameworks such as SHANSEP that require consolidation beyond gross yield (Ladd & Foot 1974). Burland (1990) indicates that the peak strength and brittleness of such soils are underestimated using the SHANSEP test procedure. The detection, identification and characterisation of atypically or unusually sensitive soils are therefore important in offshore geotechnics. The natural soil structure framework described by researchers like Burland (1990), Cotecchia and Chandler (2000) and De Gennnaro et al. (2005) can be used with other soil frameworks such as Critical State Soil Mechanics to predict the mechanical behaviour of natural soils affected, for example, by ageing, desiccation, ice-loading, erosion and mass wasting. However, BP and other offshore operators are increasingly encountering post-sedimentation processes that are more difficult to characterise using traditional methods and which may require different, possibly bespoke, soil mechanics approaches, as suggested by Gens (2010). Four examples of effects that have posed additional challenges during BP’s exploration and development activities are:

gross yield stress states in shear and compression is also important for modelling soils correctly in offshore foundation design and geohazard assessments. Aged normally consolidated clays and clays that are over-consolidated due to mechanical unloading but which are unaffected by diagenesis or biological activity, would be expected to yield close to their natural sedimentation compression lines; the SCL for most natural marine clays. The oedometer com
pression lines and vertical yield stresses, σvy , for an aged normally consolidated clay and a mechanically over-consolidated clay are shown schematically on Figure 3. The vertical yield stress for an over
consolidated soil, σvy , will be coincident with the
. geological preconsolidation stress of the soil, σvc The ratio of geological preconsolidation stress and
/σv
is traditionthe in situ vertical effective stress σvc ally termed the over-consolidation ratio (OCR). The ratio of the yield stress and in situ vertical effective
stress σvy / σv
is the called the yield stress ratio, YSR. Conversely, as shown on Figure 4, many normally consolidated clays will have been affected by diagenesis and will have diagenetically altered postsedimentation structures that will result in yield to the right of the SCL. Some over-consolidated soils that have retained diagenetic characteristics will also yield to the right of the SCL above the preconsolidation pres
sure, σvc . In this case the YSR is greater than the OCR. Oedometer compression curves for normally consolidated and over-consolidated clays with diagenetically altered structures are shown schematically on Figure 4. The natural in situ states of soils condition their responses to mechanical loading and disturbance and are therefore important geotechnical controls. Grossyield in compression and shear and the post-yield brittleness are important characteristics in offshore geotechnical engineering since they affect: (1) in situ operational shear strengths and stiffnesses of soils, and consequently the behaviour of foundations, anchors © 2011 by Taylor & Francis Group, LLC

– – – –

Shallow Gas Gas Hydrates Salt Diapirism Hydrocarbon Migration Products

Shallow Gas. Gas-charged sediments, largely of biogenic (bacterial activity) and petrogenic (thermally altered) origins, are commonplace offshore. The presence of gas, in the form of free bubbles, dissolved gas or gas hydrates can change the shear strength and compressibility characteristics of sediments under static and cyclic loading. Unlike unsaturated soils, gassy

7

soils encountered offshore contain large amount of dissolved gas that can exsolve. The degree of saturation of offshore gassy sediments is generally above 90% and free gas is generally in the form of discrete bubbles much larger than the pore spaces between individual soil particles (Sultan 2009). Shallow gas per se is a ubiquitous drilling hazard and shallow gas hazard assessments are obligatory for hydrocarbon exploration, development and production operations in most parts of the world. For example, in the GoM, the requirements for assessing shallow gas are specified by MMS (1998) and in the UK sector North Sea they are covered by guidelines prepared by UKOOA (1997). Gas also has important influences on the mechanical properties of sediments. These are less well understood but have potentially negative effects on offshore wells and structures too. Sands are especially vulnerable since the difference in pore water pressure and pore gas pressure is small, so high gas pressures equate to low effective stresses and reduced strength, as well as increased compressibility. However, the effects of gas on the physical properties of clays with low free gas concentrations, typically encountered offshore, have been the subject of limited research and are consequently less predictable. Research by Wheeler (1988) showed that the effects of gas on the mass undrained shear strength of clay depend on the initial effective and total stresses and the volume fraction of gas voids, and that these effects may be detrimental or beneficial. For example, Wheeler’s model suggests that the undrained shear strengths of gassy soils in triaxial compression at shallow depths in deep water, which have high initial total stresses and low initial effective stresses, may be less than the comparable strength of the saturated clay. Conversely, gassy clays with low initial total stresses and high initial effective stresses such as deep sediments in shallow water are estimated to be stronger than the saturated clay matrix. Both effects are predicted to increase with increasing gas saturation. However, Wheeler’s research was limited and it is clear that the mass strength of gassy clay depends, amongst other things, on the degree of soil de-structuring caused by the formation of free gas (Hight & Leroueil 2003, Sultan et al. 2009), and the direction and sense of loading. Wheeler’s soil model may provide useful upper and lower bound modifying factors for the effects of gas on undrained shear strength but the industry’s capabilities of inferring the operational properties of gassy soils from laboratory and in situ tests and for modelling their behaviour in routine geotechnical engineering are still very limited. Gas Hydrates. Gas hydrates are crystalline solids, physically resembling ice, that form when gas molecules are encaged by hydrogen-bonded water molecules. The main factors affecting hydrate formation and stability are temperature, pore pressure, gas chemistry and pore-water salinity and they can occur in the pore spaces of marine sediments under the right conditions.A change in any of these parameters such as a reduced pressure or an increased temperature could © 2011 by Taylor & Francis Group, LLC

Figure 5. Bottom simulating reflector (BSR).

affect the system equilibrium and reduce the stability of hydrates, resulting in decomposition, dissociation and dissolution, and the release of free gas and water (Sultan 2007). Many gases of low molecular weights can form gas hydrates but in marine conditions the gas is generally methane of biogenic origin. Marine gas hydrates are therefore common in organic-rich sediments at depocentres, especially where deposition has been rapid. They are therefore of relevance to most offshore hydrocarbon developments in relatively deep water where gas has infiltrated shallow sediments. Figure 5 is a seismic section from an offshore location which shows high acoustic amplitude traces below a bottom simulating reflector (BSR) that is indicative of possible gas-charged soils below a stable zone of gas hydrate-bearing sediments. A joint industry project (JIP) carried out on behalf of West African offshore operators (Sultan 2007), concluded that cementing action of gas hydrates could inhibit the normal compaction processes, resulting in sediments with more open structures and higher voids indices than would otherwise be the case. Hydratebearing sediments would therefore be expected to have higher yield stresses, elastic moduli and peak strengths, and to be more brittle, than comparable hydrate-free sediments. Sultan (2007) describes the potential physical collapse of such soils and the development of excess pore pressures, leading to loss of strength, increased compressibility and possible hydraulic fracture. The threats posed by gas hydrates to offshore exploration and production are still uncertain and speculative and views on this subject are sometimes controversial. Although problems have been reported (Nimblett et al. 2005), it is generally agreed that well drilling in hydrate-bearing sediments is possible with good planning and close control of well bore pressures and temperatures. However, experience of hydrocarbon production in hydrate-prone areas is limited so there is more uncertainty about interactions arising from longer-term offshore operations, for example raised temperatures around hot production wells. Therefore, for the present, the industry tends to avoid producing in areas where there is direct evidence or strong indirect evidence of hydrate-bearing sediments.

8

Figure 6. Salt dome.

soil overlying salt. These observations were based on the results of preliminary large strain numerical analyses performed by Imperial College/GCG (2007, unpublished) and reflect conditions often encountered offshore in regions such as Angola. Further generic and project-specific research work is clearly needed to support developments in areas where salt effects are unavoidable. Hydrocarbon Migration Products. Hill et al. (2010a, 2010b) describe areas of persistent hydrocarbon seepage in deep water Angola where the shallow predominantly clayey sediments contain beds/layers with anomalously high acoustic impedances, Figure 8a. These seismically reflective zones often have a surface expression and locally extrude above the seabed, as shown on Figures 8b and 8c. Investigations have shown that these atypical soils comprise sequences of hard carbonate-rich claystone beds with variable amounts of heavy hydrocarbons that have the consistency of thick oil and asphalt (bitumen) at room

Salt Diapirism. Hill et al. (2010a, 2010b) describe how salt diapirism in deep water Angola has uplifted older sediments which has resulted in strong soils close to the seabed that can cause installation problems for suction-installed foundations and anchors and increased seabed slopes that can lead to pipeline routing difficulties and instabilities. The seismic section and schematic shown in Figure 6 indicate how the overburden can be severely distorted by salt movement resulting in large changes to the depositional stress states of cover sediments. These changes may affect the behaviour of foundations such as suction-installed caissons that are designed using empirical methods developed from field tests in areas well away from such influences, or from centrifuge or physical models that replicate these far-field conditions. This point is illustrated in Figure 7, which summarises some of the observations made about the possible combined effects of uplift and subsequent erosion on a normally consolidated © 2011 by Taylor & Francis Group, LLC

9

Figure 7. Effects of salt uplift and surface erosion.

temperature when recovered in geotechnical samples, Figure 8d. Sediments that have been modified by hydrocarbon seepage over geological timescales are, unsurprisingly, relatively common in areas of interest for oil and gas exploration and development. These soils are inevitably highly variable and are difficult to characterise geotechnically. Hard beds arising from authigenic mineral cementation can clearly pose problems for foundation and anchor installations that will probably warrant location-specific geotechnical investigations and bespoke engineering solutions. Sediments contaminated with hydrocarbons may pose even greater engineering challenges and arguably open up a new branch of soil mechanics that requires unconventional engineering thinking and new geotechnical solutions, as suggested by Randolph et al. (2005) and Gens (2010). 2.4

Figure 8. Hydrocarbon migration products. © 2011 by Taylor & Francis Group, LLC

Development scale (Root Cause 5)

Offshore oil and gas developments range from discrete bottom-founded shallow water structures with relatively small seabed footprints at locations that are generally fixed early in the development cycle, to floating or subsea-tieback developments in deepwater that cover large areas of the seafloor and which are constantly evolving. For example, the total footprint area of the three subsea developments currently planned by BP in the WND amounts to about 1400 km2 . The footprint of a typical BP Angolan project is about 750 km2 . Scale is therefore a major challenge for deepwater geospecialists who are required to define the geotechnical conditions and geohazards for a large number of structures spread across the seabed with sufficient flexibility to guide, or accommodate, inevitable changes in scheme layout. For logistical and economic reasons, it may not be possible or practical to acquire geotechnical data at every structure location and it is often necessary to infer local design conditions from regional geotechnical data; a methodology described in some construction industry standards such as Eurocode 7 (2004). Eurocode 7 classifies structures in to three geotechnical risk categories, according to structural

10

parameter selection and geotechnical design reliability assessments.

complexity, the ground conditions, type of load, and the consequences of failure. Small and relatively simple low-risk structures constructed on uniform ground are designated Category 1. Large or conventional structures which are supported on unexceptional ground and have well defined loads and normal risks fall in to Category 2. Structures that are unusually large or complex, or which are associated with abnormal loads, ground conditions or risks are classified as Category 3. Eurocode 7 advice is that it may be acceptable to use regional (offset) geotechnical data for the detailed design of Category 1 structures but that location-specific data are needed for Categories 2 and 3. Many offshore structures fall in to Categories 2 and 3 but for the reasons alluded to earlier it may be overly constraining or uneconomic to apply Eurocode 7 advice rigidly in deep water. However, the concept may be extended by improving the reliability of the regional database through informed ground modeling and applying risk-based design to manage the residual geotechnical uncertainties. Probabilistic-based geotechnical design methods using offset geotechnical data have been discussed and presented by various authors including Gilbert and Gambino (1999), Clukey et al (2000), Jeanjean et al. (2005) and Griffiths et al. (2009). These methods generally use normal or log normal soil property distributions that are best suited for areas with relatively uniform geology and soils that have similar sedimentation and post-sedimentation histories. They are less suitable for use in areas with complex geologies or numerous geohazards. The approach may also be unconservative in uniform conditions if mechanical behaviour of the soil mass is dominated by local weak or brittle seams, layers or zones that have gone undetected. The SHANSEP normalization procedure is often used to infer the undrained shear strengths of clays of similar origins and compositions over large areas. The method is proposed in ISO 19901-4 (2003) to achieve consistent shear strength profiles and has been used successfully for geotechnical engineering and geohazard assessments in Norway (Kvalstad 2007) and in the Gulf of Mexico (Jeanjean et al. 2005). Burland (1990) and Chandler (2000) indicate that the procedure may be reliable for relatively insensitive clays that lie below the SCL. However, for the reasons described earlier and discussed by Le et al. (2008), they propose that the procedure may be unreliable for more structured clays. Jardine et al. (2005) also suggest that the procedure may be unconservative for plastic clays that develop brittle shear bands that suppress dilation, Reconsolidation to the in situ effective stresses as performed in CAU tests results in less sample compression than SHANSEP. However, this technique may also cause some ‘destructuring’ of some highly structured clays resulting in underestimates of undrained shear strengths and sensitivities. It is therefore important to recognize the significance of soil structure when preparing or ‘conditioning’ laboratory test samples and to consider factoring this into geotechnical © 2011 by Taylor & Francis Group, LLC

2.5

Data acquisition constraints (Root Cause 6)

There are no hard and fast accepted strategies for geophysical and geotechnical data acquisition in deep water. Rather acquisition plans are invariably a compromise between budget and technical requirements, and are often constrained by an industry-wide shortage of survey vessels, equipment and know-how to carry out the work. Data Acquisition Costs. Shallow geological and geotechnical characterization of the shallow subsurface, including geohazards, is generally achieved by a combination of geophysical surveying and intrusive geotechnical investigation. The precise scope of work is defined by the area, line density and resolution of the geophysical survey and by the numbers and locations of in situ tests and samples. A range of acquisition strategies is possible as illustrated on Table 2. One extreme, Strategy 1, would be to spend very little on location-specific data until the scheme layout has been finalised and to rely on regional or offset data for planning and design during the early stages of the project. The other extreme would be to carry out fully calibrated area-wide high density ultra high resolution surveys at the beginning of a project as described for Strategy 3. The delayed approach is most likely to be suitable for single projects in mature areas with uniform well understood geology and few geohazards, whereas multiple projects in geologically complex frontier areas with numerous geohazards may justify the significant front-end investment implied by Strategy 3. A phased approach somewhere between these two extremes is generally adequate for most deep water projects. The amount of money spent on shallow subsurface investigations should ideally reflect positive and negative risks, i.e. the opportunities or benefits of investing the right amount wisely and the threats from not spending enough, or spending it wastefully. The UK Institution of Civil Engineers UK indicate that the largest element of technical and financial risk on construction projects is normally in the ground (ICE 1991), and that the costs overruns and delays arising from inadequate geotechnical investigations can effect 18 to 50% of construction projects. For underground structures with significant soil interfaces such as tunnels, up to 85% of projects could be affected (USNCTT 1984). The amount spent on geotechnical investigations for onshore construction projects in the UK is typically between 1% and 5% of the capital cost but geotechnical cost overruns can be up to 50% of the project value (AGS 2003). There are few comparable statistics for offshore projects but industry surveys have shown that offshore geotechnical risks may be underestimated and that associated installation problems arise frequently in delays and cost overruns (Clayton & Power 2002). There is also little published about the costs of offshore

11

Table 2.

Data acquisition strategies.

Mature Areas/ Less Geohazards ⇔ Frontier Areas/ More Geohazards Increasing Front-end Uncertainty ⇔ Increasing Front-end Cost Purpose

Strategy 1

Strategy 2

For Concept Selection

Use exploration seismic data, Medium-density grid of 2D AUV & offset data & regional 2D UHR survey lines calibrated knowledge with geotechnical cores, boreholes & insitu tests

For Preliminary Medium to high-density design (FEED) AUV & UHR surveys with selective geotechnical calibration For Detailed Targeted geotechnical survey Design when development fixed

Strategy 3 High-density 2D AUV & 3D UHR blanket survey calibrated with geotechnical cores, boreholes & insitu tests

Additional selective ‘infill’ AUV & UHR surveys and geotechnical calibration

Possibly no additional acquisition

Selective location-specific geotechnical investigations

Selective location-specific geotechnical investigations

Table 3. The cost of data acquisition – Influencing Factors. 0.1% Capex ⇐

⇒ 1.5% Capex

– Shallow water – Mature areas – Simple geology – Few geohazards – Uniform soil conditions – Compact developments

– Deep and ultra-deep water – Frontier areas – Complex geology – Many geohazards – Highly variable soils – Laterally extensive developments – New unproven technologies

– Routine proven offshore technologies – Limited well understood – Many poorly understood soil-structure interactions soil-structure interactions – Financial risk only – Societal, Environmental and Financial Risks – 1st Party risks only – 3rd and 1st Party Risks Figure 9. Optimum spend on data acquisition.

10% loss would be 10−4 . The probabilities of a 10% loss for intermediate spends would be between 10−4 and 1.0. No attempt has been made to discount the costs in Figure 9 and consequential losses such as reduced production are not considered. Available Resources. The offshore geotechnical industry’s capabilities and resources for supporting the relatively rapid advance of hydrocarbon exploration and development in to deep water are sorely stretched. One of the main constraints for deepwater operators is the limited availability of shallow geophysical survey and geotechnical investigation vessels that can operate in these conditions and the dominance of one or two main contractors in these markets. The development of survey class autonomous underwater vehicles (AUV) (Bingham et al. 2002), has significantly improved the offshore industry’s capabilities for acquiring engineering quality ultra high resolution data (bathymetry, sidescan sonar and subbottom profiling) in deep water. However, at present only three companies offer AUV surveying services on a commercial basis in water depths greater than

geophysical and geotechnical surveys in the oil and gas industry. From BP’s experience these may range from as little as 0.1% of the capital cost for simple fixed shallow water platforms in uniform predictable soils, to 1.5% for large deep water projects in geotechnically difficult areas. Although there are no strict rules about the level of spend on site characterization, it is possible to identify those factors that may influence it, as shown on Table 3. The optimum spend on a project would be that which minimises the overall geotechnical cost, namely the front-end cost of the geophysical and geotechnical investigations plus the risked cost of geotechnical construction problems. A purely hypothetical example for a large deepwater project in a geologically complex area is shown on Figure 9. The optimal spend on investigations for this illustrative case is 1.5% of capital cost (Capex). This is based on the simple premise that: (1) if no investigations are performed the equivalent financial loss would be 10% of Capex and this would occur with certainty (probability of 1.0) during the project life and (2) at the other extreme, if 10% of Capex was spent on investigations the probability of a © 2011 by Taylor & Francis Group, LLC

12

about 500 m. Geotechnical investigation in deep water is still a significant technical challenge, time consuming and expensive, and requires significant investment by contractors. Only one contractor, Fugro, currently provides a full range of geotechnical services in deep water. The world’s fleet of deep water geotechnical vessels is also severely limited with Fugro operating the only three dedicated geotechnical drillships capable of working in water depths over about 1500 m. This shortage can mean that some deepwater basins may not have access to such vessels for years; a situation that is likely to worsen as exploration and production moves in to even deeper water and the technical, operational and safety challenges with conventional geotechnical drillships increase and acquisition costs rise. The market capability for geotechnical laboratory testing, engineering and reporting is slightly better but the numbers of companies offering independently accredited laboratories is still very small and the pool of deepwater geotechnical engineering specialists is tiny compared to that available to clients in the onshore construction business. For operators like BP, the lack of competition and resources reduces programming flexibility and often leads to geotechnical acquisition as a critical path activity. Sole-source generally means additional costs to the operator and in busy periods, poses extra burdens on the sole-provider with greater risk of log -jams and delays in post-field activities such as laboratory testing and reporting. The use of consortia as an alternative to a ‘one-stop’ service can work well but involves the management of a number of interfaces that can pose extra contractual, technical and HSSE pressures on operators’geotechnical staff. In the long term the deepwater oil and gas business would clearly benefit from the emergence of new geotechnical site investigation contractors and the use of quicker and remotely-operated investigative tools and equipment. However, these would require significant investments by contractors that ultimately can only be justified by market forces and a sustainable industry workload. Nevertheless, there has been encouraging progress in recent years; for instance through the use of Fugro’s SMARTPIPE® and SMARTSURF™ Systems (Hill & Jacob 2008; Denis & De Brier 2010), and breakthroughs made by Benthic Geotech Pty with their multi-functional seabed drilling system, PROD (Tjelta 2010). These developments have been achieved with the encouragement and support of one or more operators. Guidelines and Standards. Currently, there is a lack of international guidelines and standards for the performance of marine geophysical and geotechnical investigations, including vessel and equipment audits and best practices for data acquisition, laboratory testing and reporting. As a consequence operators often produce diverse job specifications that may have led to global inconsistencies and inefficiency, especially in the geotechnical investigation contracting business. The lack of best-practice industry guidelines has also hampered the integration of the two branches of © 2011 by Taylor & Francis Group, LLC

Figure 10. Subsea structures.

geoscience that underpin deep water site characterization; geophysics and geotechnics. Advice on offshore site investigation practices has been published by some national bodies Norsok (2004), or specialist groups such as SUT (2003 & 2004) and ISSMGE (2005), but these efforts are largely uncoordinated and are generally focused on shallow water. This industry standards deficiency is currently being addressed by the joint API RG7/ISO and SC7/WG10 committee for offshore foundations with the aim of publishing two new standards; ISO 19901-8-1, ‘Marine Soil Investigations’ and ISO 19901-8-2 ‘Geophysical Site Investigations within the next few years (Jeanjean et al. 2010). Ultimately, it is hoped that a third standard will be published that will provide guidance and encouragement for a multidisciplinary approach to shallow geohazard risk assessments and engineering for offshore developments. 2.6

Structural complexity (Root Cause 7)

Offshore Facilities. Randolph et al. (2005) provide a comprehensive review of geotechnical design practices for offshore facilities and how these have diverged from onshore practices because of the nature of marine soils, and the sizes, loads and performance characteristics of offshore foundations. As explained by Randolph et al. exploration and development in frontier regions has meant engineering in soils that are significantly different to those encountered onshore and which require new geotechnical design approaches. BP and others have reported many cases of unusual soil conditions, including extremely high plasticity structured clays in Angola (Gennaro et al. 2005; Le et al. 2010; Hill et al. 2010b), highly saline clays in the Caspian Sea (Kay et al. 2005), extremely dense sands in Norway (Alm et al. 2004) and exceptionally hard clays West of Shetlands (Evans et al. 2010). Structure Types. Figure 10 shows some typical subsea structures, ranging from relatively light surfacelaid in-field flowlines to large manifolds weighing

13

several hundred tonnes. The foundations for the heavy structures can be much larger than those built onshore. They are also required to resist much larger lateral loads and overturning moments and are often subjected to combined loads in all six degrees of freedom, including significant torsion. Large Strain Behaviour. Operationally subsea structures can generally tolerate relatively large displacements, although settlement may be a controlling factor for heavy structures such as manifolds. Many seabed interactions also involve repeated loads that can result in large cumulative strains and failures. A good understanding of large strain soil behaviour under cyclic and monotonic loading and the potential for progressive failure is therefore important for design reliability. In addition, the ability to model such behaviour accurately increases the potential for cost savings through the use of performance-based design methods that are described later. Low Stress Behaviour. Conversely, knowledge of the compressibilities and shear strengths of soils at very low stresses is also important for deep sea projects. For example, the operational normal total stresses applied by unburied in-field flowlines are generally in the range of 1 to 10 kPa, which is significantly below the range considered in onshore and shallow water engineering. The study of ‘low stress soil mechanics’ problems is difficult using conventional laboratory soils testing machines and often requires bespoke equipment to produce reliable results (Bruton et al. 2007). Uniqueness. Some deepwater structures have interactions with soil that are distinctive by having few onshore analogues (Evans et al. 2007). Examples of behaviour that are difficult to predict and continually pose engineering challenges include: (1) lateral buckling, walking and vibration of surface-laid pipelines that transport multiphase products under high pressures and temperatures, often on undulating seabeds (Figure 11a), (2) the random motions of compliant steel catenary risers (SCRs) that dangle on the seabed (Figure 11b) and flow-induced vibrations of bottom-founded flexible spools (Figure 11c). Design Codes. The background to present-day offshore foundation design practices is discussed by Jeanjean et al. (2010). Foundation design procedures in offshore codes and recommended practices are still largely based on the traditional bearing capacity methods that have been developed for onshore foundations; ISO 19904-1 (2003), API RP2A, (2000) and DNV (1992). These methods generally give safe designs for relatively simple foundations under fully undrained or fully drained conditions but provide little advice about designs under partially drained conditions or with torsional loads, and do not cover structures with more complex load-displacement characteristics such as described earlier. New Methods. The development of new reliable user-friendly geotechnical solutions for offshore engineering is proving difficult. According to Randolph et al. (2005), there are only a few theoretically exact © 2011 by Taylor & Francis Group, LLC

Figure 11. Soil-structure interactions.

plastic solutions and limit equilibrium methods are generally unconservative. Numerical methods based on finite elements have been shown to be useful for improving fundamental understanding of soil structure interactions but are difficult to adapt for routine design and, according to Randolph et al. (2005), may also be unconservative in some cases. ISO 19901-4 recommends the use of large scale load tests, model tests and field instrumentation to reduce residual uncertainties with foundation behaviour. Geotechnical engineering in geohazard-prone areas is potentially even more challenging due to more complex geology and greater natural soil variability that increases the ‘aleatory’ design uncertainties, Jeanjean et al. (2005) and possible interactions arising from active geological processes such as submarine slides. The goal for offshore operators such as BP is to develop reliable, cost-effective and robust, preferably codified, design methods that can be used with confidence by all their design and installation contractors. However, at the present time some design procedures are still under development, often through proprietary research, with very little field validation. The theoretical and empirical bases and expertise behind these methods is often exclusive to academic institutions or specialists contractors and some of the design

14

Table 4. Industry and International Guidelines – Coverage of Foundation Design and Geohazards – after Jeanjean (2010).

solutions are not readily assimilated into standard structural computer codes. Well Geotechnics. The commentaries above relate to offshore facilities but years of experience suggest that the geotechnical engineering of wells also needs serious attention. The top section of a well, or tophole, is drilled through soils and poorly lithified weak rocks. The depth of the tophole is typically between 500 m and 1500 m and it is drilled without pressure containment. Geotechnical problems are often encountered within well topholes during well conductor installations and drilling. These include wellbore instability and hydraulic fracture and drilling fluid loss in to the shallow formation. Drilling difficulties in the shallow section cause costly delays and may require expensive remedial measures. Uncontrolled shallow well drilling can also affect structures in proximity, as discussed in ISO19901-4 (2003), ISO 19902 (2007), HSE (1997) and by Hobbs and Senner (1998). Tophole problems can arise from a combination of difficult subsurface conditions and poor drilling practices. Despite many years of experience they still occur (Schroeder 2007), and the situation will only improve through increased collaboration between the offshore well drilling and geotechnical communities. Some loss-prevention initiatives have already been taken in BP with the geotechnical and drilling groups jointly developing internal best practices for tophole drilling and completions. At industry level, the Offshore Soil Investigation Forum (OSIF) and the SUT’s Offshore Site Investigation Group (OSIG) are currently preparing a guidance document on this subject.

Geohazardrelated text length

Foundationrelated text length

API RP2A ISO 19901-4 ISO 19902 API RP2T API RP2SK ISO 19901-7 ISO 19904-1 ISO19905-1

225p 34p 642p 142p 181p 130p 194p 288p

1p (0.5%) 2p (6%) 0p (0%) 0.2p (0.2%) 0.1p (0.05%) 0.1p (0.07%) 0.1p (0.05%) 2p (0.7%)

36p (16%) 34p (100%) 42p (6.5%) 6p (4.2%) 38p (21%) 10p (8%) 0.1p (0.05%) 46p (16%)

1%

20%

Average

validated. However, it is likely that these will still be limited to relatively simple structure-soil combinations. The lack of industry guidelines on geotechnical engineering in geohazardous areas is even starker. This is illustrated qualitatively on Table 4, Jeanjean (2010, unpublished), which shows that current API Practices and ISOs give little textual coverage to geohazards. This is clearly something that needs to be redressed but it is unlikely that any future code or standard will be able to give anything but general advice for engineering in extremely variable conditions. Engineering in such areas will therefore continue to call for geotechnical rigour. Contractual Responsibilities. The responsibility for geotechnical and geohazard risks should ideally be with those that are best able to manage them. At the present time, operators are probably in the best positions to manage geohazards since they employ the necessary specialists and have more time to investigate them and assess their impact. Similarly, experienced offshore design and installation contractors are better placed to manage geotechnical risks. However, the responsibility for geotechnical risks is not always clear, even in so-called non-geohazardous areas. For example, unproven, often bespoke, designs naturally carry greater than normal construction and operational risks for operators and their contractors. These risks increase for EPC contractors when they are required to adopt or adapt research-based design methods that are the intellectual properties of operators or niche specialists. In the short term, these additional risks are likely to be reflected in bid prices and in the numbers of contractual exclusions and claims but in time it is hoped that the new methods will be tested and calibrated by field monitoring and will evolve in to more reliable and routine solutions that attract less uncertainty and risks. In the meantime, the challenge for offshore operators, their specialist advisors and their contractors is to find ways of working together to manage the transition from evolving best practices to established design methods, while protecting the interests of all the stakeholders.

2.7 Inexperience and uncertainty (Root Cause 8) One of the biggest issues currently faced by offshore industry in general is working in ever deeper water in new geographical areas and with new technology. The geotechnical discipline is no different and is on a fast-track learning programme that is most evident at the delivery end of projects, namely in engineering procurement and installation. As described earlier, the ground conditions are a major source of uncertainty on all construction projects and are a common cause of delays and cost overruns. Although, great advances have been made in offshore construction there are still a lot of challenges associated with the engineering and installation subsea structures in deep water, particularly in geohazard-prone areas. Codes and Guidelines for Geohazards and Foundations. The relative lack of engineering experience is also evident by the shortage of guidelines, standards and recommended practices for the design of deep water structures. Jeanjean et al. (2005) suggested that the paucity of formal guidelines reflects the newness and uniqueness of many deepwater structures and the natural time lag required for field validation of new design practices before code acceptance. The implication is that best practices and new code-based procedures will emerge, as methods mature and are © 2011 by Taylor & Francis Group, LLC

Standard

Total length

15

As the offshore industry moves in to areas with increasing geological complexities and uncertainties the roles and responsibilities for geotechnical engineering and geotechnical risks need to be clear and equitable. In this respect, the industry may benefit by adapting the risk management practices developed in the underground construction industry where projects invariably have significant geotechnical exposure. The tunneling industry has over many decades developed best practices that centre on the production of Geotechnical Baseline Reports (GBRs) that define the interpreted or assumed reference geotechnical conditions and risks at the beginning of the contract (Essex 2007).The main purpose of GBRs is that they provide a set of geotechnical conditions against which the actual conditions encountered can be assessed and compared and therefore provide clarity on risk sharing. 2.8

Communication of geotechnical risks and costs

Figure 12. Benchmarking risk severity and manageability.

For the reasons discussed above, shallow subsurface risks in deepwater frontier areas, and therefore the time and costs of investigating and mitigating them, are often much greater than for traditional shallow water projects. Possibly the biggest challenge of all for shallow geospecialists is therefore to convey clear matter-of-fact messages about these new challenges to non-specialists who may lack this insight. The ultimate objective would be to provide timely fit-forpurpose geotechnical guidance that would allow the project team to understand and manage the geotechnical elements of the development efficiently and systematically, ideally as non-critical activities. Table 1 summarises the main natural and manmade controls over shallow subsurface conditions for offshore oil and gas developments. The main hazards are natural variable and/or weak soils and dormant or active geological features and processes, and the field activities that may aggravate these conditions. These subsurface conditions may pose significant threats to the integrity and safety of offshore developments during and after installation and construction. Equally, there are opportunities and strategies by which these threats can be managed and mitigated. It is important that all these negative and positive factors are communicated and discussed with the various stakeholders in the right ways and at the right times, so that they are interpreted correctly for inclusion in risk registers and are used effectively in budget planning, design and construction. Communication of geotechnical and geological risks (threats and opportunities) can be a major asset to the project when done properly but if done poorly, or at the wrong time, may be misleading and counterproductive. At BP advice on the shallow subsurface is generally tailored to meet the needs of six distinct project stages: Access: Geologic appraisal and prospect evaluation Appraise: Project feasibility Select: Choose the preferred project(s) Define: Develop the preferred project and fund Execute: Detailed design and build Operate: Production

Access. When considering acquisitions, investors and financial planners are most concerned in asset ‘prospectivity’ and value and the shallow subsurface has little influence on investment decisions. However, interest increases once access has been gained, especially with respect to the risks and costs associated with drilling exploration wells. The production of shortterm shallow drilling hazard risk assessments is the major activity for shallow geospecialists at this time. However, longer-term risks may also need to assessed even at this very early stage should exploration or appraisal wells be considered for use in production, ie so-called ‘Keeper’ wells. Appraise. Following exploration success the information of immediate use to project planners and managers is an assessment of the main threats to development, benchmarked against other projects, and the levels of investment that may be needed to manage these threats. A simple way of benchmarking levels of risk and complexity/manageability is illustrated on Figure 12 (HSE 2006). This figure also provides guidance on specific methodologies might be needed to assess risk, including Qualitative (Q), Semiquantitative and Quantified risk assessment (QRA) methods. Early Appraise is also the stage at which the data acquisition and risk assessment strategy should be discussed and agreed with project funders. Ultimately, at the end of the Appraise stage the shallow geo-specialists need to assure the project that there are no showstoppers and that there are unlikely to be major geotechnical surprises; in effect a ‘declaration of geotechnical feasibility’. Select and Define. During the Select and Define stages the main role of geotechnical engineers and the rest of the shallow subsurface team, is to build partnerships with project planners and facilities and well engineers to help develop economic and reliable geotechnical designs and pragmatic risk reduction solutions. Risk mitigation can be achieved by avoidance through optimised layout planning or

© 2011 by Taylor & Francis Group, LLC

16

Execute. Offshore engineering and construction are usually, although not always, carried out through lump sum EPC contracts. Whatever, the type of contract it is important that roles and responsibilities for geotechnical engineering and the associated risks are fully understood by the operators and main contractors. Good geotechnical communication is needed between the operator and the main contractors and Geotechnical Baseline Reports supported by fit-for-purpose factual Geotechnical Data Reports (Essex 2007) may be the key to this; especially if the main contractors’ views can be incorporated in the GBRs before contract award. Operate. The geotechnical conditions encountered during construction are often different to those assumed beforehand. It is important that these differences and any design changes associated with them are recorded and included in an updated as-built geotechnical report that could be used as a reference for future field expansion and maintenance, and for decommissioning. The preparation of as-built geotechnical reports is not an industry norm at present but would appear to be a natural extension of current best practices.

3

FACING THE CHALLENGES

3.1 Evolving practices The number of oil and gas facilities that have been built by BP and other operators in geohazardous and geotechnically difficult areas has increased significantly over the past 15 years. During this time BP has developed strategies and methods to maintain project reliability and cost-effectiveness in such environments. Many of these practices are basic engineering common sense while others have evolved to meet new challenges and opportunities. In their ISFOG 2005 keynote, Jeanjean et al. (2005) shared some of the best practices and methodologies that BP had developed for engineering in geohazardprone areas over the previous decade, with special reference to pioneering projects in the GoM. The authors championed the use of integrated multidisciplinary teams of geospecialists and phased geophysical and geotechnical surveys and supported the development of new enabling technology for these surveys. Amongst other things, they described (1) the use of shallow seismic data and the SHANSEP normalisation technique for inferring the undrained shear strengths of soils over large areas, (2) deterministic and probabilistic methods for assessing seabed stability, (3) model testing to calibrate non-code based geotechnical designs and (4) reliability-based geotechnical designs. The authors also recommended a consistent and systematic approach to geotechnical and geohazard risk assessment so that the geo-risks could be ranked alongside other project risks. No clear step changes have emerged in geotechnical and geohazard risk management in the past five

Figure 13. Risk assessment methods.

by geohazard-resistant design and economies can be made by exploring value engineering options such as risk based design, performance based design and observational engineering. Ideally, the relative costbenefits of the geohazard-optimised options would be assessed using probability based techniques that account for increased Capex and reduced Opex that may accrue compared to the ‘do nothing’ case. The level of residual risks should be tracked continually during this period and assessed using methodologies that are compatible with those used by the project for assessing risks from other sources. These may include Qualitative (Q), Semi-quantitative methods (SQ), in which frequency and severity/impact are quantified approximately within ranges and Quantified risk assessment (QRA) in which both are fully quantified. Examples of Q, SQ and QRA are shown in Figure 13. The objective at the end of Select is a technically feasible geohazard-tolerant field layout. The goal at the end of Define is an optimized reference scheme that is suitable for taking forward into Execute. © 2011 by Taylor & Francis Group, LLC

17

years and shallow subsurface conditions cotinue to pose difficulties for offshore exploration and development and many projects are still dealing with complex poorly defined conditions with no prescriptive solutions. However, BP’s experiences continue to show that an effective way of managing the risks and costs is through clear-ownership and teamwork, and through the early and continuous engagement of geospecialists with the BP project teams and with the contractors who plan, engineer and install offshore wells and facilities. In about 2004 prospects for multiple sequential projects emerged in Angola and Egypt that allowed BP to learn from GoM experiences and to take advantage of some important longer-term investment and learning opportunities for geotechnical engineering and geohazard risk management in these two areas. The next sections develop some of the geotechnical engineering best-practice themes described by Jeanjean et al. (2005) and describe how these have evolved and been adapted to meet new circumstances, lessons learnt and fresh challenges. Particular reference is made to the use of dedicated regional geoteams, regional ground modeling and possible methodologies for engineering facilities. 3.2

Figure 14. Centralised geotechnical and geohazard teams.

Figure 15. Shallow geospecialists – Project-life support.

Shallow geotechnical and geohazard assessment teams

Consultancy Group (GCG), Imperial College, URS Corp, Fugro William Lettis & Associates and AtkinsBoreas. The GAT approach is particularly cost-effective for multiple-project programmes by providing critical mass and know-how and the means to capture and transfer lessons learnt and develop standard design and risk assessment practices and engineering solutions. The co-location of GATs from different regions provides additional advantages in the form of economies of scale, cost-sharing, technology transfer and consistent and standardised corporate engineering practices.

Evans et al. (2007) describe how the industry-wide shortage of offshore shallow geohazard specialists, especially in engineering contracting, had motivated BP Exploration Technology Group (EPT) to set up dedicated multidisciplinary shallow geotechnical and geohazard assessment teams, GATs, to support a rolling programmes of offshore projects in Egypt and Angola. The GATs for these two regions are co-located with each other and close to their respective planning and engineering teams, as illustrated on Figure 14. The rationale behind this concept is that a GAT has the critical mass and competencies to provide sustained specialist support that is tailored to the project’s whole-life needs and activity levels as shown on Figure 15. The discipline bias of the GAT is geophysics and geology at the start of a project, gradually changing to geotechnical engineering as work progresses in to development planning and engineering. A key benefit of co-located GATs is that they encourage continuous dialogue between the geospecialists and BP’s project team and contractors so that they can manage the shallow subsurface engineering risks and costs together. BP’s Egypt and Angola GATs were established in 2004 and are presently supporting projects at various stages of development from Appraise to Execute. The teams are formed from a central core of shallow geophysicists, geologists, geomorphologists, sedimentologists and geotechnical engineers from within BP and from resident consultants Halcrow Group Limited and Fugro GeoConsulting. Additional external support is provided by other consultants and contractors, including the Norwegian Geotechnical Institute, D’Appolonia, Senergy, Geotechnical © 2011 by Taylor & Francis Group, LLC

3.3

Geotechnical and geohazard (G&G) mitigation strategy

The shallow geotechnical and geohazard (G&G) mitigation strategies BP is adopting for its rolling programmes of projects in Angola and Egypt are similar and in each case was developed early in the programme cycle following a desk-top geotechnical review or ‘Health Check’. The desk studies were performed using in-house or published regional and local geophysical and geotechnical (G&G) data and hydrocarbon exploration seismic quality geophysical data. Desk studies were the important first steps in the geotechnical and geohazard risk management process since they provided immediate insights in to the levels of risk, the resources and investment likely to be needed to manage these risks, and the best strategy for acquiring data to help this process. The results of the desk study were used to develop a fit-for-purpose geotechnical and geohazard risk management plan to investigate, characterize and mitigate geohazards and to provide geotechnical engineering support to the

18

Figure 17. Geophysical data acquisition equipment.

3 if necessary, to carry out supplementary geotechnical investigations to support detailed design and construction when the scheme layouts and architectures are defined better. Figure 16. Geotechnical and geohazard mitigation process.

The prospect of multiple projects allowed the Egypt and Angola Business Units to take a far-sighted view of surveys and make longer term arrangements with suppliers, as a hedge against the shortage of worldwide survey vessels. Integrated Surveys. As described earlier, deep water subsea tie-back schemes like those being developed by BP in the WND and others based on floating systems similar to those being built in Angola cover 100’s km2 of the seabed and huge volumes of soil. For cases like these, it is clearly not feasible or practical to investigate every seam, layer or stratum of soil or geological feature within the development footprints. Spatial interpolation is unavoidable and needs to be carried out systematically by experienced geospecialists to ensure the main geotechnical objectives are achieved at least cost. The integrated survey and ground model approaches favoured by BP in Angola and Egypt help this optimisation process by providing an initial regional view with gradual focus on areas or soils that ultimately have the greatest influence on project safety and costs. The geophysical data acquisition technologies used for imaging the seabed and shallow subsurface to provide the skeletal ground models in Egypt and Angola are shown in Figure 17, which is reproduced from Moore et al. (2007). The geotechnical acquisition priorities for a ground model strategy are to provide data for calibration of the shallow geophysics and characterisation of geohazards and to get representative geotechnical data for designing and installing wells and facilities. An important feature of BP’s optimisation strategy in Angola and Egypt is to achieve these goals with least effort and at lowest cost, by maximising the multi-functional site investigation activities, as shown on the Venn diagram in Figure 18. BP’s deepwater geotechnical investigation methods include the use of sampling and in situ testing in seabed and drilling modes. In the seabed mode BP frequently investigates the soils at a single location using a combination of sampling and testing tools, including

projects. The Geotechnical and Geohazard Mitigation (GGM) plans were based on the overall project plans, the inferred complexity of the shallow subsurface conditions and the perceived challenges these pose to developments. The prospect of multiple projects in both Egypt and Angola had a significant influence on the planning strategy and was instrumental in the decision to set up regional GATs and to adopt the regional ground model methodology. The key steps in the GGM process are shown in Figure 16. They comprise an integrated sequence of Data Acquisition, Ground Modeling, Risk Assessment and Layout Planning and Engineering designed to help the projects develop geohazard-tolerant and geotechnically robust reference schemes that can be taken forward in to EPC contracts for detailed design and construction. 3.4 Data acquisition Phasing. Although the actual data acquisition strategy and scope will depend on the nature and timing of the offshore development, it is good practice to take a staged approach. This involves phases of geohazard and geotechnical engineering quality geophysical surveying followed by geotechnical investigations to calibrate the geophysics and to provide geotechnical data for engineering. BP adopted acquisition campaigns based around Strategies 2 and 3 in Table 2 for subsea developments offshore Egypt and subsea and floating developments offshore Angola. The aim in each case was to acquire sufficient data to build up a development-wide predictive ground model. The main steps in this process were: 1 To carry out front-end development wide 2D ultra high resolution and Autonomous Underwater Vehicle (AUV) surveys, 2 to perform geotechnical reconnaissance investigations to fully calibrate the geophysics for preliminary design, and © 2011 by Taylor & Francis Group, LLC

19

Figure 18. Optimised geotechnical data acquisition.

Figure 19. Geotechnical investigations in seabed mode.

box cores, gravity cores, gravity piston cores, ‘T’bar testing and piezocone penetration tests (PCPTs) as illustrated on Figure 19. This combination, or ‘Set’, of tools is generally sufficient to fully characterise the shallow subsurface conditions for assessing shallow geohazards and for designing flowlines, pipelines, shallow foundations and anchors, including suction installed caissons. However, in recent years BP has also actively supported the development of other seabed in situ testing tools such as Fugro’s SMARTPIPE® and SMARTSURF™ systems (Figures 20a and 20b) to improve the understanding of soil conditions and pipe-soil interactions at or very close to the mudline (Denis & De Brier © 2011 by Taylor & Francis Group, LLC

Figure 20. New equipment.

2010). BP is also considering Benthic Geotech’s remotely drilling system Prod shown in Figure 20c, that has already been used by other operators such as Woodside, Total and Statoil (Tjelta 2010).

20

Figure 21. Ground model.

One innovation favoured by BP in Egypt is the acquisition of 3-m to 20-m long gravity piston cores exclusively for geological, sedimentological and ichnological logging and geochronological testing, as described by Thomas et al. (2010). The continuous geological cores were taken alongside geotechnical cores of similar length and proved to be a sound investment. The core logging significantly improved the WND GAT’s understanding of the types, frequencies and magnitude of historical submarine slides, and helped them optimise the geotechnical testing programmes for engineering shallow foundations and assessing the potential of future shallow slide activity. A key advantage of continuous core logging is that it can help to detect discontinuities and pre-existing shear surfaces, and possibly identify soils that may be especially susceptible to mechanical disturbance. Boreholes were used selectively in Angola and Egypt to perform in situ tests, and to provide samples for geological logging and geotechnical testing. The borehole data were used for calibrating regional seismostratigraphic models, exploring deeper geohazards, inferring in situ pore pressures and providing geotechnical data for engineering well topholes and deep foundations such as piles. The downhole work included high resolution geophysical logging, piezocone and temperature cone testing and piezorobe testing.

subsurface conditions anywhere within it, (2) to use the interpreted data for geohazard risk assessment, field layout planning and engineering at most locations and (3) limit site-specific geotechnical investigations to areas or structures that pose the biggest risks or offer the best value engineering opportunities to the project. The integrated geological modelling is discussed in ISO 19901-4 and the advantages of the approach are largely self evident. However, the success of the method is highly dependent on having sufficient data of the right quality to populate the model, a competent interpretation team and purposeful deliverables that are communicated clearly with end-users. The challenge is particularly onerous in geologically and geotechnically complex areas but the multidisciplinary GAT and phased data acquisition concepts described earlier have proven to be well suited to the task. The ground modelling methodology adopted by BP is illustrated on Figure 21. The procedure involves the integration of environmental, geophysical and geotechnical data and the detailed mapping, interpretation and calibration of these data to develop a seismostratigraphic-soil model of the area of interest, and to capture this in GIS-format for interfacing with the project team and its engineering contractors. The main outputs from the regional model are shown on Figure 21 and in an idealised flow diagram on Figure 22. These may be summarised as:

3.5

– Terrain Units. These are developed from the geomorphological interpretation of Multibeam Echo Sounder, Side Scan Sonar, 2D UHR and AUV geophysical data. Each Terrain Unit has a distinctive geological setting and depositional history with similar geological forms and processes. – Geohazards. Terrain units that are associated with past geohazard events such as submarine slides, and therefore require more detailed assessments, are generally distinguished from those that have

Ground modelling

The predictive ground model approach that BP is adopting for its deep water projects offshore Egypt and Angola is centered on the creation of a 3-D block model that captures the geomorphology, main stratigraphic units, geological features, geohazards and representative geotechnical conditions across the development footprint. The main objectives of the method are to: (1) build a model that is reliable enough to infer the shallow © 2011 by Taylor & Francis Group, LLC

21

Figure 22. Ground model outputs for geohazard risk assessment and engineering.

sediment types, one layered and the other more disturbed, are shown Figure 23(b). The hemipelagites (Set 1) have high Iv values that are at, or above, the SCL and they are relatively compressible. Conversely, Set 2, interpreted from geological evidence as mass flow deposits, have Iv’s close to the ICL and they are less compressible in the stress ranges of interest for most offshore foundations. These findings are consistent with the inferred geological stress histories. However, a second group of soils, Set 3, also interpreted as mass flow sediments, have higher Iv values closer to the SCL. This suggests that these soils may have retained some of their sedimentation structures during deformation. Alternatively the facies interpretation may be imprecise and may be need to be reconsidered and refined. Figures 24 shows similar plots for recently deposited shallow sediments from two regions in deepwater Angola, Regions 1 and 2. The data envelope shown in Figure 24(a) is for oedometer tests performed on samples of hemipelagic soils from Region 1. These soils are interpreted to be unaffected by geological processes such as salt diapirism and fluid expulsion and the test data broadly straddle the SCL. Figure 24(b) shows the results of oedometer tests on four samples of shallow soils from Region 2, in an area influenced by salt and erosion. These results suggest that the effects of uplift and erosion may include ‘destructuring’ and overconsolidation as implied in Figure 7. The results of the oedometer tests in Figures 23 and 24 clearly show the potential insights in to the stress histories and mechanical properties of marine soils that may be gained from using frameworks such as proposed by Chandler (2000). These frameworks offer the means to calibrate terrain units

more benign origins. These geohazard terrain units are subject to further detailed sedimentological, ichnological and geochronological assessment and geomechanical back analyses to estimate the frequency and magnitude of these historical events (Thomas et al. 2010). The results of this work are fed in to the future geohazard risk assessment. – Soil Provinces. Soil Provinces are linked closely to Terrain Units and are idealised areas of the seabed and shallow subsurface that have broadly similar geotechnical characteristics and properties to depths of interest for engineering. Each Terrain Unit may comprise one or more Soil Provinces and some Soil Provinces may be common to a number of Terrain Units. – Soil Units. These are the main soil layers and strata within each Soil Province that are interpreted to have broadly similar mechanical properties. Some Soil Units are common to more than one Soil Province and they are interpreted by dividing the ‘geophysical’ column into significant seismostratigraphic units separated by regional unconformities and defining the sediments that make up these units as one or more distinct soil facies. The soil facies represent different depositional and post-depositional processes and histories and they are interpreted from sedimentological, lithological, ichnological and chronological logging and analyses and geotechnical data obtained by selective sampling and in situ testing. As described earlier, it can be instructive to consider the results of incremental oedometer tests on some of the interpreted soil facies in void index – vertical effective stress space. Figure 23(a) shows envelopes of oedometer results from tests performed on WND normally consolidated hemipelagic clays and others inferred to be mass flow deposits (debrites, mudflows and slides). The contrasting soil fabrics for these two © 2011 by Taylor & Francis Group, LLC

22

Figure 23. One-dimensional consolidation tests – WND Soils.

Figure 24. One-dimensional consolidation tests – Angolan soils.

and soil units and may provide the geologicalgeomechanical link that is required for a truly integrated approach to geotechnical engineering, particularly in geologically complex areas. – Geotechnical Modifiers. Some of the Soil Units may have local depositional fabrics or structures or have undergone local post-depositional changes © 2011 by Taylor & Francis Group, LLC

similar to those discussed earlier in this paper. These features and characteristics may not necessarily be resolvable by geophysics but nevertheless may have a significant effect on soil behaviour. Examples include sediments with high void indices and soils affected by cementation, hydrocarbon

23

contamination, bioturbation and pre-existing shear surfaces – Regional Soil Parameters. Each of the main Soil Units is assigned representative mass soil parameters, such as undrained shear strengths and unit weights, which are inferred from the results of the geotechnical calibration programmes. These regional soil parameters are defined as credible ranges to reflect the levels of uncertainty with the data interpretation. The SHANSEP technique was not used to derive regional soil parameters in Angola and Egypt since the soils of greatest interest for engineering were considered to be too brittle or too variable or complex to apply the method reliably over large areas. However, the technique has been used selectively to infer operational undrained shear strengths for specific geotechnical designs or geohazard assessments where there are abrupt changes in OCR/YSR. – Operational Geotechnical Parameters.The geotechnical parameters used for engineering wells and facilities and carrying out geohazard assessments are necessarily problem-specific since they need to account for: (1) the mass properties of the soils, including the effects of geotechnical modifiers, (2) the type of structure or geohazard, (3) the imposed loads, and (4) the methods of analysis or design (code-based, reliability-based etc). For most of BP’s work in Egypt and Angola the regional soil parameters derived from the ground model databases have been used as the reference data for deriving operational design parameters. Location-specific data are used where local conditions are significantly different to the regional model and/or when geotechnical conditions may have an important bearing on safety or cost. Operational geotechnical parameters are key input to Geotechnical Baseline Reports used as reference for detailed design and construction contracts and as such, should be selected with support from the project engineering teams. The use of regional geotechnical data as the basis for design generally leads to higher factors of safety to achieve that same level of reliability as obtained using location-specific data. However, the strategy is pragmatic in largely uniform marine conditions and is broadly in line with one of the design approaches described in Eurocode 7 (2004). It also provides the flexibility to progress the geotechnical planning and design productively despite inevitable scheme changes. Nevertheless, significant effort is needed to identify atypical conditions that may control the mechanical behaviour of soil and invalidate the regional model. 3.6

Scheme definition – geohazard screening and risk assessment, and scheme layout planning

The geohazard management strategy being adopted in Angola and Egypt may be summarised as: – Assess the zones of influence of potentially active geohazards and where possible avoid them. © 2011 by Taylor & Francis Group, LLC

24

– Where, for overriding reasons, ‘geohazardous’ areas are unavoidable, assess the risks in terms of probability of occurrence and impact and consider special geohazard-resistant engineering measures to reduce the risks to tolerable levels. Geohazard Screening and Risk Assessment. The first step in this process is to screen the development area for geohazard ‘hotspots’ that denote major risk drivers.The method involves considering potential event scenarios associated each of the geohazards in the ground model database and eliminating those that are not significant because they are very unlikely to occur during the project life, would have little impact on the development or could be accommodated by routine design. The geohazard scenarios considered in this study include natural incidents similar to those in the past, reactivated geohazards, completely new events and man-made effects. The magnitudes and frequencies of the historical events, such as submarine slides, were assessed by multidisciplinary expert consensus opinion informed by the ground model terrain evaluation work, as illustrated on Figure 25. This process was supported by geomechanical back analysis and forecasting using simplified inhouse methods that are suitable for screening large areas. For example, the potential for shallow submarine slides was investigated using an undrained sliding block model (Newmark, 1965) that has been adapted to account for the effects of residual excess pore pressures, earthquake loading and soil brittleness. Similarly, the return periods for delayed failures of scarps (Potts et al. 1997, Leroueil 2001), formed in previous slides were estimated by a simple pore pressure equilibration model. Regional geotechnical parameters were generally used in the screening process but the local effects of exceptionally weak or brittle layers, or pre-existing shear surfaces (geotechnical modifiers) were also considered. Complementary engineering studies were performed to assess the potential impacts of geohazards such as submarine slides on typical seabed infrastructure (Parker et al. 2008, 2009). The risk assessment process is being facilitated by a series interactive multidisciplinary workshop sessions and the results are integrated in to the overall project risk assessment using the Qualitative, Semi-quantitative and Quantitative Risk Assessment methods shown on Figure 13. Scheme Layout Planning. The results of the geohazard screening and risk assessment are used with other criteria to optimise the scheme layouts. This is a multidisciplinary task that depends on a number of factors such as reservoir access, flow assurance and cost, as well as geohazard threats. The ideal objective for the GAT is to avoid all the buffer zones associated with the major geohazards. However, compromises are inevitable, and for technical or economic reasons it may be necessary to encroach some of the buffer zones. In such cases more detailed location- and structure-specific assessments are performed to define the geohazard risks more accurately.

Figure 25. Assessment of frequencies and magnitudes of submarine slides – Fully-developed slab slides.

Specialised geohazard-resistant engineering measures are also being considered to reduce the risks in these areas. Optimised Reference Scheme. The layout planning is an iterative process performed throughout the Appraise to Define stages of a project with the ultimate objective of developing a geohazard-tolerant reference scheme that can be taken forward in to detailed engineering and construction, generally performed through an EPC contract. 3.7

solutions, often derived from first principles, are being sought for complex structures for which there is little or no experience. Geohazardous Areas. Similar design approaches are also being applied in geohazardous areas, although additional special geohazard-resistant design measures are also considered. Although sometimes difficult to achieve, the overall objective is to develop safe and functional standard engineering solutions in both non-geohazardous and geohazardous areas and to limit the use of bespoke designs. Code-based Designs. As discussed earlier industry and international guidance for the design of shallow and deep offshore foundations has improved considerably in recent years and is beginning to address the complex behaviour of many subsea structures (Jeanjean et al. 2010). However, codes still provide solutions for relatively simple structures designed to avoid failure, with less consideration for serviceability criteria such as deflections. BP’s normal practice is to use published codes such as ISO 19901-4 (2003), ISO 19902 (2007) and API RP2A (2000) where possible but to consider alternative methods should these standard methods not be applicable, or if they are known to be excessively conservative. Value-engineering Solutions. Value engineering is a process by which design function and/or design reliability are increased at no additional cost, or cost

Engineering

The design strategy being followed by BP in their West Nile Delta and Angolan deep-water programmes is shown on Figure 26. The methods are based on the nature of the seabed and shallow subsurface conditions and the types of structures, and are driven by the scheme layout and architecture. Non-geohazardous Areas. In areas designated as non-geohazardous, simple structures with relatively predictable performance are being designed using codified methods and where such methods do not exist or are uneconomic or unsuitable, more pragmatic value engineering solutions are being adopted. The use of centrifuge tests to calibrate the design of suction anchors in layered soils for the GoM Mad Dog project described by Jeanjean et al. (2005) is an example of such a pragmatic design approach. New design © 2011 by Taylor & Francis Group, LLC

25

Figure 26. Design strategies for non-geohazardous and geohazardous areas.

longer term it is important to validate and/or refine these methods through field monitoring. The idealised flow diagrams in Figures 27 and 28 illustrate just two possible options for developing and calibrating these new designs methods. Figure 27 shows an approach that is based on the observational method. In this case a structuresoil interaction behavioural model is developed and used with project-specific soil and structures data to produce a job-specific prototype design to meet the project’s performance objectives. Provision is then made to monitor the structural response during field operations. If observed behaviour matches expectations the design may be verified. Otherwise, preplanned intervention and remedial measures may have to be implemented to assure acceptable life-of-field performance. Data from the field monitoring is also used to update and improve the behavioural model for future designs. Monitoring of flowline buckling and walking in Angola and the design and assurance process applied for the West of Shetland Clair Phase 1 Platform piles (Evans et al. 2010), are examples of the use of the observational method in BP. The method shown in Figure 28 is often used in onshore geotechnics to confirm design assumptions at sites or areas where there is little or no previous experience and is one of the ways recommended in ISO 19901-4 to manage geotechnical uncertainty Typically, it involves carrying out a field trial in advance of construction contract to validate or calibrate design assumptions for a specific site or types of soil and to update these accordingly. An example would be preliminary trials to confirm the design methodologies for predicting the axial capacities and load-deflection responses of piles which would then be used in projectspecific designs. It would be very rare for a single operator to do this in a frontier deep water area for a single project because of the high cost. However, the idea is more appealing when multiple projects are planned and trials can be performed opportunistically

is reduced without affecting function and reliability. Keaton and Eckhoff (1990) describe the approach in geologic hazard risk management terms as eliminating or reducing those aspects of a system that add cost without reducing risk. In BP the application of value engineering includes such things as challenging conventional thinking, doing things differently, design innovation, lateral thinking, adapting existing practices, extending the performance envelope and risk-based design with rigorous validation. In recent years BP has spent considerable effort investigating and developing cost-effective and safe engineering solutions based on these principles. Installation is a major cost driver and there is a big incentive to reduce the size and weight of subsea foundations and anchors. BP is tackling this challenge by developing reliability- and function-based (performance-based) designs that require the project to set performance objectives, such as failure probability, system displacements and tolerable levels of damage. These methods may also involve performance monitoring to manage the residual uncertainties, as practised in the observational engineering approach (Eurocode 7 2004). Other cost-saving initiatives presently being pursued include the development of composite foundations, the adaptation of wellconductor jetting techniques for installing piles and the use of numerical and physical modeling to challenge conventional design practices such as the use beamcolumn analyses with API RP2A p-y springs (Jeanjean 2009). New Design Methods. Many subsea structures have complex foundation behaviours and seabed interactions for which there is very little or no operational experience. In such cases it is necessary, at least in the short term, to develop design methods largely from first principles using mathematical and physical modelling and the results analogous structures or research together with expert judgement. In the © 2011 by Taylor & Francis Group, LLC

26

Figure 27. New design solutions using observational methods.

Figure 28. New design solutions using field trials.

by increasing component flexibility and/or strength. The key conclusions drawn from these studies are:

in the early projects to benefit later ones. For example in Angola BP is planning on leveraging selective deepwater pile design and installation experiences on a later project currently in Execute to extend the foundation design options on later projects. Investment in frontier trials would also more attractive when performed through joint industry projects (JIPs) sponsored by a number of operators, contractors and other interested parties with common interests. As engineering experience with new behavioural models increases and design confidence grows the plan would be to adopt them as standard company practices. In the longer term, this field experience may also provide the bases for industry best-practices, particularly if it could be pooled with similar data from other offshore operators. Geohazard-resistant Design. Preliminary studies for BP (Parker et al. 2008, 2009), have shown that the susceptibility of wells and bottom-founded facilities to geohazards such as submarine slides may be reduced © 2011 by Taylor & Francis Group, LLC

– Pipelines. The vulnerability of pipelines in landslide areas may be reduced by laying them in curves, a common practice in the GoM. – Sub-seabed soil movements. Subsea structures may be made more resistant to sub-seabed soil movements (landslides, erosive run-outs, fault movement etc) by increasing foundation capacities; driven piles being more effective for this purpose than suction-installed caissons due to the smaller area exposed to soil forces. Conversely, structural connections may be made more tolerant of this type of movement by increasing their flexibilities rather than strength. – Supra-seabed soil movements. Stronger foundations would also increase resistance to impact forces above the mudline (turbidity currents, mud volcano outflows, fluid non-erosive debris flows)

27

Procurement and Construction (EPC) contracting. In the short term, it may even be necessary for Operators to accept more responsibility than normal for the operational performance of designs based on methods that are not field-calibrated. Irrespective, EPC tender and contract documents will need to be adapted to ensure that the risks for new design methods are fully defined and roles and responsibilities for the work are clearly understood. As mentioned earlier, one possible way of defining and agreeing contractual risks for non-standard design solutions is through the use of Geotechnical Baseline Reports (GBRs) which are discussed further below. 3.8

Earlier in this paper, the communication of the shallow subsurface conditions to stakeholders was identified as a big challenge for geo-specialists. The approach taken by the Egypt and Angola GATs has been to merge the results of their work in a single document called the Geotechnical Engineering and Geohazard Mitigation (GGM) Report that is used to assist project decision making and to underpin more purposeful deliverables that support engineering and construction, as shown on Figure 30. The GGM Report evolves as more geophysical and geotechnical data are gathered and the project’s development plans emerge and represents the consensus state-of-awareness of the shallow subsurface conditions and how these affect the planned development. The GGM Report is updated whenever there is a significant change in the state-of-knowledge or project status, triggered for example by a programme decision gate, additional survey data or major scheme changes. The GGM Report embodies the ground model and amongst other things, describes the interpreted geological and geotechnical conditions and geohazards and, with input from the engineering teams, provides general guidance on geohazard avoidance, layout planning and engineering.The GGM database is captured in GIS format for ease of communication and interfacing with others in the project team. The GGM Report also provides the platform for the production of more engineering-focused and contractual documents such as the Geotechnical Basis of Design (BoD) that is specifically tailored to the scheme taken forward through FEED in to an EPC contract. The two key elements of the Geotechnical BoD are the Geotechnical Data Report (GDR) that includes all the factual data gathered to support the project and the Geotechnical Baseline Report (GBR) that contains interpreted or assumed ‘baseline, or ‘reference’ statements about the ground conditions and risks. EPC tenderers are encouraged to propose supplementary surveys and geotechnical investigations to support the BoD should they consider the data provided at the bid stage to be insufficient to support their specific design and installation plans. The GBR represents the project’s consensus assessment of the ground conditions and geohazards and how these will be taken in to account in design and for

Figure 29. Geohazard impact framework.

and piles and caissons are equally suitable for this purpose. Structural connections also need to be strengthened but this is a design challenge since it conflicts with the need for increased connector compliance to accommodate sub-seabed ground movements – ‘Weak-link’ Design. When safety and environmental considerations are accounted for, the most pragmatic geohazard-resistant design strategy may be to provide weak links that are relatively easy to repair. Equipment, structures, foundations and anchors used routinely offshore will have some inherent reserve capacities to resist geohazard-related loads However, the resistance thresholds may be increased and risks reduced by additional engineering as illustrated on Figure 29. The investment in geohazard-resistant design would be justified if it is less than the additional operating costs and production losses that would arise if the measures were not taken. Probability-based cost-benefit analyses would support this decision. Additional Considerations. The motives behind value-engineering and new design solutions are to enhance design by increasing reliability or reducing costs, or to enable developments to proceed in controlled and responsible ways in the absence of significant field experience. However, the methods are not universally applicable and therefore need to be used selectively and with care. For example, the observational method may not be suitable for safetycritical structures or where soil-structure interactions are insufficiently ductile to give warning and time to implement planned interventions. Performance-based methods may also need to be avoided if there is potential for brittle behaviour. There may also be some contractual considerations since many of the non-routine design methods are not ideally suited to traditional offshore Engineering, © 2011 by Taylor & Francis Group, LLC

Deliverables

28

Figure 30. Geotechnical documentation.

that may not be justified or practical for a single project. These include opportunities for:

construction. It also provides reference conditions for tendering or setting cost targets and for comparing with actual conditions encountered to support the interpretation of any ‘Differing or Unforeseen Site Condition’ clauses included in contracts. GBRs are most effective when they are prepared jointly by the Operator’s design team and EPC contractors and are agreed before contract award.

4

– Long term agreements with geophysical and geotechnical survey contractors. – Early geophysical and geotechnical data acquisition – The use of regional ground models to support the systematic assessment of geotechnical and geohazard risks and the geotechnical engineering for wells and facilities. – Use of observational methods and field trials to support value engineering solutions or new design methods for foundation and anchor design and for predicting complex soil-structure interactions. – Applying best-practices from onshore civil engineering projects with high geotechnical exposures, such as tunnelling, to develop strategies for managing offshore geotechnical risks within an EPC contracting framework that satisfy all parties.

CONCLUDING REMARKS

This paper describes the problems faced by offshore oil and gas companies operating in geohazardous and geotechnically-challenging areas and some of the ways that BP’s geotechnical engineers and other geospecialists are tackling them. Jeanjean et al. (2005) described some guiding principles and best-practices from the company’s pioneering deep water developments in geohazard-prone areas in the Gulf of Mexico. These included: (1) operator whole-life ownership, (2) multidisciplinary geoteams, (3) phased development-wide geophysical surveys and geotechnical site investigations and (4) the development of new engineering solutions using numerical and physical modeling and reliability-based design. This paper develops these ideas and concepts with particular reference to strategies and practices that are unfolding from deep-water development programmes in the West Nile Delta and Angola. Multiple projects are planned in both these regions which allowed BP to take a long-term holistic approach to geotechnical and geohazard risk management and to take advantage of a number of investment and leveraging opportunities © 2011 by Taylor & Francis Group, LLC

The approach that BP is taking to manage and mitigate geotechnical and geohazard risks in Egypt and Angola is clearly not the only way and may not necessarily be suitable or practical for all projects. Other, equally valid, management and mitigation strategies that embody the best practices and principles described by Jeanjean et al. (2005) and in this paper are being adopted by the company in other regions such as GoM and in the Caspian Sea. Some of the ideas and practices discussed in this paper are work-in-progress and others are still aspirational. However, to date the methodology adopted has been relatively successful. Arguably, the most significant steps were: (1) the establishment of semipermanent geotechnical and geohazard assessment

29

teams, GATs, for both regions to help provide projectlife geotechnical support and (2) to co-locate the GATs centrally, close to their respective project teams. The centralised GATs have technical and commercial advantages over more ad-hoc arrangements generally made for single projects, including: critical mass, proximity to project teams, technology transfer, shared best-practices and learning, standardisation, training opportunities and economies of scale. Perhaps the most important benefit of the GAT approach is that it has improved the interactions between the shallow geospecialists, deep geospecialists and the engineering teams and encouraged the holistic teamwork that is so often the hallmark of a successful project.

stress design, API recommended Practice RP-2A-WSD, 21st edition. Bingham, D., Drake, T., Hill A & Lott, R. 2002. The Application of Autonomous Underwater Vehicle (AUV) Technology in the Oil Industry – Vision and Experience, TS4.4, Hydrographic Surveying II, Fig XXII, International Congress, Washington DC, USA Brunsden, D. 2010. ‘A review of literature on submarine slope instability processes’. BP Egypt GAT Internal Project Report, unpublished. Bruton, D.A.S., Carr, M. & White, D.J. 2007. ‘The Influence of Pipe-Soil interaction on lateral Buckling and walking of Pipelines – The Safebuck JIP’, 6th International Conference, Offshore Site Investigation and Geotechnics, SUT, London. Burland, J.B. 1990. ‘On the compressibility and shear strength of natural clays’. 30th Rankine Lecture, Géotechnique, 40, No 3, 327–378. Chandler, R.J. 2000. Clay Sediments in Depositional Basins: the Geotechnical Cycle, Quarterly Journal of Engineering Geology and Hydrogeology, 33, 7–39. CIRIA. 1999.The Observational Method in ground engineering: principles and applications, Construction Industry Research and Information Association, Report 185, London. Clayton,C. & Power,P. 2002. Managing geotechnical risk in Deepwater., 5th International Conference, Offshore Site Investigation and Geotechnics, SUT, London. Clukey, E.C., Banon, H., & Kulhawy, F. 2000. Reliability Assessment of Deepwater Suction Caissons. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 12192. Cotecchia, F & Chandler, R.J. 2000. ‘A general framework for the mechanical behaviour of clays’. , Géotechnique, 50, No 4, 431–447. De Gennaro, V., Delage, P. & Puech, A. 2005. On the compressibility of deepwater sediments of the Gulf of Guinea. First symposium in Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds). Denis, R. & De Brier, C. 2010. Deep Water Tool for Insitu Pipe-soil Interaction Measurement: Recent Developments and System Improvement. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 20630. DNV. 1992. Foundations, Classification Notes No 30.4, Det Norske Veritas. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origins of near-seafloor “crust zones” in deepwater. First symposium in Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds). Essex, R.J. 2007. Geotechnical Baseline Reports for Construction. Technical Committee on Geotechnical Reports of the Underground Technology Research Council, American Society of Civil Engineers. Evans, T.G., Usher, N. & Moore, R. 2007. Management of Geotechnical and Geohazard Risks in the West Nile Delta, Proceedings 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. Evans, T.G., Finnie, I., Little, R., Jardine R.J. & Aldridge, T.R. 2010. BP Clair Phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till, Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Eurocode 7. 2004. BS EN 1997-1:2004 Geotechnical Design. General Rules, BSI. Gens,A. 2010. Soil -environment interactions in geotechnical engineering. Géotechnique 60, No 1, pp 3–74. Gilbert, R.B, & Gambino, S.J. 1999. Reliability-based Approach for Foundation Design without Site-Specific

ACKNOWLEDGEMENTS The author is grateful to BP Exploration and Production Technology (BP EPT) for permission to publish this paper, although the views expressed are wholly his own and not necessarily those of BP. The author wishes to thank his BP geotechnical colleagues, Junius Allen, Jim Clarke, Ed Clukey, Paul Dimmock, James Hansen, Kevin Hampson, Andy Hill, Hugo Galanes-Alvarez, Philippe Jeanjean,Attasit Korchaiyapruk, Eric Liedtke and Tony Lusted for their views and support, and Bryn Austin and Mike Fiske for leading the GAT geophysical interpretation work. Special thanks to Mike Sweeney for his support and encouragement. Also, thanks to Giles Thompson and his colleagues at Senergy Survey & Geoengineering, Tore Kvalstad of NGI, Eric Parker and his colleagues at D’Appolonia and Alan Niedoroda of URS Corp for strengthening the team when it matters. Finally, this paper would not have been possible if had not been for the dedication, hard work and professionalism of a diverse team of specialists from Halcrow Group Limited and Fugro GeoConsulting Limited who have underpinned BP’s Angola and Egypt GATs over the past 6 years. The individuals are too many to mention here but I will single out three; Roger Moore and Denys Brunsden of Halcrow and Steve Thomas of Fugro, who have been here since the beginning. Many thanks. REFERENCES AGS. 2003. Response to UK Government Minister for Construction. Association of Geotechnical and Geoenvironmental Specialists. Aldridge, T.R., Carrington, T.M., Jardine R.J., Little, R., Evans, T.G. & Finnie, I. BP Clair Phase, 2010. ‘Offshore foundation design in extremely hard till’, Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Alm,T., Snell, R.O., Hampson, K.M.& Olaussen, A. 2004. Design and Installation of the Valhall piggyback structures. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 16294. API. 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – working © 2011 by Taylor & Francis Group, LLC

30

Jeanjean, P., Liedtke, E., Clukey, E.C., Hampson, K. & Evans,T. 2005. ‘An Operator’s perspective on offshore risk assessment and geotechnical design in geohazard-prone areas’, First International Symposium on Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds) Jeanjean, P., Watson, P.G. Kolk, H.J. 7 Lacasse, S. 2010. RP GEO: The new API recommended Practice for Geotechnical Engineering. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 20631. Kay, S., Goedemoed, S.S. & Vermeijden, C.A. 2005. Influence of Salinity on Soil Properties. First International Symposium on Frontiers in offshore Geotechnics: ISFOG 2005– Gourvenec & Cassidy (eds). Keaton, J.R. & Eckhoff, D.W. 1990. Value Engineering Approach to Geologic Hazard Risk Management. Transportaion Research Record 1288, Geotechnical Engineering, 1990, 168–174, Transport Research Board, Washington DC. Kvalstad, T.J. 2007. What is the Current “Best Practice” in Offshore Geohazard Investigations? A State-of-the -Art Review’, Proceedings Offshore Technology Conference, Houston, Texas. OTC 18545. Kuo, M.Y.H., Bolton, M.D., Hill, A.J. & Rattley, M. 2010. Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Ladd, C.C. & Foot, R. 1974. New Design Procedures for Stability of Soft Clays. ASCE, Journal of the Geotechnical Engineering Division, Vol. 100, No. GT7, 763–786. Leroueil, S. 2001. ‘Natural slopes and cuts: movement and failure mechanisms’. 39th Rankine Lecture, Géotechnique, 51, No 3, 197–243. Le, M-H., Nauroy, J-F., De Gennaro, Delage, P., Flavigny, E., Thanh, N., Colliat, J-L., Puech, A. & Meunier, J. 2008. Characterization of Soft Deepwater West Africa Clays: SHANSEP Testing is Not Recommended for Sensitive Structured Clays. Proceedings, Offshore Technology Conference, Houston, Texas. OTC 19193. Locat, J. 2001. ‘Instabilities along ocean margins: a geomorphological and geotechnical perspective’. Marine and Petroleum Geology, 18, 503–512. Locat, J. & Lee, H.J. 2000. Submarine Landslides: Advances and Challenges. Proceedings of the 8th International Symposium on Landslides, Cardiff, UK, June. Mithchell, J.K. 1976. Fundamentals of Soil Behaviour, New york, Wiley. MMS. 1998. US Department of the Interior Minerals Management Service. Notice to lessees and Operators of Federal Oil, Gas and Sulphur Leases in the Outer Shelf Gulf of Mexico OCS Region, Shallow Hazard Requirements, NTL 98-20, September 15, 1998. Moore, R., Usher, N. & Evans, T. 2007. ‘Integrated Multidisciplinary Seismic Geomorphology Assessment of West Nile Delta Geohazards’, Proceedings 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. Newmark, N. 1965. Effects of earthquakes on dams and embankments. Géotechnique, 15, No 2, 139–160. Nimblett, J.N., Shipp, R. C. & Strijbos, F. 2005. Gas Hydrate as a Drilling Hazard: Examples from Global Deepwater Settings. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 17476. NORSOK. 2004. Marine Soil Investigations, NORSOK Standard G-001. Power PT & Clayton, C.R. 2003. ‘Managing Geotechnical Risk in Deepwater off West Africa’, 7th Annual offshore West Africa Conference and Exhibition, Windhoek, Namibia.

Soil Borings. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 10927. Griffiths, D.V., Huang, J. & Gordon, A. F. 2009. On the reliability of earth slopes in three dimensions. Proceedings of the Royal Society, A 2009 465, 3145–3164. Hadley, D., Peters, D. & Vaughan, A. 2008. Gumusut-Kaakap Project: Geohazard Characterisation and Impact on Field Development Plans, IPTC 12554, International Petroleum Technology Conference. Hight, D.W., Bond, A.J. & Legge, J.D. 1992. Characterization of the Bothkennar clay:: an overview. Géotechnique, 33, No 2, 327–340. Hight, D.W. & Lerouiel, S. 1993. Characterisation of soils for engineering purposes. Characterisation of soils for engineering purposes. Tan et al. (eds), Vol 1, 255–362. Hill, A.J. & Jacob, H. 2008. In-situ Measurement of Pipe-Soil Interaction in Deep Water. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 19528. Hill, A.J., Fiske, M., Fish, P.R. & Thomas, S. 2010a. Deepwater Angola: Geohazard Mitigation. Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia, November 2010. Hill, A.J., Evans, T.G., Mackenzie, B. & Thompson, G. 2010b. Deepwater Angola: Geotechnical Challenges. Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Hobbs, R. & Senner, D.W.F. 1998. Safety Implications for Offshore Foundations of Conductor and Shallow Well Drilling. 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. HSE. 1997. Drilling and Installation Effects on Foundations. Offshore Technology Report-OTO94 005, UK Health and Safety Executive, June 1997. HSE. 2006. Guidance on Risk assessment for Offshore Installations, Offshore Information Sheet No 3/2006, UK Health and Safety Executive. ICE. 1991. Inadequate Site Investigation. Institution of Civil Engineers, Thomas Telford, London. Imperial College Consultants/Geotechnical Consulting Group. 2007. Geotechnical Analysis of Salt Diapir Effects on Properties and Stress States of Deepwater Sediments. Report to BP Exploration, April 2007, unpublished. ISO 19901-4. 2003. Petroleum and Natural Gas Industries – Specific Requirements for offshore Structures – Part 4: Geotechnical and Foundation Design Considerations. ISO 19902. 2007. Petroleum and Natural Gas Industries – Fixed Steel Offshore Structures. ISSMGE. 2005. Geotechnical & Geophysical Investigations for Offshore and Nearshore Developments. Technical Committee 1, International Society for Soil Mechanics and Geotechnical Engineering. Swan Consultants Ltd (ed). Jardine, R., Chow ,F., Overy., R. & Standing, J. 2005. ICP Design methods for Driven Piles in Sands and Clays. Thomas Telford, London. Jeanjean, P. 2009. Re-assessment of P-Y Cirves for Soft Clays from Centrifuge Testing and Finite Element Modeling. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 20158. Jeanjean. 2010. Private communication. Jeanjean, P., Hill, A. W. & Taylor, S. 2003. The Challenges of Siting Facilities along the Sigsbee Escarpment in the Southern Green Canyon Area of the Gulf of Mexico, Framework for Integrated Studies. Keynote lecture, Proceedings, Offshore Technology Conference, Houston, Texas, OTC 15156. © 2011 by Taylor & Francis Group, LLC

31

Parker, E.J., Traverso, C., Moore, R. & Evans, T. 2008. Evaluation of Landslide Impact on Deepwater Submarine Pipelines. Proceedings, Offshore Technology Conference, Houston, Texas. OTC 19459. Parker, E.J., Traverso, C., Del Giudice, T., Evans, T. & Moore, R. 2009. Evaluation of Landslide Impact on Deepwater Submarine Pipelines. Proceedings, Offshore Technology Conference, Houston, Texas. OTC 19459. Potts, D.M., Kovacevic, N. & Vaughan, P.R. 1997. Delayed collapse of cut slopes in stiff clay. Géotechnique, 47, No 5, 953–982. Randolph, M., Cassidy, M., Gourvenec, S. & Erbrich, C. 2005. Challenges for offshore geotechnical engineering. International Conference on Soil Mechanics and Geotechnical Engineering, ICSMGE. Osaka, Japan, 123–176. Schroeder, F.C., Jardine, R.J., Kovacevic, N. and Potts, D. M. 2007. Potential Effects of Well Drilling Operations on Foundation Piles in Clay. Proceedings 6th International Conference, Offshore site Investigation and Geotechnics, SUT, London. SUT, Society of Underwater Technology. 2003. Guidance Notes on Geotechnical Investigations for Subsea Structures. Offshore Soil Investigation and Geotechnics Group. SUT, Society of Underwater Technology. 2004. Guidance Notes on Geotechnical Investigations for Marine Pipelines, Offshore Soil Investigation and Geotechnics Group.

© 2011 by Taylor & Francis Group, LLC

Sultan, N. 2007. Gas hydrates stability Law, Hydrate Fraction and acoustic Velocities of Gas-Hydrate Bearing Sediments: Theoretical Study and empirical Expressions, Gas Hydrate Research Project. West Africa Deepwater Operators, (WADO), Deliverable 1. Sultan, N., Adam, S., De Gennaro, V., Lakshmikantha, M.R. & Puech, A. 2009. Acoustic properties and mechanical behaviour of marine sediments partially saturated by gas. IFREMER Scientific Report Task 2, Joint Industry Funded Project. Thomas, S., Bell, L., Ticehurst, K. & Dimmock, P. S. 2010. ‘An investigation of past mass movement events in the West Nile Delta’. Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Tjelta,T. I. 2010. ‘Prod probes Statoil’s seabed soils’. Offshore Engineer, February 2010. UKOOA. 1997. UK Offshore Operators Association. ‘Guidelines for the Conduct of Mobile Drilling Rig Site Surveys’, Volume 1, March 1997 and Volume 2, April 1996. USNCTT. 1984. U.S National Committee on Tunneling Technology, Geotechnical Site Investigations for Underground Projects, National Research Council, Washington D.C. (2 Volumes) Wheeler, S.J. 1988. The undrained shear strength of soils containing large gas bubbles. Géotechnique, 38, No 3, 399–413.

32

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Recommended best practice for geotechnical site characterisation of cohesive offshore sediments D.J. DeGroot University of Massachusetts Amherst, Amherst, MA, USA

T. Lunne Norwegian Geotechnical Institute, Oslo, Norway

T.I. Tjelta Statoil, Norway

ABSTRACT: The paper presents recommendations for conducting site characterisation programs to determine the geotechnical properties of cohesive offshore sediments with the primary focus being on clays. The engineering behaviour of clays is complex and characterisation of their in situ properties is magnified by additional challenges presented by offshore environments. Site investigation programs are best conducted using well calibrated in situ tests and laboratory testing of high quality undisturbed samples. Results from these test programs should be coupled with geophysical data and collectively evaluated in the context of a regional geological framework. The paper reviews clay behaviour, describes the unique conditions of offshore environments, and lists key cohesive soil parameters required for infrastructure design and geohazards analysis. Best practice recommendations founded on these fundamentals are then presented including drilling methods, in situ testing and instrumentation, soil sampling, laboratory testing and evaluation of test data. The paper concludes with an assessment of present and future challenges. 1

INTRODUCTION

regional geologic framework that encompasses the past and current geologic states. The paper discusses characterisation of cohesive offshore sediments, which can encompass clays, plastic silts and organic soils, although the main focus is on soft to medium consistency clays. This is in part due to the authors’ experience is primarily with clay deposits, but also because a large portion of offshore sites consist of soft clay deposits, especially in deeper waters. There are equally important challenges with characterisation of other offshore deposits such as sands, calcareous soils, and non-plastic silts. While some of the site characterisation methods described are to varying degrees relevant to such soils, they are not the focus of the paper. The paper begins with a review of clay behaviour since such knowledge is essential to designing and implementing an appropriate site characterisation program. The unique and complex challenges presented by offshore environments such as over pressured zones, ultradeep waters, high salinity, ice gouging, etc. are described. Specific soil parameters required for design of offshore infrastructure such as anchor systems, pipelines, conductor installation, and geohazards analysis are listed. These background sections provide the context for the main objective of the paper which is to present recommended best practice for geotechnical site characterisation programs to determine reliable

Offshore geotechnical site characterisation programs are used to determine soil stratigraphy, in situ pore water pressure, and soil parameters for foundation design of infrastructure and geohazards analysis. Cohesive soils can have highly varied geologic histories making systematic quantification of their stressstrain behaviour complex. They exhibit significant stress history effects, can have a high degree of anisotropy, and exhibit strain rate effects. They are difficult to sample and test without causing excessive and irreversible disturbance. These challenges are magnified for offshore site characterisation programs since many offshore regions have complex geologic environments and site investigations are increasingly being conducted in deeper waters. Geotechnical site characterisation programs ideally combine in situ testing, in situ instrumentation, collection of high quality samples and follow-on laboratory testing. Assessment of sample quality is essential for evaluating the accuracy of laboratorymeasured mechanical properties. Coupling laboratory data with results from well calibrated in situ tests greatly enhances the reliability of site characterisation programs. The resulting geotechnical data should be studied concurrently with geophysical data and the collective data set should be evaluated within a

33 © 2011 by Taylor & Francis Group, LLC

soil parameters. Recommendations are given for all geotechnical aspects of the site characterisation process including drilling methods, in situ testing and instrumentation, soil sampling, laboratory testing, and evaluation and selection of design profiles. This paper follows other related overview papers on offshore site characterisation, notably Randolph (2004), Kolk & Wegerif (2005), and Andersen et al. (2008). The paper stresses the importance of geophysical and geological work as part of a comprehensive site characterisation program but these topics are not within the authors’ expertise. ISSMGE (2005) reviews offshore geophysical methods.

2

FUNDAMENTAL CLAY BEHAVIOUR

An understanding of clay behaviour is necessary to formulate and conduct a successful site characterisation program. It is important for the planning and execution of the in situ testing, soil sampling, and laboratory test programs so that the most relevant tools and test methods are utilized. It is also valuable during the process of comparing and analyzing the measured data sets and ultimately in development of design profiles for key soil properties such as in situ vertical effective
stress (σv0 ), preconsolidation stress (σp
), consolidation behaviour, undrained shear strength (su ), and stiffness. This section reviews the fundamentals of intact (i.e. undisturbed) clay behaviour, the influence of sample disturbance on this behaviour, and the behaviour of remoulded clays. Leroueil & Hight (2003) and Ladd & DeGroot (2003) each provide a detailed treatise on clay behaviour; Lunne & Andersen (2007) emphasize behaviour of soft, deepwater clays. The focus herein is on monotonic testing behaviour; Andersen (2004, 2009) provides details on cyclic behaviour of clays.

Figure 1. Fundamentals of 1-D consolidation behaviour: compressibility, hydraulic conductivity, coefficient of consolidation vs. vertical effective stress (after Ladd & DeGroot 2003).

Key aspects of intact clay behaviour are stress history, consolidation, strength anisotropy, non-linear stressstrain behaviour, and strain rate effects.

Figure 2 plots su versus σp
data for 27 low to medium OCR clays worldwide (22 from offshore locations) and with a majority having a plasticity index (PI) between 10 to 50% (NGI 2002). The su values were determined from triaxial compression tests conducted on good to excellent quality samples that were anisotropically consolidated to estimated in situ effective stresses prior to undrained shear (CAUC). The σp
values were determined from 1-D consolidation tests conducted on companion test specimens taken from the same sample. The strong link between su and σp
implied in Figure 2 is often expressed in terms of normalised soil parameters such that

2.1.1 Stress history All significant aspects of clay behaviour are influenced by stress history. It is quantified through the preconsolidation stress and the corresponding overconsolidation
ratio OCR = σp
/σv0 . The preconsolidation stress is essentially a yield stress, which separates small, mostly elastic vertical strains (εv ) from large, mostly plastic strains, as illustrated in Figure 1. It is more appropri
ately referred to as the vertical yield stress (σvy ) but the familiar σp
notation is used in this paper. In terms of consolidation behaviour, Figure 1 also illustrates the significant changes in the coefficient of consolidation (cv ) as a function of stress levels relative to σp
. The hydraulic conductivity (kv ) decreases with an in increase in σv
with an approximate linear relationship between void ratio (e) and logkv .

with a common value of m = 0.8 ± 0.1 and S depending on the particular su measurement method/mode of shear being investigated (e.g. Ladd 1991, Ladd & DeGroot 2003). Mesri (1975) and Terzaghi et al. (1996) reason that su is directly proportional to σp
(i.e. m = 1) such as for example, in the case of stability problems with su (ave) = 0.22σp
, where su (ave) = the average mobilized su . Linear regression to the data set shown in Figure 2, which assumes m = 1, results in su (CAUC) = 0.28σp
. Jamiolkowski et al. (1985) describe several mechanisms that cause an OCR = 1 soil deposit to become overconsolidated including vertical stress relief, desiccation, ageing/drained creep (often referred to

2.1 Intact clay behaviour

34 © 2011 by Taylor & Francis Group, LLC

Figure 4. su (CAUE) vs. σp
for low to medium OCR and plasticity index offshore clays (from NGI 2002).

Figure 2. su (CAUC) vs. σp
for low to medium OCR and plasticity index offshore clays (from NGI 2002).

Figure 5. Schematic of shear modulus versus shear strain (modified after Jardine 1992, Leroueil & Hight 2003).
same database as the CAUC (σ1f = 0◦ ) data plotted in Figure 2. Linear regression to these data results in su (DSS) = 0.23σp
and su (CAUE) = 0.17σp
. The regression coefficients imply su anisotropy ratios of Ks = su (DSS)/su (CAUC) = 0.82 and su (CAUE)/ su (CAUC) = 0.61. While there is scatter in the data sets presented in Figures 2-4, the trends are evident and similar to that presented for terrestrial soils (e.g. Ladd 1991, Terzaghi et al. 1996) showing significant su anisotropy for these low to moderate plasticity soils.

Figure 3. su (DSS) vs. σp
for low to medium OCR and plasticity index offshore clays (from NGI 2002).

as apparent preconsolidation), and physicochemical effects (e.g. cementation). This highlights the importance of developing an understanding of the geologic history of a site so that stress history data from in situ and/or laboratory testing can be properly evaluated and understood.

2.1.3 Non-linear stress-strain behaviour The stress-strain behaviour of clays is highly nonlinear. Clay stiffness, expressed as Young’s modulus (E) or the shear modulus (G), is greatest at small strain levels (less than ∼ 0.01% for clays) and degrades with increasing strain. Jardine (1992) presented a conceptual model and examples of soils in which the stiffness-strain degradation curve is divided into several zones, as shown schematically in Figure 5. This includes linear-elastic behaviour (up to Pt. A), followed by non-linear elastic behaviour (Pts. A to

2.1.2 Anisotropy Undrained shear strength is said to be anisotropic when its magnitude depends on the orientation of
the major principal stress at failure (σ1f ), with su
decreasing as σ1f rotates from vertical (same direction as deposition) to horizontal. Figures 3 and 4 plot results from consolidated direct simple shear
∼ 45◦ ) and anisotropically consolidated (DSS; σ1f
triaxial extension tests (CAUE, σ1f = 90◦ ) for the

35 © 2011 by Taylor & Francis Group, LLC

B), plastic straining (Pts. B to C), and large plastic straining (beyond Pt. C). The shear stiffness within the region up to Pt. A is the maximum value (nearly constant) and is commonly referred to as the small strain shear modulus Gmax (or G0 ). It is often determined using elastic shear wave measurements such that

where ρt = total density and Vs = shear wave velocity. Like su , Gmax is also anisotropic and thus it should be reported with appropriate subscripts Gij to explicitly indicate the direction of propagation (i) and polarization (j) of the shear wave. For example, downhole seismic testing results in a measurement of Vvh and a computed Gvh while cross-hole testing yields either Ghh or Ghv . Figure 6. Normalised static undrained shear strength versus rate of shear strain for clays (from Lunne & Andersen 2007).

2.1.4 Strain rate effects Clays are sensitive to the rate of strain or loading and can exhibit a significant increase in su at fast rates of undrained shearing. Figure 6 shows an example of the influence of shear strain rate on su of several clays. The data show not just an increase in su with an increase in strain rate, but furthermore that the change in su per log cycle increases as the strain rate increases. The magnitude of the effect can depend on many factors including clay type, OCR, stress path, and may depend on whether the load is strain- or stresscontrolled and whether the shearing is static or cyclic (Lunne & Andersen 2007). Nevertheless, the effect is present for all such conditions and must be considered when evaluating laboratory and in situ test data relative to field conditions, all of which typically have significant differences in the rate and duration of loading. The rate effect shown in Figure 6 is often modeled using a simple log function with most data showing an increase in su between 5 to 20% per log cycle increase in strain rate. The effect can also be modeled using an inverse hyperbolic sine function (e.g. Randolph 2004), which allows for decaying strain rate effects (as is evident in Figure 6) below a specified threshold rate.

Figure 7. Hypothetical stress path for a low OCR clay element during tube sampling, specimen preparation and undrained shear (from Ladd and DeGroot 2003).

Each stage of the sampling process, from initiation of drilling to preparation of laboratory specimens, causes potential disturbance. The most important effect of sample disturbance in low to moderate OCR clays is a destructuring of the soil, which is accompanied by a significant reduction in the sample effective stress (σs
). For example, Figure 7 shows how the reality of sampling and testing can vary unpredictably from
). This figure shows the anticithe perfect sample (σps pated stress path for a low overconsolidated clay as stresses change from the in situ stress state (Point 1) to the stress state at laboratory testing (Point 9) as a result of disturbance caused by sampling, storage and handling. There are similar impacts on the small strain shear modulus, as depicted schematically in Figure 8 in terms of the shear wave velocity. Stress relief alone results in a reduction in Vvh , which is presumably recoverable during laboratory reconsolidation, however, destructuring results in an irreversible decrease in Vvh . It is common in geotechnical engineering practice to rely on the behaviour measured from laboratory

2.2 Influence of sample disturbance Sample disturbance is the most significant issue affecting the quality and reliability of laboratory test data for clays. It causes changes in the natural soil state and structure and as a result all key design parameters such as E, G, σp
and su are adversely influenced by sample disturbance. Figure 1 schematically shows this for the one-dimensional consolidation properties with sample disturbance resulting in a more rounded compression curve with greater εv at all stress levels. This tends to obscure and usually lower σp
, especially for more structured soils. The only parameters not significantly affected by sample disturbance are cv well beyond σp
and the e–logkv relationship, unless there is severe disturbance.

36 © 2011 by Taylor & Francis Group, LLC

Figure 8. Hypothetical reduction in sample effective stress and shear wave velocity of a low OCR clay for the perfect sample and a disturbed sample.

index strength tests (e.g. unconfined compression tests, UUC). In such tests, the specimen is at an effective stress state represented by Point 9 in Figure 7 when, in fact, the in situ behaviour under field loading starts at Point 1. Moreover, not only is the effective stress state incorrect, but destructuring that occurs during poor sampling and handling further magnifies differences between measured laboratory and in situ behaviour. As an example of these effects, Figure 9 plots data showing the influence of different samplers on the undrained shear behaviour of the Onsøy, Norway, clay. The results are from CAUC tests conducted on Sherbrooke Block, 76 mm tube and 54 mm tube samples. The Sherbrooke Block sampler generally collects high quality samples followed by decreasing sample quality for the 76 mm tube sampler and especially for the 54 mm tube sampler (e.g. Lunne et al. 1997a, 2006). There are significant differences in the strain-strain-strength behaviour as sample disturbance increases, including: decrease in su , decrease in rate of strain softening, and increase in large strain shear strength. Lunne et al. (2006) show additional examples from DSS and CAUE tests for the Onsøy clay and several other clays.

Figure 9. CAUC recompression tests conducted on samples of Onsøy, Norway, clay from 14.5 m: a) stress-stress curves, and b) effective stress paths (from DeGroot et al. 2007).

been performed using the field vane and more recently full-flow penetrometers. As an example of some of these issues, Figure 10 presents results of motorized laboratory vane (MLV) tests conducted on an intact sample of Troll clay from the Norwegian sector of the North Sea. The sample was first tested to measure the intact su and the post peak residual shear strength. It was subsequently remoulded through multiple rotations of the vane (VR) and again tested. Thereafter the sample was thoroughly remoulded by hand (HR), tested, and followed by a final remoulding through multiple vane rotations (HRVR). Different values of sur also resulted from other methods including that of the fall cone (FC). In fact, for the FC, there are different calibration factors in current use (e.g. Norwegian Standard, Swedish Standard, ISO) and different values result from these various methods. The resulting sensitivity values St = su /sur depend on which measure of sur , is used to compute the sensitivity, and in the case presented in Figure 10, ranges from 2 to 12 for the MLV. Lunne & Andersen (2007) present data from UUC tests conducted on remoulded clays at different strain rates. The results indicate that the rate effect clearly observed for intact clays (Figure 6) also occurs for remoulded clays. Preliminary tests conducted at NGI also showed sur may be anisotropic.

2.3 Remoulded clay behaviour Some design solutions (as outlined in Section 4.0) require information on the remoulded undrained shear strength (sur ) and also its strength and stiffness evolution with time after remoulding. Laboratory measurement of sur has the significant advantage over measurement of su because sample disturbance is generally not an issue. However, it remains a relatively complex parameter to evaluate because results can vary greatly depending on the degree of remoulding and the measurement method used. Some test devices require a remoulded test specimen to be prepared (e.g. fall cone) while other devices can be used to remould an intact soil (i.e. device remoulding) and also directly measure sur (e.g. laboratory or in situ vane test). In situ methods of estimating sur , which all employ device remoulding by default, have long

37 © 2011 by Taylor & Francis Group, LLC

Table 1.

Summary of key aspects of clay behaviour.

Clay behaviour

Significance

Stress history
(σv0 , σp
, OCR) su anisotropy

Most important factor, controls all significant aspects of clay behaviour Very significant for low to medium PI clays, su (CAUC) > su (DSS) > su (CAUE)† Significant degradation in stiffness (E or G) with increasing strain Increase in su with increase in strain rate Value depends significantly on how measured and method of remoulding. Exhibits rate effects and anisotropy Most significant issue affecting quality and reliability of laboratory test data

Non-linear stress-strain Rate effects Remoulded sur Sample disturbance

Note: † except for varved clays where su (DSS) < su (CAUE)

should always be considered when selecting equipment and test methods to conduct a geotechnical site investigation. Figure 10. Shear stress vs. angular rotation for motorized laboratory vane tests conducted on Troll clay from 4.1 m depth.

3

UNIQUE CONDITIONS IN OFFSHORE ENVIRONMENTS

The various aspects of fundamental clay behaviour presented in the previous section challenge any site investigation program. However, these challenges are magnified in the offshore environment because of numerous unique logistical issues and geologic conditions that are often present. This section presents some of these conditions and describes how they add another layer of complexity to characterising the engineering properties of offshore sediments. Many of these conditions are classified as offshore geohazards, a topic for which keynote papers were presented at ISFOG 2005 by Jeanjean et al. (2005) and at ISFOG 2010 by Evans (2010). 3.1

Figure 11. Fall cone undrained shear strength (left axis) and shear modulus (right axis) vs. time for thixotropic test conducted on remoulded Troll clay.

Remoulded clays, if left undisturbed at constant water content and temperature will typically exhibit significant thixotropic hardening. Figure 11 plots results for the same Troll clay in Figure 10 showing the significant gain in su and Gvh (as measured using bender elements) with time after thorough hand remoulding. The rate in gain in su and Gvh are similar implying a near constant rigidity index IR = Gvh /su during thixotropic hardening. 2.4

Summary

Table 1 lists a summary of the fundamental aspects of clay behaviour described in this section. It is evident that clay behaviour is complex and this complexity

38 © 2011 by Taylor & Francis Group, LLC

Non-hydrostatic pore pressures

Many offshore regions have in situ pore pressures that are in excess of hydrostatic (i.e. overpressures). These overpressure zones can be caused by rapid sedimentation, mud volcanoes, pore fluid flow, manmade activities, and other mechanisms as summarized by Dugan et al. (2010). Given that the in situ vertical effective stress profile is critical to determining soil design parameters, detection of overpressured zones and measurement of the overpressure is an essential part of a site characterisation program. Overpressures are believed to be a contributing factor to the massive Storegga submarine landslide in the Norwegian Sea (Kvalstad et al. 2005) and have caused major difficulties during drilling operations in the Caspian Sea (Allen et al. 2005). Figure 12 presents an example of overpressures for the West Azeri field in the AzeriChirag-Gunashli (ACG) development in the Caspian Sea as measured using pore pressure dissipation tests

Figure 13. Schematic of reduction in effective stress state for consolidated OCR = 1 clay subject to post deposition influx of pressurised pore water.

platform. There are many such cases in SE Asia where poor cementing results in gas charging of shallow permeable layers. 3.2 Figure 12. Pore water pressure versus depth, West Azeri, Caspian Sea (after Allen et al. 2005).

(Section 5.5). The magnitude of overpressure equals about 800 kPa at 230 m below the seabed, and over the profile is supporting 40 to 50% of the soil buoyant unit weight. Determining the genesis of overpressures is also important. If it is a result from ongoing self-weight consolidation in a rapid sedimentation environment then the deposit is considered “under-consolidated”
relative to the eventual final equilibrium σv0 . In such cases the deposit will be relatively weak and assuming hydrostatic pore pressures could lead to significant errors in selecting consolidation stresses for laboratory strength testing and evaluating final su design profiles. Conversely, a deposit that has excess pore pressures that evolved after self-weight deposition pore pressures have come to equilibrium should not be categorized as under-consolidated. One mechanism that can cause this is upward or lateral fluid flow from adjacent overpressure regions. If the onset of fluid flow excess pore pressure occurs after a previous equilibrium effective stress state, then the deposit will be overconsolidated (Figure 13). In some regions excess pore pressure from both rapid sedimentation and fluid flow from adjacent overpressure zones cans occur concurrently which makes for a complex in situ state to measure and evaluate. Although Dugan et al. (2010) note that recent advances in modeling overpressures have better defined different mechanisms and spatial scales. Overpressures can also result from manmade activities. For example, Lunne et al. (1996) report on gas charging of permeable layers from deep gas leaking upwards through weak cement around conductors in the Duyong B field, Malaysia. It was discovered during geotechnical drilling for a new nearby platform when a blow-out occurred. The gas charging built up pore pressures in adjacent clay layers threatening the safety of the foundation system of the existing

3.3

Highly irregular seafloor topography

There are numerous past and present geologic events that create highly irregular seafloor topography. Examples include mud volcanism, seabed slumping, small to mega-submarine landslides, debris flows, channel erosion, faulting, salt diapirism, and ice gouging. Such complexity in the seafloor topography presents obvious challenges to locating and installing offshore infrastructure, especially for routing of pipelines. It also implies that there can be significant spatial variability in soil properties, which requires careful planning and layout of the geotechnical site exploration program. 3.4

Gas exsolution

Stress relief from sampling, especially in deep waters, can cause exsolution of even small amounts of gas dissolved in the pore water. The subsequent expansion, if undrained, can damage the soil structure which in turn will impact the quality of the measured soil behaviour

39 © 2011 by Taylor & Francis Group, LLC

Low effective stress state

For flowlines and pipelines, the largest category of offshore infrastructure, soil-structure interaction takes place in the upper few decimeter of soil where there are
very low effective stress (e.g. σv0 ∼ 2 to 3 kPa at 0.5 m depth). This, along with temperature induced cyclic loading effects (from episodic flows causing pipe contraction and expansion), complicates characterisation of the in situ soil properties such as modulus and shear strength. Laboratory measurement using consolidated test specimens (e.g. CAUC) is difficult to do reliably at such low effective stresses. For in situ testing methods, such as the piezocone and full-flow penetrometers, the tip force acting on the device from hydrostatic water pressure at the sea bottom can be very large relative to the penetration resistance of the upper few decimeters of soil. This creates measurement inaccuracy issues; a problem that is especially significant for soft soils and/or as water depth increases (e.g. Lunne 2010).

and design parameters. Lunne et al. (2001) conducted a laboratory test program that simulated deepwater sampling of clay with various amounts of dissolved gas. The measured CAUC stress-strain-strength behaviour was influenced by the degree of gas saturation, with a large reduction in su with increasing gas saturation. When disturbance caused by gas exsolution was coupled with simulated tube sampling disturbance, the combined effect was significant. CAUC reconsolidation volumetric strains were larger implying greater disturbance from the combined effect. But the stressstrain curves tended to strain harden, presumably due to the large volumetric strains and reduction in void ratio prior to undrained shear. The results indicate that disturbance caused by gas exsolution depends on the state of soil structure prior to exsolution with an intact structure being better able to resist gas exsolution than an initially disturbed structure.

3.5

Surficial crust

Figure 14. Evidence of surficial crust zone based on CPTU data, offshore Nigeria (from Ehlers et al. 2005).

Ehlers et al. (2005) describe the presence of a thin crust zone near the seabed in several deepwater regions as shown for example in Figure 14. This su profile, with the relatively high strength within the upper meter, is unusual compared to the more familiar deepwater soft clay profile which starts with the lowest values at the seabed and increases near linearly with depth. It is hypothesized that these crusts are a product of bioturbation and geochemical transformation, primarily into pyrite (Ehlers et al. 2005). Detecting the presence of these thin surficial crusts and properly characterizing the strength and stiffness properties is especially important to design and installation of shallow foundation systems, flowlines and pipelines (Section 3.2).

3.6 High salinity Some offshore regions have been found to have extremely high pore fluid salinity. Figure 15 plots data for the West Azeri field in the Caspian Sea (same site as Figure 12). The pore fluid salinity reaches values as high as 250 g/kg in comparison with Caspian Sea water of 12 g/kg. This high pore fluid salinity is believed to be associated with mud volcano activity as there appears to be higher pore water salinity at locations that are close to mud volcanoes (Kay et al. 2005). The upward migration of materials within the mud volcanoes includes pore water from greater depths and may also form conduits for further fluid flow. Excess pore water pressure in the vicinity of the mud volcanoes provides a mechanism for migration of dense brine vertically and laterally. High pore water salinity needs to be taken into account for computation of in situ pore water pressure. It also impacts many aspects of laboratory testing including measurement of index properties that rely on an oven dry mass and the need to match water used for saturation and testing of consolidated specimens to the pore fluid salinity.

Figure 15. Pore fluid salinity vs. depth below seabed for West Azeri region, Caspian Sea (after van Paassen & Gareau 2004).

In the Caspian Sea ACG development, there are high plasticity clay units that were found to be slickensided (Allen et al. 2005). This indicates that the sediments were subjected to failure conditions with repeated shearing. These soil units were not exposed to subaerial conditions and thus have not been desiccated. The likely mechanisms are mechanical, such as deformation of the soil units over the crest of the underlying Apsheron Ridge, or physio-chemical. Synaeresis cracks can occur in submerged sediments due to volume changes induced by changes in pore fluid salinity (Collinson & Thompson 1982). These post-deposition salinity changes are present at several of the ACG fields where the original pore fluid has been displaced by high salinity pore fluid migrating

40 © 2011 by Taylor & Francis Group, LLC

Figure 16. Normalised preconsolidation stress versus temperature (from Leroueil & Marques 1996).

from adjacent overpressure zones. It is likely that the fluid migration is cyclic, and in combination with the high plasticity swelling clay minerals present, is a condition conducive to synaeresis. Figure 17. Preconsolidation stress for ice-gouged and non ice-gouged sites in Smith Bay, Beaufort Sea (modified after Young 1986).

3.7 Temperature changes In offshore polar regions and in deep water, the seabed temperature often approaches 0◦ C (and even slightly below) with the pore fluid freezing prevented by the salt concentration. For example, Bugge (1983) reported that on the Norwegian continental margin at latitude of 64◦ N, the sea water temperature equals 0◦ C at 850 m and reaches an equilibrium temperature of −0.9◦ C at 1300 m. In the Mexican part of the Gulf of Mexico, Vidal et al. (1994) report temperatures lower than 5◦ C in 1200 m water depth. Collecting soil samples from a 0◦ C degree environment and testing them at room temperature (∼20◦ C), which is the common practice, raises the issue of temperature effects on the measured behaviour. Another temperature effect is that during production, hot gas or oil in pipelines can cause heating of surrounding soil. Leroueil & Marques (1996) studied the influence of temperature on σp
and found a nearly 1% decrease in σp
per ◦ C increase for temperatures ranging from 5 to 40◦ C, as shown in Figure 16. Since su is linked to σp
, as described in Section 2.1.1, a similar temperature effect on su is expected (Perkins & Sjursen 2009). 3.8

out to sea at approximately 20 m water depth. Both sites involve the same Pleistocene deposit with Site W being heavily overconsolidated with nearly constant σp
suggesting mechanical precompression due to erosion. Site T is at a greater water depth and in a zone of extensive ice gouging, which was determined in part from geophysical data. The repeated reworking at Site T caused significant softening within the upper 2 meters of the initially heavily overconsolidated deposit. Below 2 meters the σp
values rapidly increase with depth as the effects of ice gouging diminish and coincide with those of Site W. Another interesting feature in offshore arctic regions is the presence of subsea relict permafrost below existing unfrozen sediment. For example, in regions of the Beaufort Sea, the surficial soils in shallow waters are often soft sediments that have a temperature just below freezing, but are unfrozen due to higher pore fluid salinity. These deposits were found to be underlain by permafrost, which formed during the last ice age when the sea level was much lower than present and the then subaerial deposits were subject to extreme freezing temperatures. During subsequent sea level rise, some of the top portion of the permafrost thawed and in some regions was covered with Holocene sediments. As a result, a veneer of soft sediment exists over a semi-rigid layer of permafrost. Foundation analysis for gravity based structures in such circumstances will involve modeling squeezing of a relatively thin layer of soft sediments. Additional problems, such as settlement, can result if oil and gas production melts the permafrost. In polar regions, ice loading can impose large lateral forces on offshore infrastructure such as gravity platforms. Ice loading depends on the nature of the ice encountered (i.e. landfast, transition or polar pack) and will have both mean and cyclic components. When

Ice gouging, subsea relict permafrost, and ice loading

Offshore regions that have been, and continue to be, influenced by the presence of sea ice often have several unique geologic features. Sea ice keels that interact with the seabed can gouge the surficial sediments. Repeated ice gouging reworks the seabed and can significantly reduce σp
and su . It can also result in significant variations in seafloor topography. Figure 17 plots an example for the Smith Bay region of the Beaufort Sea, offshoreAlaska.The Sites W andT are located about 17 km from each other with Site W closer to land at approximately 10 m water depth and Site W further

41 © 2011 by Taylor & Francis Group, LLC

Table 2. Some unique offshore conditions and implications for conduct of site characterisation programs.

combined with gravity forces imposed by the structure, it creates a complex stress condition within the foundation soil. In some cases it can include significant shear stress reversals that are difficult to characterise (e.g. DeGroot et al. 1996).

Condition

Site characterisation remarks (details in Section 5)

Overpressures

3.9

Essential to detect and measure, use in situ dissipation tests and/or install piezometers Low effective Use in situ tools with high sensitivity stress conditions or hydrostatically compensated CPTU, collect high quality surficial samples and test on the vessel deck (e.g. box coring) Highly irregular Select in situ testing and sampling topography borehole locations to ground truth geophysical data and check for anomalous soil properties, carefully assess soil spatial variability Gas exsolution Use water sampling probe (e.g., BAT) to measure dissolved gas, consider use of pressure sampler Surficial crust Conduct accurate shallow depth in situ testing, collect thick (∼1m) box cores (if possible) and test on vessel deck
High salinity Use corrected γw for computation σv0 , correct measured soil properties for salt content, could be indicator of overpressures Temperature Research ongoing, results may lead to changes development of data correction procedure Ice gouging, Select in situ test and soil sampling subsea relict borehole locations to ground truth permafrost geophysical data, carefully assess soil spatial variability Shallow gas and Use soil sampling boreholes to ground gas hydrates truth geophysical data, consider use of pressure samplers Cyclic loading Must collect undisturbed samples and conduct laboratory cyclic test program

Shallow gas and gas hydrates

Shallow gas and gas hydrates are hazards that can be encountered during site investigation and production drilling. Shallow gas can cause dangerous blow-outs during drilling operations (e.g. Lunne et al. 1996). The presence of gas hydrates, which have been found in many deep water regions, is a potential hazard because of the large volume expansion that can occur if they melt or dissociate. Geophysical methods are typically used during a site investigation program to detect the presence of shallow gas and gas hydrates. Ideally follow-on soil sampling should be conducted to ground truth findings from the geophysical surveys. This may require special pressure sampling equipment that can maintain in situ conditions in samples during their retrieval and storage (e.g. Kolk & Wegerif 2005). 3.10 Cyclic loading Cycling loading of offshore infrastructure can result from wind, waves, ice, and earthquakes. In production facilities such as pipelines, cyclic loading can be induced by temperature changes due to episodic flow of product. In soft cohesive soils, cyclic loading causes a reduction in effective stress due to pore pressure generation and hence a reduction in su . The behavioural response during cycling loading depends on the stress path and the combination of average and cyclic shear stresses. It is another level of complexity beyond that of the monotonic shear response of cohesive soils as covered in Section 2. Andersen (2004, 2009) presents comprehensive reviews of the cyclic behaviour of soils and provides guidance on testing and design applications.

3.12

Table 2 lists the unique and demanding offshore conditions discussed in this section. It also provides some comments that are a preview of the recommendations given in Section 5 for conduct of geotechnical site characterisation programs to evaluate these conditions.

3.11 Deep to ultra deep waters Site investigations in deep (>500 m) to ultra deep (>2000 m) water depths pose a multitude of logistical and soil testing and behaviour issues. Many of the challenges noted in the previous sections are exacerbated when working in deep to ultradeep waters. Some of these include: the regions are usually remote and thus have limited support from onshore facilities, handling of equipment can take a long time due to the large water depth, large hydrostatic water pressures may limit the accuracy of some in situ tools (Section 3.2), the degree of stress relief for samples is very large and there is a higher potential for gas exsolution (Section 3.4) and dissociation of gas hydrates (Section 3.9), the soil units are typically very soft (although sometimes a thin surficial crust exists Section 3.5), and the seabed temperature is often near 0◦ C in some regions (Section 3.7).

4

REQUIRED SOIL PARAMETERS FOR DESIGN AND ANALYSIS

This section lists soil parameters required for design and analysis of offshore infrastructure and assessment of geohazards involving cohesive soils. This information when coupled with a fundamental knowledge of soil behaviour and an awareness of the unique conditions that can often exist at offshore sites should be used to formulate the required outcomes from a geotechnical site characterisation program. Table 3 provides a list of soil parameters and examples of design concepts and geohazards for which these

42 © 2011 by Taylor & Francis Group, LLC

Summary

Table 3. Summary of cohesive soil properties required for various design concepts and geohazard analyses (modified after Andersen et al. 2008). Soil property

Requirements

Soil classification Index strength (e.g. FC, MLV, UUC, torvane, etc.)

– required in all cases – not required in many cases but almost always performed because tests are quick and easy to conduct – used in empirical design methods – required in all cases

/σv0 for some – need K0 = σh0 applications (e.g. allowable drilling pressure) but difficult to measure – CR for set-up analysis of some anchor systems and skirted seabed structures; analysis of pipelines and risers – CR and Cαε for settlement computations (e.g. gravity based structures) – set-up analysis of some anchor systems, skirted seabed structures, piles; analysis of pipelines and risers – settlement time rate computations (e.g. gravity based structures) – anisotropic su required for most applications (e.g. tip resistance of anchors, pile shaft friction, shallow foundation bearing capacity, jack-up spudcans, slope stability) – Gmax and anisotropic stress-strain behaviour required for slope stability, skirted seabed structures – sur required for penetration resistance of suction anchors; capacity of anchor systems and skirted seabed structures; analysis of flowlines, pipelines, risers and debris flow – thixotropy and reconsolidated sur required when calculating set-up effects (e.g. anchor systems, skirted seabed structures) – set-up analysis of some anchor systems (e.g. suction, torpedo) and skirted seabed structures

In situ stress state
– σv0 – u0 – σp
Compressibility – consolidation CR = ε/logσv
– creep Cαε = ε/ logσv


Intact soil flow properties – kv and cv – rk = kh /kv

Undrained shear strength – su anisotropy – Gmax and anisotropic stress-strain behaviour (including strain softening) Remoulded undrained shear strength – sur or St = su /sur – thixotropy – reconsolidated sur

Remoulded compressibility and flow properties – CR, kv and cv Cyclic properties – permanent/cyclic shear strains, cyclic pore pressures Dynamic properties – strength, modulus and damping Rate effects Temperature effects

Figure 18. Map of soil tests and design applications (from Randolph et al. 2007).

parameters are required. The information presented in this table is abstracted from the comprehensive design focused papers by Randolph et al. (2005) on challenges of offshore geotechnical engineering and Andersen et al. (2008) on deepwater geotechnical engineering. Clearly all site characterisation programs will involve tests necessary to classify the various soil units encountered. In terms of specific soil parameters, the
in situ stress state (σv0 , u0 , and σp
) is the most important and required in all cases. For cohesive soils, su is also required for almost all cases. The need for other parameters depends on the specific design concept and geohazard being analyzed. Some of the factors that influence clay behaviour have a compensating effect, although to variable and typically unknown degrees. It must be considered when linking results from laboratory and in situ tests with soil response conditions anticipated for a design application. Figure 18 illustrates this interplay for the undrained shear strength. It maps the relationship between the influence of remoulding (decreases su ) versus the influence of strain rate (increases su ) for lab and in situ testing relative to that of various offshore systems.

5

– required of all applications involving cyclic loading

Competent site investigation programs utilize an integrated approach that engages a multidisciplinary team of geo-specialists. This is especially important for assessment of complex geohazards as described for example by Jeanjean (2005), Kvalstad (2007), and Evans (2010). Geophysical, geological and geotechnical investigation methods are often conducted in a phased approach to take advantage of the best that each method offers and as project requirements evolve. A collective evaluation of the resulting data sets by a multidisciplinary team provides the greatest opportunity

– required for all applications involving earthquake loading – consideration required in most cases – required for all cases with soil temperature change > about 10 to 20◦ C

43 © 2011 by Taylor & Francis Group, LLC

BEST PRACTICE RECOMMENDATIONS FOR GEOTECHNICAL SITE CHARACTERISATION

detailing the requirements for the project. It is also important that the field and laboratory factual reports explicitly describe the equipment and test methods used.This documentation is especially important when multidisciplinary teams, which are often composed of members from different organizations and far removed from the work conducted, need to use the factual reports to interpret the results and develop design parameters. In terms of notation for soil parameters, examples of poor practice in site characterisation reports persist. One such example is reporting of su values. As discussed in Section 2, su is not a unique parameter and its value depends on how it is measured. It is therefore essential to explicitly indentify the measurement method used when reporting su values, be it laboratory measured, e.g. su (FC), su (CAUC), etc. or interpreted from in situ test data using N factors relative to a reference su measurement, e.g. Nkt,CAUC , NT-bar,CAUC , etc. (see Section 5.8 for definitions of these N factors). Recommendations: While publication of the proposed ISO standard on Marine Soil Investigations will not be a panacea, and it may not necessarily be thoroughly adopted worldwide, it will provide a muchneeded reference framework. Project specifications should be developed detailing required site investigation activities. Field and laboratory factual reports should thoroughly document equipment, procedures, and data processing methods used. Notation used for soil parameters should be more explicit than is often the case to avoid ambiguity.

for an effective and reliable site characterisation outcome. In the sections to follow, best practice recommendations are given for the geotechnical component of site characterisation. The focus is primarily on presentation of key recommendations with many of the details provided in the references cited. These recommendations should be tailored to the scope of the investigation (i.e. preliminary, intermediate or final), anticipated challenges (Table 2), applications being considered (Table 3), schedule, and budget. 5.1 Work scope Several non-engineering factors such as availability of equipment and personnel, time constraints, and budget, often dictate the final work scope. Nevertheless, proper planning of the investigation involves making decisions on the type, depth, and number of in situ tests and/or borings that should be planned for. Most often this is governed by the depth of stress influence of the design concept and soil spatial variability. In terms of depth, the extremes include pipelines for which only the upper few meters are of interest to analysis of large submarine landslides that may require borehole depths of 100 m or more. Recommendations: Specific recommendations on number and depth of boreholes are not given here because it is highly project dependent (i.e. application, location, soil variability, anticipated geohazards present, etc.). In principle, it is recommended that both in situ tests and high quality sampling, followed by laboratory testing, should be performed. SUT-OSIG (2000, 2004) and ISSMGE (2005) provide general guidance on developing exploration plans. 5.2

5.3 Vessels and deployment modes ISSMGE (2005) gives an overview of the type of platforms and vessels that can be used for offshore investigations. In shallow waters, the common options are anchored barges, jack-up rigs, and anchored vessels, which all must provide a stable platform. In deeper waters, options include specialized soil drilling vessels with a moon pool or geophysical survey vessels that handle geotechnical tools (i.e. soil sampling and in situ testing devices) using an A-frame, crane or winch. Use of dynamic positioning is more efficient than vessel anchoring systems in shallow to moderate water depths, and is required for deep water investigations. The draft ISO standard on Marine Soil Investigations (ISO 2010) defines deployment modes as being either “drilling mode” or “non-drilling mode” (traditionally referred to as downhole and seabed mode, respectively). In drilling mode, a borehole is advanced into the seabed using rotary drilling. At selected depths a geotechnical tool is lowered into the drill string and advanced into the seabed from the bottom of the borehole. The drilling can be performed from a vessel or stable platform (“vessel based drilling”) or from a seabed system (“seabed drilling”), which involves a remotely operated drill rig that is placed on the seabed. Non-drilling mode involves using geotechnical tools that are advanced from the seabed. This includes simple free fall devices or more sophisticated equipment

Standards, reporting and notation

There is a large body of standards that cover geotechnical site investigations with most of these developed for in situ and laboratory testing of terrestrial soils. These standards have been developed by various countries and organizations (e.g. ISO, BS, ASTM, etc.) and in some cases there is a significant difference amoung such standards for conduct of the same test. This creates opportunities for confusion and mistakes in using the resulting data. At present the only comprehensive standard that was specifically prepared for offshore investigations is NORSOK G-001 (2004). This standard was originally developed in the 1980s for projects offshore Norway but evolved over time with input from an international group of North Sea operators. In 2007, ISO formed an international committee to develop a new standard on Marine Soil Investigations (ISO 2010), using NORSOK G-001 as a basis. At the time this paper was written (May 2010) a draft of the full standard was completed and under review. In spite of standardization efforts, some standards leave considerable latitude on how to conduct a test and in some cases no standards exist. For each soil investigation, project specifications should be prepared

44 © 2011 by Taylor & Francis Group, LLC

Table 4.

that is landed on the seabed prior to commencement of testing. The important distinction here is that no borehole drilling takes place and thus the depth of penetration is limited. There is a large variety of nondrilling mode equipment being used and the quality of data obtained can vary significantly. More details and recommendations on this are given in Section 5.4 for in situ testing and Section 5.6 for sampling methods. The technology for seabed drilling systems continues to evolve with the development of increasingly versatile and sophisticated robotic units. For example, the Portable Remotely Operated Drill (PROD) as originally described by Carter et al. (1999), is currently used in commercial applications. Upon landing of the system on the seabed, rotary drilling, in situ testing, and sampling are controlled via robotic assembly of the drill string. Casing, rods, in situ tools, and samplers are stored on a carousel. In situ tools can be deployed directly from the seabed or at the bottom of a drilled borehole. Borehole depths in excess of 100 m have thus far been achieved using this equipment. In deep water, this type of seabed drilling system is more efficient than vessel drilling due to less handling of the drill string. Recommendations: Depth accuracy during drilling, in situ testing, and sampling is critical. For vessel based drilling, it is important to utilize a heave compensation system to stabilize the drill string. Good systems use the so-called “hard-tie” method that stabilizes the drill string against surface motions by using an installed seabed template as the reference point. Set-down of seabed systems and seabed templates can disturb the upper seabed and also apply a stress field to it. To minimize these problems, the set-down must be carefully controlled and the equipment should be designed to minimize imposed seabed stresses within the region of in situ testing and soil sampling (see Lunne et al. 2010).

Deployment method Vessel based drilling Seabed drilling Non-drilling mode – seabed system ROV mounted Free fall

Rod type and potential penetration depth (m) drop-in or wireline (>100 m) carousel w/ casing and drill string (75 m) or wireline (150 m) straight (40–50 m), carousel (40–50 m) or coiled (15–30 m) straight, coiled or split (all 3 m+) 15–20 (?) m

Figure 19. Schematic of recommended sequence for recording deck-to-deck reference readings for seabed in situ testing (from Lunne et al. 2010).

not necessary for coiled rod systems. Remotely operated vehicles (ROV) have the significant advantage of being able to maneuver to specific locations. The obvious disadvantage is limited thrust capacity and hence penetration depth; adding suction anchors to the ROV will increase capacity but such a system has not yet been deployed in practice. Free fall penetrometers are either expendable or retrievable with a cable system and are relatively easy to deploy. Recommendations: Long continuous push strokes are preferred for push-in tools such as the CPTU and full-flow penetrometers. It is also important when deploying these tools to monitor zero reference readings at all stages of the test as shown in Figure 19 (i.e. on deck prior to deployment, at sea bottom before and after push, back on deck). This deck-to-deck logging is considered an important quality control procedure and should always be conducted (Lunne et al. 2010).

5.4 In situ testing 5.4.1 In situ testing deployment systems Table 4 lists the various deployment systems that are available for in situ testing. Vessel based drilling with downhole in situ testing is conducted using pressurized drilling mud or a hydraulic system to advance the tool (the depth limitation is normally 500 to 600 m in the latter case). For tools such as the piezocone (CPTU), stroke lengths of up to 3 meters are possible. A disadvantage of using pressurized mud that does not use a data umbilical line is that data are stored on a memory module and cannot be viewed in real time. Seabed drilling systems that use pushing by a drill string have the advantage of being able to push an in situ tool 10 m or more from the seabed, or the bottom of the borehole. Seabed drilling systems that use wire line deployment are in development (e.g. Boggess & Robertson 2010) and should be able to test at greater depths, although the push stroke for in situ tools will likely be limited to about 3 m. Non-drilling mode seabed systems that use straight rods require the use of a constant tension winch or a tower to support the rods, which is

5.4.2 In situ tools Table 5 lists the main in situ tools available for site investigations with the CPTU being the most common tool. There is a large body of experience in using the

45 © 2011 by Taylor & Francis Group, LLC

Deployment systems for in situ testing tools.

Table 5.

In situ testing tools.

In Situ Test/ Measurements

Comments

Piezocone (CPTU) – tip resistance (qc ), sleeve friction (fs ), pore pressure (u), inclination

– most common tool – best to use differential measurement but equipment still in development

Seismic CPTU – same as CPTU + Vvh

– requires a seabed shear wave velocity source

Full-flow penetrometers T-bar, ball – intact and remoulded penetration resistance

– higher resolution than CPTU, minor overburden correction, cyclic testing can be used to measure shear strength degradation

Field vane – intact and remoulded shear resistance

– remoulded shear strength depends on number of rotations used to remould soil

Piezoprobe – in situ pore water pressure

– data often difficult to inter-pret (covered in Section 5.5) – collects pore water sample to analyze for dissolved gas

Deep water gas probe – water sample, temperature

standard CPTU it is a more sensitive device (for same capacity load cell). Full-flow probes are deployed in the same manner as the CPTU using the same penetration rate of 20 mm/s. Cyclic testing using short strokes (e.g. ± ∼0.5 m) at selected intervals during the penetration phase of the test allows for measurement of shear resistance degradation upon remoulding. The field vane test (FVT) has been used extensively in soft Gulf of Mexico soils. For terrestrial soils, it has been calibrated for use in analyzing stability problems to account for anisotropy, rate effects and insertion disturbance (Bjerrum 1973). However, in offshore practice, the Bjerrum correction factor is typically not applied (Kolk et al. 1988, Randolph 2004) and the measured results are directly used to report su (FVT). This is justified in part because the reduction in shear resistance due to insertion disturbance is partially compensated by the high strain rates applied by the test (Randolph 2004). Expendable penetrometers and recoverable free-fall penetrometers continue to be studied (e.g. Aubeny & Shi 2006, Mosher et al. 2007) with some promising results. However, more research is needed before it can be recommended as a reliable tool. Recommendations: Lunne et al. (2010) provide guidelines for use of the CPTU, full-flow penetrometers, and the FVT in deepwater soft clays. Most of these recommendations are also relevant to their use in shallow water depths. DeJong et al. (2010a) outline recommended practice for onshore full-flow penetrometer testing, much of which is also relevant for offshore practice.

CPTU and in interpreting the measured data to determine soil parameters (e.g. Lunne et al. 1997b). Saturation of the pore pressure measurement system is not a problem offshore, in contrast to terrestrial applications. However, there are accuracy problems when testing in deep waters because of the high hydrostatic pore water pressure acting on the sensors at the seabed – a problem that is exacerbated when testing soft sediments. Measuring the differential cone resistance and pore pressure would solve this problem but no such cone has yet been deployed in routine practice. Lunne (2010) points out that due to differences in design, the sleeve friction depends on the type of cone penetrometer used. This makes sleeve friction values less reliable compared to the tip resistance and pore pressure. The seismic CPTU uses geophones or accelerometers to measure shear wave velocity arrival times which can be used to estimate the small strain shear modulus via Equation 2. The shear wave source is usually deployed at the seabed. Most seismic CPTU systems use a single set of sensors and the depth specific travel time is estimated as the difference between the arrival times for successive push intervals (pseudo-interval). However, a more accurate set-up is to use two sets of sensors located 1 m apart in the cone to record the shear wave arrive at both locations simultaneously (true-interval). In recent years, full-flow penetrometers such as the T-bar and ball have started to be used in offshore practice. Because they allow the soil to “flow” around the device, the vertical stress correction (as required for full displacement devices such as the CPTU) is minimal and with an area typically 10 times that of the

– The CPTU is recommended as the main tool to be used for in situ testing. It is the best tool for soil profiling and identification of soil behaviour type. It can be used for estimating σp
and su but is not reliable for estimating sur . – The seismic CPTU should be used more often to measure Vvh (ideally using true-interval equipment), which can be used to estimate Gvh . – Full-flow penetrometers have proved to be as reliable as the CPTU for estimating su of soft clays and in some circumstances are better, e.g. shallow test depths and/or very soft sediments. They are the best option for measuring soil shear degradation and estimating sur . Both penetration and extraction resistance should always be measured when conducting full-flow tests. – The FVT remains a good tool for estimating su and sur (provided the soil is remoulded through multiple fast-rotations of the vane). However, the CPTU and full-flow penetrometer tests are quicker to perform and give a continuous profile of penetration resistance. 5.5

Determining the in situ pore water pressure condition is critical to conduct of a reliable site characterisation program. As described in Section 3.1, there are a number of offshore regions where overpressured zones

46 © 2011 by Taylor & Francis Group, LLC

Measurement of in situ pore water pressure

exist and have caused significant problems. Simply assuming hydrostatic pore water pressure conditions without verification is a risk. There are two options available at present for directly measuring in situ pore water pressure: piezoprobes and piezometers. 5.5.1 CPTU and piezoprobes The CPTU can be used to measure in situ equilibrium pore water pressure (u0 ) during a pause in penetration. Excess pore pressures that are generated during penetration will eventually dissipate back to u0 with the rate depending on cv . In clays with a low cv , this can take several days for the standard diameter CPTU (d = 36 mm) which is not practical for most site investigations. Since the rate of dissipation is proportional to d2 , small diameter piezoprobes have been developed to speed up dissipation. These probes consist of a tapered section that narrows down to a diameter of about 5 to 8 mm with a filter element near the tip. However, Whittle et al. (2001) showed that the pore pressure developed by the regular diameter push rod, to which the piezoprobes are attached, eventually reaches the smaller diameter tip and influences the dissipation time (and it can take as long as with a regular diameter CPTU). Whittle et al. (2001) thus developed an interpretation method that uses data from a dual element piezoprobe which consists of two pore pressure filters, one at the small diameter tip and one at the normal location on the shoulder of the cone (Figure 20a). The simultaneously recorded dissipation data are interpreted to estimate u0 . Piezoprobes have successfully been used in site investigations but unresolved issues remain. The lengthy, small diameter extension, which ideally should be as long as possible to reduce influence of the larger diameter push rod, is vulnerable to damage during penetration. When deployed in vessel drilling mode, imperfect heave compensation of the drill string can produce erratic results that are difficult to interpret. The use of a dual-element piezoprobe and the corresponding interpretation theory of Whittle et al. (2001) are still been researched (e.g. Flemings et al. 2008) and have not yet been validated for use in commercial practice.

Figure 20. a) T2P dual element piezoprobe (from Flemings et al. 2008), and b) example of multi-piezometer string installed in borehole (from Strout & Tjelta 2007).

Figure 21. Pore water pressure measurement options for piezometers (from Strout & Tjelta 2007).

and the soil formation. This is especially important for shallow depth installations. Three methods can be used to measure the in situ pore pressure at the depth of installation of the filter element (Figure 21, Strout & Tjelta 2007): a) direct measurement of u0 at the filter; b) differential pressure measurement at the filter relative to the seabed; and c) differential pressure measurement at the seabed. Practical considerations include saturation of the connecting fluid column for differential pressure measurements and the density of this fluid column needs to match that of the in situ pore water, sensors must be protected against corrosion, and the logistics of collecting data from the instrumentation package. Recommendations: Strout & Tjelta (2007) conclude that the most reliable method for measuring in situ pore pressure in low permeability sediments is to use piezometers. The key advantage is that they provide time history data. The cost is not trivial and new

5.5.2 Piezometers Piezometers can now be installed as part of a soil investigation in a predrilled borehole (i.e. drilling mode) or via non-drilling mode using a seabed push frame. In drilling mode, the creation of a borehole allows for the installation of a single piezometer or a string of piezometers. The piezometers are grouted in place and an instrumentation package is connected to the top of the string at the seabed (Figure 20b). An important practical consideration is the effectiveness of the grout, which must create a stable hydraulic seal around the piezometers with a hydraulic conductivity that is equal to or less than that of the formation soil. For push-in piezometers, the push stroke must be straight and concentric so as to not potentially create an open void, and thus a short circuit between the piezometer rise pipe

47 © 2011 by Taylor & Francis Group, LLC

Table 6.

solutions are being developed and this effort needs to continue. Seabed sediment overpressures was the focus of a 2009 workshop (Sheahan & DeGroot 2009) that identified challenges in dealing with overpressures. For current practice, workshop recommendations were made for measurement and installation options as follows:

Soil sampling equipment.

Deployment

Sampler and diameter (mm)

Non-drilling, DWS 110 seabed mode

– for shallow depth (<∼30 m below seabed or up to depth possible to penetrate from the seabed using a seabed system) push-in piezoprobes or piezometers can be used. They should be deployed at the beginning of an investigation to immediately start collecting data. Provisions could be made to retrieve the units at the end of the investigation, or later on if possible, for reuse. – for investigations using drilling mode (i.e. deep investigations), options depend on whether a piezoprobe can be pushed without damage; if not then piezometers are the only option. If so, both options are a consideration. Piezoprobes have the potential to give immediate data (depending on soil type, ship time available for waiting during dissipation). If time history data is required, then piezometers must be used. – advanced seabed drilling platforms, such as the PROD, provide new opportunities for more efficient installation in deep water, and to significant depths below seabed, as described by Tjelta and Strout (2010). – it is probably not yet possible with complete certainty to measure u0 with an accuracy greater than
10% of σv0 .

Potential penetration m Comments 20–25

Piston fixed relative to seabed, penetration measured reliably, underpressure measured behind piston

STACOR 105

20

Piston fixed relative to seabed, penetration not measured reliably

Kullenberg 80–110

3 - >50

Piston relative to vessel, problems with recovery and sample quality

Box corer typically 500 mm square

0.5–1.0

Miniature in situ tests can be conducted in box upon retrieval, subsamples can be collected

Seabed drilling

Piston or push, 44/75 Percussion or rotary, 75

75–150

Depth depends on system casing/rod storage capacity with wireline systems thus far being able to go deepest

Vessel based drilling

Piston, push, >100 or percussion 75

Advanced by push, drive or percussion using mud pressure or hydraulics with power from an umbilical.

5.6 Soil sampling, preservation and storage 5.6.1 Sampling methods Table 6 lists the different types of soil sampling equipment presently available. Section 2.2 discussed the issue of sample disturbance and its impact on the reliability of laboratory measurement of design parameters such as σp
and su . When such advanced laboratory tests are to be conducted it must be a key objective of the site characterisation program to use a sampler that can collect good quality samples. Lunne & Long (2006) outlined the key features that an offshore sampling system should have in order to have the potential to collect good quality samples. These include (Figure 22): sharp cutting shoe, stationary piston relative to the seabed, minimal inside and outside friction, small area ratio, sample diameter of 100 to 120 mm, and a retractable core retainer. In terms of deployment, the sampler should be pushed at a steady rate of ∼20 mm/sec, the push rate should be measured, and the underpressure developed behind the piston should also be measured. The Deep Water Sampler (DWS) was developed with the features displayed in Figure 22 as a goal for its design and deployment (Lunne et al. 2008). It has thus far been used on several onshore and offshore investigations, including the Troll site (Norwegian North Sea), and has shown a capability to collect good to

© 2011 by Taylor & Francis Group, LLC

Figure 22. Design features of ideal sampler for soft to medium-stiff clays (from Lunne & Long 2006).

high quality samples (Lunne at al. 2008). The depth of these investigations was less than 10 m and its good performance remains to be verified for the design full penetration depth of 20 to 25 m. The STACOR gravity sampler (Fay et al. 1985) uses a piston that is fixed relative to a small frame that sets-down on the seabed during deployment. Thus if the frame remains stationary during sampling then so

48

does the piston. Penetration is not directly measured. Numerous cases have reported collecting good quality samples using the STACOR sampler (e.g. Borel et al. 2005). The Kullenberg type of sampler also penetrates the seabed by gravity and can collect very long samples of up to 50 m or more, which can be an efficient way of collecting long sample cores. However, there are problems with this type of sampler including uncertain penetration depth and low recovery, and it has been found that sample quality tends to decrease with depth. These latter two problems are largely a result of the use of a thick walled, blunt tipped, cutting shoe and the fact that the piston is fixed with a wire to the vessel and not positioned relative to the sea bottom. Downhole sampling in vessel based drilling mode has long been shown to be able to collect good quality samples providing a fixed piston sampler is used in conjunction with a thin walled sample tube having a sharp cutting edge (e.g. 5◦ for soft clays), small area ratio and zero inside clearance ratio, and furthermore that the drilling is carefully conducted with an efficient heave compensation system. As noted in Section 5.3, several multipurpose seabed drilling systems have been developed. For the PROD system, which uses a drill string to deploy a sampler, potential sampling depth depends on the number of rods the carousel can hold (which is a function of the diameter casing being used). Box cores are good for characterizing the upper 0.5 m of the seabed. Common box cores normally collect a 0.5×0.5×0.5 m sample and can be tested immediately after retrieval to the vessel. For soft soils, miniature laboratory vane and full-flow probes can be used to test the sample while it is still in the box (e.g. Low et al. 2008). Because the effective stress of these surficial sediments is quite low, it is generally not possible to conduct reliable advanced laboratory tests on these samples; thus, the use of these miniature probes is a good, practical solution. Figure 23 shows an example of such data recorded for a soft seabed sediment. Subsamples can be collected for follow-on onshore laboratory testing. The large volume of collected soil can be saved from multiple box cores and resedimented in the laboratory for scale model testing of shallow soil structure interaction problems as pipelines and risers (e.g. 1g model tests described by Andersen et al. 2008). Recommendations: Handling of long recovery samples is difficult and the vessel must be equipped to handle such samples without disturbance. Box core testing using miniature full-flow probes and MLV is recommended for characterizing the upper (∼0.5 m) sediments for shallow depth investigations. Tube samplers are ranked as follows relative to their potential to obtain as good a quality sample and as high a recovery ratio as possible:

Figure 23. a) Penetration, extraction and cyclic resistance versus depth for miniature T-bar test conducted in box core sample of a soft seabed sediment. b) degradation plot for cycle 2.

– For non-drilling mode seabed based sampling: 1) piston sampler fixed relative to seabed, favorable geometry, steady penetration with real time measurement of penetration, and accurate measurement of recovery (e.g. Figure 22, DWS), 2) gravity sampler with piston fixed relative to seabed (e.g. STACOR), 3) gravity sampler but piston fixed relative to vessel (e.g. Kullenberg), and 4) gravity sampler without fixed piston. 5.6.2 Sample handling and storage Sample handling and storage must avoid handling disturbance and must maintain sample moisture and temperature conditions. Samples should not be subject to excessive temperature changes from warm climates, overheated wax and especially freezing temperatures. This is especially important for all samples that may be used for advanced laboratory testing (Section 5.7) Evidence suggests that soft clay samples can survive commercial transport when properly protected (and not subjected to catastrophic impact such as dropping of the sample box). For example, Figure 24 plots results of constant rate of strain consolidation (CRS) tests performed on specimens trimmed from Sherbrooke block samples of the Onsøy clay at NGI in Oslo, Norway and again at UMass Amherst, USA after commercial air transport.

– For downhole sampling in drilling mode (vessel or seabed based): 1) thin walled fixed piston sampler with favorable geometry as noted above; 2) thin walled push sampler (although sampling without a fixed piston is not recommended).

© 2011 by Taylor & Francis Group, LLC

49

Table 7.

Figure 24. Comparison of CRS tests conducted on Sherbrooke Block samples of Onsøy clay at NGI, Oslo, Norway and again after air transport to UMass Amherst, USA (from DeGroot et al. 2009).

Recommendations: Sample tubes/liners should be sealed with a mechanical seal and/or with wax and capped. The samples should be stored vertically and should not be able to move inside the sample tube or liner. Soft clays should be kept in the tube or liner for transport to the onshore laboratory. For stiffer clays, the sample may be extruded in the field and sealed in a cardboard cylinder using plastic, aluminum foil and wax. Samples should as quickly as possible be placed in a constant temperature controlled container that has been set at the estimated in situ temperature. Samples should be transported upright in containers that damp shock and vibration. Once at the onshore laboratory, samples should be stored in a container or room with humidity and temperature control.

Test Category

Test Types

Comments

Classification and Basic Index Testing

– – – – –

– simple and relatively quick to perform – necessary part of any site characterisation program – cannot provide design parameters

Express/Index Strength Testing Parameters measured: su , sur

– pocket penetrometer – fall cone – torvane – laboratory vane, – unconfined compression – unconsolidated undrained triaxial

– simple and relatively quick to perform – equipment commonly available – often gives scattered results – successful use in design requires soil/ site specific correlations

Advanced Laboratory Testing Parameters measured: CR, Cαε , σp
, cv , kv , su , sur , St c
, φ
, Gmax

– IL and CRS Oedometer – CAUC, CAUE & DSS – ring shear – permeability – resonant column – cyclic CAUC & DSS – bender elements

– provides best control of soil state and conditions in the laboratory – relies on good quality samples – provides direct measure of design parameters – equipment more complex but automation very efficient

water content density Atterberg Limits grain size specific gravity

Advanced laboratory tests, as listed inTable 7, are an essential component of any site characterisation program requiring design parameters for compressibility, consolidation/flow, and strength behaviour. The need for such testing is highlighted in Table 3 for the various design concepts and geohazards listed. Section 5.7.3 to follow gives an overview of recommended scope and methods for conduct of advanced laboratory test programs. For some projects, geochemical an dating tests may also be required for assessing the geologic origin and history of sediments.

5.7 Laboratory testing Table 7 divides laboratory tests into three primary categories: classification and basic index testing, index strength tests, and advanced laboratory tests. Every design concept will first and foremost require classification of the various soil units encountered and such testing should always be conducted as part of any site investigation. The express/index strength tests listed in Table 7 are popular because they are generally quick, easy to perform and can readily be conducted offshore. However, it is important to note that these tests generally use fast shear rates, different modes of shear, and test small soil volumes, and the results are greatly affected by sample disturbance. Undrained shear strength profiles developed using these devices often show significant scatter. Therefore, the data from these devices represent, at best, relative strengths rather than values suitable directly for design. They should be relied upon only to indicate the general consistency of soil layers. Reliable determination of su values for design should focus on appropriate in situ tests and use of laboratory equipment that can conduct consolidated-undrained (CU) tests.

5.7.1 Field laboratory For logistical reasons, offshore laboratory testing is usually limited to soil description, classification and index strength tests. In some cases it can be advantageous to conduct CRS tests offshore in order to get an early indication of stress history, which provides a preliminary indication of the soil state and aids in planning the advanced laboratory test programme. The CRS data can also be used to evaluate sample quality as described in Section 5.7.2 to follow. If box cores are collected they should also be tested offshore as described in Section 5.6.1. 5.7.2 Evaluation of sample quality While no definitive method exists for determining sample quality, valuable information can be obtained

50 © 2011 by Taylor & Francis Group, LLC

Laboratory tests for fine grained soils.

Table 8. Evaluation of sample quality for low to medium OCR clays using Lunne et al. (2006).  e/e0 at in situ stresses for sample qualities 1 to 4 OCR

1 (very good to excellent)

2 (fair to good)

3 (poor)

4 (very poor)

1 to 2 2 to 4

< 0.04 < 0.03

0.04 – 0.07 0.03 – 0.05

0.07 – 0.14 0.05 – 0.10

> 0.14 > 0.10

from making use of both qualitative and quantitative methods. Visual observations on the vessel during sample processing and testing in the offshore laboratory can provide useful qualitative information on sample quality (e.g. evidence of cracking). Qualitative (visual) assessment of sample quality is best made by examination of sample X-rays. Radiographic methods allow for non-destructive visual evaluation of sample quality, layering and the presence/quantity of stones, cobbles and other inclusions. It is also possible to obtain computerized tomography (CT) scans of soil samples that are contained in non-metallic sampling tubes or liners. CT scans provide a visual image along the length of a sample and cross-sectional images.The sample radiographs and/or CT scans should be used to select the location within a tube or liner sample of the most critical advanced laboratory tests. Quantitative assessment of sample quality for intact, low to medium OCR clays, can be done by measuring volume change during laboratory recon
solidation to the estimated in situ stress state (σv0 ,
σh0 ). The normalised sample quality parameter e/e0 of Lunne et al. (2006) is computed as

Figure 25. Example application of e/e0 sample quality evaluation for three samplers used at Troll field (from Lunne et al. 2008).

where e = change in void ratio during reconsolidation to in situ stresses, e0 = initial void ratio, and εvol = volumetric strain (= V/V0 ) from reconsolidation to in situ stresses. These data are available from 1-D consolidation tests (IL and CRS) and the consolidation phase of CU strength tests (triaxial and DSS). The corresponding sample quality is determined using Table 8. Figure 25 plots an example application for samples collected from the Troll field using the DWS, borehole drilling with tube sampling, and a gravity core sampler. The technique of evaluating sample quality using shear wave velocity continues to be developed (e.g. Landon et al. 2007, Donohue & Long 2007). The method involves measuring Vvh of a short sample section using bender elements mounted in a jig such as the one shown in Figure 26. The advantage of this method is that by using small strain shear waves generated by the bender elements, the measurement is nondestructive and can be performed immediately after sample collection. Measurements can also be repeated on the same samples again in the onshore laboratory to potentially track any additional disturbance from sample sealing, transport and storage.

Figure 26. Schematic of device for combined sample shear wave velocity and soil suction measurements.

Figure 27 shows the proof of concept for terrestrial soils by comparing Vvh normalised by the in situ Vvh , in this case measured with downhole seismic CPTU, versus e/e0 . Ideally this method could be used to quickly and non-destructively screen samples on the vessel and again later on in the onshore laboratory prior to making decisions on selection of samples for

51 © 2011 by Taylor & Francis Group, LLC

that are important to proper and reliable conduct of these tests. Lunne & Andersen (2007) describe an approach used at NGI for large testing programs to optimize the use of in situ and laboratory testing, and when su anisotropy data are required. A brief summary is as follows: 1) select a key representative borehole and conduct sufficient number of CAUC tests on the best quality samples to establish a reliable su (CAUC) profile, 2) in the same borehole conduct some CAUE and DSS tests to establish anisotropy ratios, 3) correlate su (CAUC) values with the results of CPTU and/or full-flow tests to develop a site specific correlation for the continuous in situ data (generic correlations for these in situ tests are discussed in Section 5.8), and 4) if possible conduct a few CAUC and DSS tests in other borings. When selecting the location of a test specimen, soil from within 1 to 1.5 times the tube/liner diameter from the tube top and bottom should not be used because of greater disturbance near the sample ends (Lacasse & Berre 1988). Sample sides should be trimmed during specimen preparation to remove potentially disturbed material. Sub-sampling should not be used for preparation of final specimen dimensions. Very soft clays will need to be supported during trimming and mounting without touching the specimen by hand during preparation. The best test to measure compressibility, flow and σp
is the CRS. The CRS test is preferred over incremental loading (IL) consolidation tests because of the ability for continuous measurement of deformation, vertical load, and pore pressure for direct calculation of the stress-strain curve, kv and cv . The loading strain rate should be selected such that normalised base excess pore pressure (ub /σv ) is less than about 15% in the normally consolidated stress range. Sandbækken et al. (1986) report that rates of 0.5 to 1%/hr are adequate for most clays. Experience at NGI indicates that σp
from CRS tests gives a value about 10 to 15% greater than that determined from IL tests that use 24 hr load increments. The CRS tests should be conducted as early as possible. Stress history information is important for preliminary assessment of various design concepts and is also for conduct of the CAUC/E and DSS strength tests. The CRS results will also give a direct indication of sample quality via the e/e0 method described in Section 5.7.2 and Table 8. Evaluation of undrained shear strength (su ) anisotropy is often a critical aspect of many offshore design concepts as discussed in Section 2.1.2 and listed in Table 3. In the laboratory, this is best evaluated using a combination of CAUC, DSS and CAUE test. This is especially important for low to medium plasticity soft clays for which the undrained shear strength anisotropy is typically the most significant (e.g. Figures 2–4). The rate of strain for undrained shear should be selected to account for strain rate sensitivity of clays and typical field loading rates. Recommended rates include 0.5 to 2.0% per hour for triaxial tests and 5%/hr shear strain for DSS (Germaine & Ladd 1988, Lacasse & Berre 1988).

Figure 27. Normalised sample shear wave velocity versus CRS measured e/e0 for samples of three soft clays (from DeGroot et al. 2007).

advanced laboratory testing. Once samples are tested, confirmation e/e0 data can be collected during the consolidation phase. The method does require in situ Vvh , with downhole seismic CPTU being the best option (Section 5.4.2). Measurement of sample suction may be used as an additional refinement of this non-destructive sample evaluation scheme. Suction in a clay sample is the negative equivalent of the sampling effective stress (σs
in
Figure 7) and its value relative to σv0 is an indication of sample disturbance. Measuring both Vvh and σs
could be used in the framework shown in Figure 8 or that presented by Donohue & Long (2007) to assess sample quality. Sample suction can be measured using a portable suction probe such as that described by Ridley & Burland (1993). The suction probe in Figure 26 at the base of the bender element jig is based on that of Poirier & DeGroot (2010). While these non-destructive methods of evaluating sample quality show promise, more research and development is needed before it can be recommended for regular use in offshore practice. 5.7.3 Advanced laboratory test procedures The most important advanced laboratory tests for characterisation of clays include the CRS test for measurement of compressibility, flow and stress history, and CAUC, DSS, and CAUE for measurement of stress-strain-strength behaviour and anisotropy. Other measurements listed in Table 3 such as small-strain behaviour, remoulded behaviour, and cyclic response are also important in many cases but the CRS and CAUC/E and DSS tests are critical to almost all applications. The following discusses procedure issues

52 © 2011 by Taylor & Francis Group, LLC

Ideally triaxial and DSS specimens should be anisotropically consolidated to the estimated in situ

effective stress state (σv0 , σh0 ), i.e. the Recompression method (Bjerrum 1973). In the absence of measured K0 data, the Brooker & Ireland (1965) correlation amoung K0 , PI and OCR can be used to estimate K0 and compute the laboratory consolidation stress

σhc ∼ σh0 . The quality of su data from Recompression tests depends on sample quality. Increasing sample disturbance can destroy structural bonds and subsequent loss in measured su ; however, large volume changes during reconsolidation can result in an increase in measured su . If sample quality is found to be poor to very poor then consideration should be given to conducting SHANSEP (Ladd 1991) tests, which involves laboratory K0 consolidation to a OCR = 1 state of stress and then to various mechanically overconsolidated stress states. Otherwise the SHANSEP testing protocol (e.g. evaluation of anisotropy, shear rates) is the same as that of the Recompression method. The advent of reliable automated triaxial stress path cell systems greatly enhances the efficiency of conducting K0 consolidation. SHANSEP triaxial and DSS test data for each major soil unit are used to evaluate the S and m parameters for Equation 1 and the best estimate OCR profile is used in Equation 1 to compute the in situ su profile. For under-consolidated and normally consolidated clays, OCR = 1 SHANSEP testing is the only viable option. Ladd (1991) and Ladd & DeGroot (2003) provide further details on the Recompression and SHANSEP methods and describe the advantages and disadvantages of both. For laboratory measurement of sur , the motorized laboratory vane (MLV) is the most versatile device. It covers a large range of shear strength, can use different rates, and can measure shear resistance of samples in various states. Several measures of the sur are possible using the MLV as discussed in Section 2.3 and it is important that the method(s) required or used be clearly documented. The fall cone (FC) is also recommended to measure sur . While not as versatile as the MLV, there is a large amount of experience with the FC, it can cover a large range of shear strength (through the use of different size cones), and uses only a relatively small volume of soil. However, calibration factors for converting FC penetration to sur differ among standards (e.g. ISO, Norwegian, and Swedish standards use different factors), and thus the calibration factor used should be documented. Measurement of sur using UUC tests is not recommended if sur is less than about 5 kPa because the tests are difficult to perform and the membrane and area corrections are significant. The UUC test is better suited for stiffer soils and has the advantages of being suitable for investigating rate effects and to investigate sur anisotropy. Recommendations: Major recommendations for conduct of advanced laboratory tests include:

– it is essential to evaluate sample quality (Table 8) and the e/e0 value should be reported for all CRS (or IL) and consolidated strength tests – 1-D compressibility, flow, and σp
are best determined using CRS tests – CAUC/E and DSS tests should be used for measurement of stress-strain-strength behaviour and su anisotropy. For good quality samples, specimens should be consolidated to the estimated in situ effective stress state. – for poor to very poor quality samples the triaxial and DSS test program should involve conduct of SHANSEP tests. SHANSEP testing should be used for under-consolidated and OCR = 1 soils. – sur is best measured using the MLV, and the method of remoulding must be specified. Companion tests may be run using the FC. – Andersen et al. (2008) give recommendations for cyclic, thixotropy, creep, and reconsolidated remoulded testing and present example results. 5.8

5.8.1 Stress history recommendations
profiles are developed using soil total unit In situ σv0 weight measurements and estimated in situ pore pressures. In the early stages of a site investigation, it is common to assume that the in situ pore pressure profile is hydrostatic with a sea water unit weight γw = 10.1 kN/m3 . However, measurement of in situ pore pressure (Section 5.5) and pore fluid salinity will be required in regions with non-hydrostatic pore pressures and high salinity. There are many methods of estimating σp
using CRS or IL data and the K0 consolidation phase of SHANSEP tests with no method being definitively better than any other. The authors typically use two to three methods including Casagrande, constrained modulus (Janbu 1969) and strain energy (Becker et al. 1987) and report values from all methods in the factual
laboratory report. The Andresen et al. (1979) su /σv0 , PI and OCR correlation is also often used. CPTU data can be used to estimate σp
using the correlation (Lunne et al. 2007b).

where qnet = CPTU net penetration resistance and k is typically within the range of 0.25 to 0.35. These data should be used with the laboratory σp
values and geologic history information to develop a best estimated in situ σp
profile. This profile should then be used to compute the corresponding OCR profile. If the

– samples should be X-rayed to check for features such as layering and visual signs of disturbance

53 © 2011 by Taylor & Francis Group, LLC

Data evaluation and interpretation

The following presents recommendations for data evaluation and interpretation with a focus on stress history and su . The evaluation of sample quality should be used to determine if σp
and su . from the advanced laboratory tests are reliable. Generally it is desired to use data from tests conducted on samples of e/e0 quality rating 1 or 2 (Table 8).

Table 9. Recommended soft clay CPTU and T-bar N-factors for evaluation of su and sur (modified after Low et al. 2010). N-factor N factor, Nrem factor Nkt,CAUC Nkt,suave Nu NT-bar,CAUC NT-bar,suave NT-bar,rem,UU NT-bar,rem,fc NT-bar,rem,MLV

Definition

Mean

Range

qnet /suCAUC qnet /suave or qnet /suDSS (u2 – u0 )/suCAUC qT-bar /suCAUC qT-bar /suave or Tbar /suDSS qT-bar,rem /sur,UU qT-bar,rem /sur,fc qT-bar,rem /sur,MLV

12.0 13.5

10.0–14.0 11.5–15.5

6.0 10.5 12.0

4.0–9.0 8.5–12.5 10.0–14.0

20.0 14.5 14.0

13.0–27.0 12.5–16.5 12.0–16.0

Notes: su N values for clays with St < 8. Use NBall = NT-bar . Table 10. Typical SHANSEP S and m values for su (ave) (modified after Ladd and DeGroot 2003). Soil Description

S

m

Sensitive cemented marine clays (PI < 30%, LI > 1.5) Homogeneous sedimentary clays of low to moderate sensitivity (PI = 20 to 80%) Sedimentary silts and organic soils (Atterberg Limits plot below A-line) and clays with shells Lacustrine varved clays

0.20

1.0

0.22

0.8

0.25

0.8

0.16

0.75

Figure 28. Undrained shear strength data for low OCR soft clay site, Haltenbanken area of the Norwegian Sea.

as those presented in Figures 2–4, are also useful in evaluating the data. For low OCR soils, reference lines of minimum expected su ratios should be plotted, i.e. assume OCR = 1 in Equation 1 and select an S value for the su of interest [e.g. Table 10 presents typical S and m values for su (ave)]. The recommendations given in this paragraph are for evaluating su data and not intended to be a substitute for in situ and laboratory testing. Figure 28 plots an example data set for a lightly overconsolidated, soft, clay site in the Haltenbanken area of the Norwegian Sea. Site and soil layer specific CPTU Nkt factors were developed in reference to the su (CAUC) tests which were conducted on good quality samples. The most notable aspect of the data is the scattered su values from the strength index tests and that many of these values are very low. This result is
represents a lower a common occurrence. The 0.28σv0 bound su (CAUC) reference line assuming OCR = 1
). (i.e. Figure 2 with σp
= σv0

Note: PI = plasticity index, LI = liquidity index

σp
data are considered reliable, then a site specific k coefficient should be developed. 5.8.2 Undrained shear strength recommendations The CPTU and full-flow penetrometer data can be converted to su using an appropriate N factor

N values are specific to the reference su and in situ test conducted. Low et al. (2010) and DeJong et al. (2010b) present recommended N factors for interpretation of CPTU, T-bar, and ball data for soft clays. Table 9 presents the Low et al. (2010) factors, which were developed using a database that included several offshore sites. The laboratory su data should be plotted together with that estimated from the in situ tests using the N factors in Table 9. Depending on the quality of the laboratory su data, site specific N values should be developed. The su profiles should always be evaluated in conjunction with the stress history data. As described in Section 2.1.1, su and σp
are strongly linked and Equation 1 can be used to help in evaluating the consistency of the su profiles relative to the σp
profile. Andresen
et al. (1979) present correlations amoung su /σv0 , PI and OCR based on Equation 1. Database correlations, such

6

Many innovative systems and test methods have been developed for geotechnical characterisation of offshore sediments. Significant progress has been made in implementing improved deployment systems, in situ tools, and samplers in practice. However, challenges remain, some of which are a matter of convincing practice to implement more widely what are already known best practice methods. For example, sample quality is

54 © 2011 by Taylor & Francis Group, LLC

PRESENT AND FUTURE CHALLENGES

so critical to the reliability of laboratory data that the use of proper sampling tools must be made a priority. Another example is project specifications, which must be developed with an understanding of appropriate procedures and standards to specify. This largely involves clear communication amoung all parties and use of proper QA/QC programs. Universal adaptation of the proposed ISO standard on Marine Soil Investigations will provide a valuable reference framework. At the other end of the spectrum, some challenges are more daunting, such as for example developing geophysical techniques that yield reliable soils parameters. Other important challenges and related questions include:

the quality and reliability of in situ and laboratory test data. The paper describes these phases and gives best practice recommendations for each. Some of the most important recommendations include: accurate depth control relative to the seabed during deployment of in situ and sampling tools, CPTU and full-flow penetrometers for in situ testing, piezometers for measurement of in situ pore pressure, non-drilling seabed sampling with a piston sampler that is fixed relative to the seabed and with an appropriate geometry for sampling of cohesive sediments, evaluation of sample quality, and conduct of CRS, CAUC/E and DSS tests to measure consolidation, stress history and anisotropic stress-strain-strength behaviour.

– implementation of new seabed based drilling systems is an exciting development and the innovation needs to continue. Their use should be more widespread, especially for deep waters. – implementation of hydrostatically compensated CPTU for deep water investigations and increased use of seismic CPTU to obtain small strain shear modulus profiles. – in situ pore pressure u0 : 1) piezometers should become standard practice for projects where knowledge of u0 is essential, 2) develop inexpensive reliable drop-in piezometers, 3) develop reliable indirect indicators of overpressures, e.g. could geophysical measurements be used? – characterisation of the upper 1 to 2 m of sediment, especially in deep waters. Is box core testing good enough and how to evaluate the quality of a box core sample? How accurate are in situ tools in characterizing this surficial zone? – in situ testing of intermediate soils such as silts that can undergo partial drainage during testing. Can variable rate testing become a reliable and practical approach for offshore investigations? – do ambient pressure and temperature samplers need to be used more often or can we compensate for stress relief and temperature change in laboratory measurements? 7

ACKNOWLEDGEMENTS The authors thank their colleagues at UMass Amherst, NGI and Statoil for their contribution to the numerous research and consulting projects from which results and findings are presented in the paper, and Thomas Sheahan who reviewed the manuscript.The first author thanks NGI and the University of Western Australia for providing sabbatical opportunities to work on offshore site characterisation and acknowledges the US National Science Foundation for its support on grant OISE-0530151.

REFERENCES Allen, J.D., Hampson, K., Clausen, C.J.F., & Vermeijden, C. 2005. Well deformations at West Azeri, Caspian Sea. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 999–1004. Andersen, K.H. 2004. Cyclic clay data for foundation design of structures subjected to wave loading. Proc. Int. Conf. on Cyclic Behaviour of Soils and Liquefaction Phenomena, CBS04, Bochum, Germany, 371–387. Andersen, K.H. 2009. Bearing capacity under cyclic loading – offshore, along the coast and on land. 21st Bjerrum Lecture presented in Oslo, 23 November 2007. Can. Geotechnical J. 46: 513–535. Andersen, K.H.A., Lunne,T., Kvalstad, T. & Forsberg, C.F. 2008. Deep water geotechnical engineering. Proc. XXIV Nat. Conf. of the Mexican Soc. of Soil Mechanics, Aguascalientes, 26–29 November, 2008. Andresen, A., Berre, T., Kleven, A. & Lunne, T. 1979. Procedures to obtain soil parameters for foundation engineering in the North Sea. Marine Geotechnology, 3(3): 201–66. Aubeny, C.P. & Shi, H. 2006. Interpretation of impact penetration measurements in soft clays. J. Geotech. and Geoenvironmental Eng. 132(6): 770–777. Becker, D. E., Crooks, J.H.A., Been, K. & Jefferies, M.G. 1987. Work as a criterion for determining in situ yield stresses in clays. Can. Geotechnical J. 24: 549–564. Bjerrum, L. 1973. Problems of soil mechanics and construction on soft clays. Proc. 8th Int. Conf. on Soil Mech. and Foundation Eng., Moscow, 3: 111–159. Boggess, R. & Robertson, P.K. 2010. CPT for soft sediments and deepwater investigations. Proc. 2nd Int. Symp. on Cone Penetration Testing, Los Angeles. 9–11 May 2010.

CONCLUDING REMARKS

This paper presented best practice recommendations for geotechnical characterisation of offshore cohesive sediments. The particular focus has been on clays which have a complex mechanical behaviour, are not easily characterized, and are vulnerable to irreversible disturbance when sampled. This complexity is exacerbated for offshore investigations because of the many unique and challenging conditions that are present offshore. Many tools and methods have been developed to meet these challenges. Competent geotechnical site characterisation programs should combine the best of in situ testing and laboratory testing, i.e. using well calibrated in situ tools and collection of high quality undisturbed samples for advanced laboratory testing. Each phase of the site characterisation process, from drilling to evaluation of laboratory data, can influence

55 © 2011 by Taylor & Francis Group, LLC

International Standards Organization. 2010. Marine Soil Investigations. ISO TC 67/SC 7 N. In preparation. Jamiolkowski, M., Ladd, C.C., Germaine, J.T. & Lancellotta, R. 1985. New developments in field and laboratory testing of soils. Proc., 11th Int. Conf. on Soil Mech. and Found. Eng., San Francisco, 1: 57–154. Janbu, N. 1969. The resistance concept applied to the deformation of soils. Proc. 7th Int. Conf. on Soil Mech. and Found. Eng., Mexico City. 1: 191–196. Jardine, R.J. 1992. Some observations on the kinematic nature of soil stiffness. Soils and Foundations. 32(2): 111–124. Kay, S., Goedemoed, S.S. & Vermeijden, C.A., 2005. Influence of salinity on soil properties. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 1087–1093. Kolk, H.J., Hoope, J. & Imms, B.W. 1988. Evaluation of offshore in situ vane test results. Vane Shear Strength Testing of Soils: Field and Laboratory Studies, ASTM STP 1014: 339–353. Kolk, H. & Wegeriff, J. 2005. Offshore site investigations; New frontiers. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 145–164. Jeanjean, P., Liedtke, E., Clukey, E.C., Hampson, K. & Evans, T. 2005. An operator’s perspective onoffshore risk assessment and geotechnical design in geohazard-prone areas. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 115–143. Kvalstad, T.J. 2007. What is the current “best practice” in offshore geohazard investigations? A state-of-the-art review. Proc. Offshore Technology Conf., Paper 18545-MS. Kvalstad, T.J., Andresen, L., Forsberg, C.F., Berg, K. & Bryn, P. 2005. The Storegga Slide: Evaluation of triggering sources and slide mechanisms. Marine and Petroleum Geology, 22 (1-2): 245–256. Lacasse, S. & Berre. T. 1988. State-of-the-Art: Triaxial testing methods for soils. Advanced Triaxial Testing of Soil and Rock, ASTM STP 977, 264–289. Ladd, C.C. 1991. Stability evaluation during stage construction. J. of Geotechnical Eng., 117(4): 540–615. Ladd, C.C. & DeGroot, D.J. 2003. Recommended practice for soft ground site characterization: Arthur Casagrande Lecture. Proc. 12th Panamerican Conf. on Soil Mechanics and Geotechnical Eng., MIT, 1: 3–57. Landon, M.M., DeGroot, D.J., & Sheahan, T.C. 2007. NonDestructive sample quality assessment of a soft clay using shear wave velocity. J. of Geotechnical and Geoenvironmental Eng., 133(4): 424–432. Leroueil, S. & Hight, D.W. 2003. Behaviour and properties of natural soils and soft rocks. In Tan et al. (eds.) Proc., Characterisation and Engineering Properties of Natural Soils, Balkema, 1: 29–254. Leroueil, S. & Marques, M.E.S. 1996. Importance of strain rate and temperature effects in geotechnical engineering. Measuring and ModelingTime Dependent Soil Behaviour. ASCE Geotechnical Special Publication GSP 61, 1–60. Low, H.E., Lunne,T.,Andersen, K.H., Sjursen, M.A., Li, X. & Randolph, M.F. 2010. Estimation of intact and remoulded undrained shear strength from penetration tests in soft clays. Géotechnique. doi: 10.1680/geot.9.P.017. Low, H.E., Randolph, M.F., Rutherford, C., Bernie, B.B. & Brooks, J.M. 2008. Characterisation of near seabed surface sediment. Proc. OTC., Houston, USA, Paper 19149. Lunne. T. 2010. The CPT in offshore soil investigations – a historic perspective. Proc. 2nd Int. Symp. on Cone Penetration Testing, Los Angeles. 9–11 May 2010.

Borel, D., Pench, A., Dendani, H. & Colliat, J.L. 2005. Deepwater geotechnical site investigation practice in the Gulf of Guinea. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 921–927. Brooker, E.W. & Ireland. H.O. 1965. Earth pressures at rest related to stress history. Can. Geotechnical J. 2(1): 1–15. Bugge, T. 1983. Submarine Slides on the Norwegian Continental Margin, with special emphasis on the Storegga area. IKU, Publication 110, 152 p. Carter, J.P., Davies, P.J., & Krasnostein, P. 1999. The future of offshore site investigation – robotic drilling on the sea-bed. Australian Geomechancis. 34(3): 77–85. Collinson, J. D. & Thompson, D. B. 1982. Sedimentary Structures. George Allen & Unwin, London, U. K. DeGroot, D.J., DeJong, J.T., Yafrate, N.J., Landon, M.M. & Sheahan, T.C. 2007. Application of recent developments in terrestrial soft sediment characterization methods to offshore environments. Proc. OTC, Houston, Paper18737MS. DeGroot, D.J., Ladd, C.C. & Germaine, J.T. 1996. Undrained multidirectional direct simple shear behavior of a cohesive soil. J. of Geotechnical Eng. 122(2): 91–98. DeGroot, D.J., Landon, M.M. & Lunne, T. 2008. Synopsis of recommended practice for sampling and handling of soft clays to minimize sample disturbance. Proc. 3rd Int. Conf. on Site Characterization, Taipei, Taiwan. DeJong, J.T., Yafrate, N.J. & DeGroot, D.J. 2010b. Evaluation of undrained shear strength using full-flow penetrometers. J. of Geotechnical and Geoenvironmental Eng. In press. DeJong, J.T., Yafrate, N., DeGroot, D.J., Low, H.E., & Randolph, M. 2010a. Recommended practice for full-flow penetrometer testing and analysis. Geotechnical Testing J. 33(2): DOI: 10.1520/GTJ102468. Donohue, S. & Long, M. 2007. Rapid determination of sample quality using shear wave velocity and suction measurements. Proc. Conf. on Soil Invest. and Found. Behaviour, SUT, London, Sept. 2007, 63–72. Dugan, B. Sheahan, T.C., Thibault, J.M. & Evans, T.G., 2010. Offshore sediment overpressures: overview of mechanisms, measurement and modeling. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. In press. Ehlers, C. J., Chen, J. Roberts, H. H., & Lee, Y. C. 2005. The origin of near-seafloor “crust zones” in deepwater. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis. 927–933. Evans, T. 2010. A systematic approach to offshore engineering for multiple-project developments in geohazardous areas. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. ISFOG 2010. In press. Fay, J.B., Montargès, R., Le Tirant, P., & Brucy, F. 1985. Use of the PAM self-boring pressuremeter and the STACOR large sized fixed piston corer for deep seabed surveying. Proc. Conf. Offshore Site Invest., SUT, London, 187–199. Flemings, P.B., Long, H., Dugan, B., Germaine, J., John, C., Behrmann, J.H., Sawyer, D., & Expedition 308 Scientists, 2008. Pore pressure penetrometers document high overpressure near the seafloor where multiple submarine landslides have occurred on the continental slope, offshore Louisiana, Gulf of Mexico. Earth and Planetary Science Letters, 269(3-4): 309–32. Germaine, J.T. & Ladd C.C. 1988. State-of-the-Art: Triaxial testing of saturated cohesive soils. Advanced Triaxial Testing of Soil and Rock, ASTM STP 977, 421–459. ISSMGE 2005. Geotechnical and geophysical investigations for offshore and nearshore developments. Int. Society for Soil Mechanics & Geotechnical Engineering. Handbook.

56 © 2011 by Taylor & Francis Group, LLC

Lunne, T. & Andersen, K.H. 2007. Soft clay shear strength parameters for deepwater geotechnical design. Proc., 6th OSIG, SUT, London, UK, 151–176. Lunne. T., Andersen, K.H., Low, H.E., Randolph, M. & Sjursen, M. 2010. Guidelines for offshore in situ testing and interpretation in deepwater soft clays. Can. Geotechnical J., In press. Lunne, T., Berre, T., Andersen, K.H., Strandvik, S. & Sjursen, M. 2006. Effects of sample disturbance and consolidation procedures on measured shear strength of soft marine Norwegian clays. Can. Geotechnical J., 43: 726–750. Lunne, T., Berre, T. & Strandvik, S. 1997a. Sample disturbance effects in soft low plastic Norwegian clay. Proc. Recent Developments in Soil and Pavement Mechanics. Rio de Janeiro, Brazil, 81–102. Lunne, T., T. Berre, S. Strandvik, K.H. Andersen & T.I. Tjelta 2001. Deepwater sample disturbance due to stress relief. Proc. Int. Conf. on Geotechnical, Geological and Geophysical Properties of deepwater Sediments. College Station, Texas, 64–85. Lunne, T., Isa, O. & Tan, M. 1996. Shallow gas problem at Duyong B offshore Malaysia. Proc. 11th Offshore South East Asia Conference. Singapore. Lunne,T. & Long, M. 2006. Review of long seabed samplers and criteria for new sampler design. Marine Geology. 226: 145–165. Lunne, T., Robertson, P.K., & Powell, J.J.M. 1997b. Cone Penetration Testing in Geotechnical Practice. Spon Press, London. Lunne, T., Tjelta, T.I., Walta, A. & Barwise, A. 2008. Design and testing out of deepwater seabed sampler. Proc. Offshore Technology Conf. Houston, USA, Paper 19290. Mesri, G. (1975). Discussion of ‘new design procedure for stability of soft clays. J. of Geotech. Eng., 101(GT4): 409–412. Mosher, D.C., Christian, H., Cunningham, D., MacKillop, K., Furlong, A. & Jarrett. K. 2007. The Harpoon free fall cone penetrometer for rapid offshore geotechnical assessment. Proc. 6th OSIG, SUT, London, UK, 81–90. NORSOK Standard 2004. Marine soil investigations. G-001, Rev. 2, October 2004. Norwegian Geotechnical Institute 2002. Early soil investigations for fast track projects: Assessment of soil design parameters from index measurements in clays. Summary Rep./Manual, NGI Report 521553-3, 15 Jan. 2002. Perkins, S. & Sjursen, M. 2009. Effect of cold temperature of unfrozen Troll clay. Can. Geotech. J. 46(12): 1473–1481. Poirier, S.E. & DeGroot, D.J. 2010. Development of a portable probe for field and laboratory measurement of low to medium values of soil suction. Geotech. Testing J. 33:3.

© 2011 by Taylor & Francis Group, LLC

Randolph, M.F. 2004. Characterisation of soft sediments for offshore applications. Proc. 2nd Int. Conf. on Site Characterization, Porto, Portugal, 1 209–231. Randolph, M. F., M. J. Cassidy, S. Gourvenec & C. J. Erbrich 2005. Challenges of offshore geotechnical engineering, Invited State of the Art paper. 16th Int. Conf. on Soil Mechanics and Foundation Eng., Osaka, Japan, 123–176. Randolph, M.F., Low, H.E. & Zhou, H. 2007. In situ testing for design of pipeline and anchoring systems. Keynote lecture. Proc. of the 6th OSIG, SUT, London, UK, 251–262. Ridley, A.M. & Burland, J.B. 1993. A new instrument for the measurement of soil suction. Géotechnique 43(2): 321–324. Sandbeakken, G., Berre, T., & Lacasse, S. 1986. Oedometer testing at the Norwegian Geotechnical Institute. Consolidation of Soils: Testing & Evaluation, ASTM STP 892, 329–353. Sheahan, T.C. & DeGroot, D.J. 2009. Seabed Sediment Pore Pressure: Genesis, Measurement and Implications for Design/Analysis. Final Report Int. Workshop for US Natl. Sci. Fdn. Project, www.offshoregeohazards.org, 114 pp. Strout, J.M. & Tjelta, T.I. 2007. Excess pore pressure measurements and monitoring for offshore instability problems. Proc. OTC, Houston, TX. Paper18706. SUT-OSIG 2000. Guidance Notes on Geotechnical Investigations for Subsea Structures. OSIG-Rev. 2, Prep. by Subsea Working Group of Offshore. Soil Invest. Forum. SUT-OSIG 2004. Guidance Notes on Geotechnical Investigations for Marine Pipelines. OSIG-Rev. 3, Prep. by Pipeline Working Group of Offshore. Soil Invest. Forum. Terzaghi, K., Peck, R.B., & Mesri, G. 1996. Soil Mechanics in Engineering Practice. John Wiley and Sons, New York. Tjelta, T.I. & Strout, J. 2010. Piezometer installation in deepwater Norwegian Sea. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. ISFOG 2010. In press. van Paassen, L.A. & Gareau, L.F. 2004. Effect of pore fluid salinity on compressibility and shear strength development of clayey soils. In Hack et al. (eds). Eng. Geology for Infrastructure Planning in Europe. Springer-Verlag. 327–340. Vidal, V.M.V., Vidal, F.V., Hernandez, A.F., Meza, E. & Zambrano, L. 1994. Winter water mass distributions in the western gulf of mexico affected by a colliding anticyclonic ring, J. of Oceanography, 50: 559–588. Whittle, AJ, Sutabutr, T, Germaine, JT, & Varney, 2001. A Prediction and interpretation of pore pressure dissipation for a tapered piezoprobe. Géotechnique. 51(7): 601–617. Young, G.A. 1986. The strength deformation properties of Smith Bay arctic silts. SM Thesis, MIT, Cambridge, MA.

57

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Gulf of Guinea deepwater sediments: Geotechnical properties, design issues and installation experiences J.-L. Colliat & H. Dendani Total, Pau and Paris, France

A. Puech Fugro France, Nanterre, France

J.-F. Nauroy Institut Français du Pétrole, Rueil-Malmaison, France

ABSTRACT: The paper describes the geotechnical properties of deepwater sediments encountered on the continental slopes of the Gulf of Guinea, in water depth ranging from about 400 m to 2000 m. After more than 10 years of site investigations, a large database on the behaviour of these sediments is now available and is tentatively summarized, with illustrations from various Gulf of Guinea sites located between Nigeria and Angola. The main characteristics of the deepwater West Africa clays are addressed, comprising plasticity, carbonate and organic contents, particle size distribution, mineralogy, sensitivity and thixotropy, compressibility and shearing behaviour in relation to the clay microstructure. It is shown that specific laboratory and in-situ testing procedures are required for proper determination of fine particle sizes and for the measurement of the relatively high soil sensitivity. The presence and origin of a near seabed “crust” is highlighted. Its origin is questioned, and potential implications on the design of soil-pipeline interactions are emphasized. Typical results of installation of suction piles, driven piles and VLA anchors are presented, which further illustrates the clay behaviour.

1

INTRODUCTION

(CLub pour les Actions de Recherche sur les Ouvrages en Mer) and with the financial support of CITEPH (Programme de Concertation pour l’Innovation Technologique dans l’Exploration et la Production des Hydrocarbures). As a key regional characteristic, the Gulf of Guinea is a “geotechnically remote” area, meaning that no soil investigation drilling vessel is permanently present in West Africa. In 1998, the first deepwater Gulf of Guinea soil investigation, at the Girassol site in 1300 m of water offshore Angola, was done in conventional drilling mode. Following this, the soil investigation strategy adopted for the Gulf of Guinea was based on the use of 30–40 m penetration seabed-based in-situ testing tools and 15–25 m long gravity piston corers. In soft deepwater sediments, such seabed-based equipments are limited by their stroke range, not thrust capacity. The requirements for geotechnical engineering purposes for suction piles and anchors (main type of foundation in deep offshore West Africa clays) are then covered. The drilling of deep soil borings (needed for e.g. driven piles for TLPs) currently still represents a small minority of cases in the Gulf of Guinea, which is a major difference with the Gulf of Mexico practice. The physical and geotechnical properties of the Gulf of Guinea deepwater sediments are described in detail in this paper. The main characteristics of these soft clays are addressed, comprising index properties

In about ten years, the deep offshore Gulf of Guinea has become a mature oil province with rapid development of numerous oilfields, like in the Gulf of Mexico or offshore Brazil. Compared to the Gulf of Mexico, the Gulf of Guinea is known to present benign environmental conditions. If subsea currents are indeed generally low and maximum storm waves limited to less than 8–10 m, strong squalls may impose severe design loads for the moorings of floaters (i.e. for temporary moorings of MODUs, or for permanent moorings of FPUs or FPSOs). The Gulf of Guinea is also known to present relatively gentle seafloor conditions, but the full list of geohazards can be encountered off Angola and Nigeria, such as faults, pockmarks, shallow gas, gas hydrates, salt and mud diapirs, seabed slope instabilities, and sub-seabed mass transport deposits. These deepwater geohazards, which may represent serious challenges in terms of geotechnical design, have been studied, principally by Total and Ifremer, since the early 1990’s, with several scientific cruises in Nigeria, Gabon, Congo and Angola (e.g. Cochonat et al. 1996, Sultan et al. 2007 & 2010). Geotechnical characterisation of the Gulf of Guinea deepwater sediments has been the topic of various collaborative R&D projects, carried out under the umbrella of CLAROM

59 © 2011 by Taylor & Francis Group, LLC

(with the noticeable exceptions of Nigeria and ultra deepwater Angola, see Sultan et al. 2007, Hill et al. 2010, or Evans 2010), a new soil investigation strategy has emerged with the use of 30–40 m stroke seabedbased in-situ testing tools and large-diameter gravity piston corers allowing the recovery of 15–25 m long samples (Borel et al. 2002 & 2005). A first application for the Kuito FPSO anchors in 430 m of water offshore Cabinda is described in Alhayari et al. (2000). Adopting this survey strategy (also applicable to other remote deepwater provinces, such as the Australasia area) may allow a reduction of mobilisation times by performing the soil investigation from vessels of opportunity. Although still rarely used, vessels of opportunity may become important in fast track development projects. Deep soil borings are of course still required, either for the design of long driven piles (e.g. for the Kizomba TLP or the Benguela-Belize CT in Angola, Labbe & Perinet 2004, Will et al. 2006), for the installation of jetted well conductors (Evans et al. 2002), or in areas where more sophisticated testing and sampling might be needed in relation to specific geohazards such as in ultra deepwater offshore Angola (Hill et al. 2010, Evans 2010). But the drilling of 100 m+ deep soil borings represents a small minority of cases in the Gulf of Guinea, which is a key difference with the Gulf of Mexico practice. Currently, three geotechnical drilling vessels are capable of drilling in up to 2000 m of water, and two more can reach 3000 m of water. The PROD (Portable Remotely Operated Drill) is an available alternative, with a seabed-based system that combines the ability to take samples (piston sampling or rotary rock coring) and perform penetration testing (CPTs BPTs). The PROD1 and PROD2 systems are capable to operate in up to 2000 m and 3000 m of water, respectively (Kelleher & Hull 2008). By saving the time for tripping the drill-string, such a seabed-based system can be more efficient at deepwater sites than conventional drilling tools deployed from a vessel deck at the sea surface (Osborne et al. 2010). In 2009, the PROD1 was applied for the first time in the Gulf of Guinea for a soil investigation in water depths of 1100–1400 m offshore Angola.

(water content and plasticity, carbonate and organic contents), particle size distribution, mineralogy, sensitivity and thixotropy, compressibility and shearing behaviour in relation to the clay microstructure. Having possible implications for the design of pipelines, the near seabed “crust” of stiffer clay, locally encountered at several sites in Nigeria and Angola, has been the subject of several studies. Some key results are discussed, but the paper shows that the origin of this “crust” remains largely unexplained. Where relevant, specific laboratory and in-situ testing procedures are emphasised and illustrated by case examples taken from a number of West Africa deepwater sites located between Nigeria and Angola. The examples presented in the paper, covering water depths in the range of 400 to 2200 m, come from several sites operated by Total (containing both published and proprietary data) and from data published by the other three operators actively involved in the Gulf of Guinea, i.e. BP, Chevron and ExxonMobil. Some data presented in the paper were previously published but are completed by a larger database. Other unpublished data come from recent R&D studies. Some preliminary results from ongoing work on new topics of interest, such as strain rate effects, interface friction and modelling, are also presented. The design issues related to soil-pipeline interactions are briefly discussed. Finally, typical results of installation experiences are presented, covering suction piles, driven piles and VLA plate anchors. Such installation case studies provide a further illustration of the behaviour of Gulf of Guinea deepwater sediments, in particular with respect to friction degradation (i.e. clay sensitivity) and regain of strength with time (i.e. thixotropy and set-up). 2

SOIL INVESTIGATION STRATEGY

2.1 A strategy adapted to regional requirements The first deepwater soil investigation in the Gulf of Guinea was done in 1998 for the Girassol site at a water depth of 1300 m offshore Angola. It was carried out in conventional mode, i.e. from a specialised geotechnical drilling vessel and with down-hole sampling and in-situ CPT tools operated through the drill-string. Extending the shallow water soil investigation strategy to deep water (as was done in the Gulf of Mexico too) presented operational efficiency drawbacks, like the long time required for tripping the drill-string and the use of a 1.5 m stroke CPT tool, thus making the acquisition of data in deep borings rather time consuming. With FPSO and subsea wells being the preferred option for deepwater oilfield developments in West Africa, most of the foundations and anchors to be designed and installed only require a detailed investigation of the first 20–30 m of sediments, with emphasis on the top 1–2 m for the soil-pipelines or soil-risers interaction. Coupled with relatively gentle seafloor conditions in many Gulf of Guinea development areas

2.2

Large-diameter (100 mm) piston corers have been shown to provide quality of soil cores similar to 75 mm (3

) thin-walled pushed-in piston samplers with the conventional down-hole method from a specialised geotechnical drilling vessel (Young et al. 2000). However, gravity piston corers may by-pass some of the top 1.5–2 m of very soft seafloor sediments. This may also be the case in extremely soft seabed sediments for the STACOR® which uses a base plate and a truly fixed piston (Wong et al. 2008). In its current configuration, the STACOR® corer has a 25 m long core barrel and will generally recover 12–25 m long samples, depending on the water depth or the clay strength gradient (Borel et al. 2002, Enjaume et al. 2010). Longer gravity piston corers may be able to recover up to 30–35 m long

60 © 2011 by Taylor & Francis Group, LLC

Scope to 30–40 m for foundations or anchors

cores in specific conditions, i.e. with a heavier ballast weight and in extremely soft sediments or highly sensitive clays (Montarges et al. 1987), but the operation of such long corers is not a routine practice and can be detrimental to the sampling quality. Using accelerometers for accurate monitoring of the coring operation is a significant improvement for better understanding the corer behaviour and improving core recovery and quality (Buckley et al. 1994, Bourillet et al. 2007). In addition to the CPT, a number of in-situ testing tools can be deployed from seabed-based systems, the most commonly used being the field vane (VST), and the full-flow penetrometers (T-bar TPT and ball BPT; Randolph et al. 1998 & 2005). In soft sediments, the maximum investigation depth is not limited by the thrust capacity but by the maximum achievable stroke (generally 30–40 m). Based on years of practice, the undrained shear strength of the clay is determined by correlation with the cone resistance (Lunne & Andersen 2007). Other in-situ testing tools have the advantage of either a direct measurement (VSTs), or an increased accuracy in very soft sediments (TPTs and BPTs). Typical examples of Gulf of Guinea measurements are provided in subsequent sections of the paper. The use of various in-situ testing tools offers a superior data quality and a better definition of the soil strength intercept at the seafloor. Borel et al. (2010) propose a soil investigation strategy, where “clusters” of in-situ tests (CPT, VST and cyclic TPT or BPT) are performed at a number of selected locations, in combination with high quality sampling for advanced laboratory testing and calibration of the Nk factors. At other locations, the scope can be limited to penetrometer (CPT or full flow) testing and a reduced number of sampling for assessing the variability in soil conditions over the site. An accurate calibration of the Nk (alternatively Nt ) factor then allows a relatively precise determination of the clay strength, and the main geotechnical design parameters can be defined shortly after the soil investigation on the basis of the penetrometer test results.

Figure 1. Evidence of seabed crust from lightweight CPT testing and laboratory testing in box core (after Borel et al. 2005).

soil disturbance and frame penetration in the soil. An example of results obtained with the SEASCOUT® is shown in Figure 1. Piston gravity corers have been extensively used for pipeline soil investigations in deep waters. The analysis of the top decimetres of samples however often discloses partial loss or remoulding of the material which cannot be used for accurate measurements at very low stresses. Concerns have also arisen with giant piston corers regarding their capacity to capture subtle changes in the geotechnical parameters in the top decimetres below seabed, due to possible washing-out of the extremely soft seabed materials during the initial phase of free-fall penetration. Until recently, box corer sampling was the best option for recovering intact seabed samples. The standard box corer, typically weighing about 200–300 kg, penetrates 40–50 cm into the seabed sediments under its own weight. A trigger mechanism then releases a latch which causes the swivel base to close-off the captured sample, before the whole unit is recovered on board the vessel. The recovered samples are of high quality, but the extremely soft nature of seabed soils cannot allow any sub-sampling without disturbing the material. “In situ” testing inside the box therefore appears as the only reasonable way to obtain relevant shear strength data. This can be done with a minivane, but using a mini T-bar has the advantages of (a) obtaining continuous profiles and (b) measuring the soil sensitivity by performing quick cyclic tests. A manually operated mini T-bar (called DMS) was developed at the Centre for Offshore Foundation Systems (COFS, Perth, Australia) for strength profiling in box cores (Randolph et al. 2007). In 2008, Fugro developed the DECKSCOUT™ system (Fig. 2), in which the tool can be fitted with full flow penetrometers (T-bar or ball probes) with an actuator which can apply constant rates of penetration in the range of 0.01 to 2 cm/s (Puech et al. 2010); cyclic testing for sensitivity measurement can be automatically monitored.

2.3 Scope to 1–2 m depth for pipelines or risers Soils encountered in deep waters generally present shear strength profiles linearly increasing with depth (strength gradient between 1.0 and 1.5 kPa/m), starting at very low values at seabed (typically 1 to 2 kPa). In these very soft soils, assessing the longitudinal and lateral restraints of pipelines or flowlines laid on the sea bottom is a challenging issue for the industry. This requires both a very accurate measurement of the intact and remoulded undrained shear strengths over the first decimetres (accuracy in the order of or better than 1 kPa) and a precise soil-pipe interaction model. In the Gulf of Guinea the random presence of the so-called seabed “crust” (see details in section 4.9) increases difficulties in assessing accurate shallow shear strength profiles and impact on pipe penetration. In-situ tests (CPTU and T-bar tests) have been performed from lightweight frames in order to reduce

61 © 2011 by Taylor & Francis Group, LLC

Figure 3. SMARTPIPE® and SMARTSURF™ modules. Table 1. Country

Site

Water depth (m)

Nigeria

A K O Q∗ U M Z

350–650 1250–1500 1500–2000 1400–1500 750–800 550–1100 1300–1400 400 1300–1400 700–1100 650–700 800–1400 1400–2250 1200–1500 1350 1300–1400

Figure 2. Fugro DECKSCOUT™ system in operation on box corer.

It is now widely accepted that soil-pipe interaction laws in very soft soils should better be obtained directly on the sea bottom. The Fugro SMARTPIPE® system measures directly in-situ the penetration of an instrumented pipe section and the longitudinal and lateral restraints resulting from static or cyclic loads (Hill & Wintgens 2009, White et al. 2010). The system is equipped with a 1 m stroke mini T-bar to characterise the seabed sediments at the test location. The second version of the equipment was successfully used offshore Angola at the end of 2009. As the SMARTPIPE® will necessarily be used at a limited number of discrete locations to obtain detailed information on the soil-pipe behaviour, a companion and complementary tool called SMARTSURF™ has been designed (Puech et al. 2010). It is aimed at providing accurate information on seabed soil properties required at a large number of locations regularly distributed all along the pipeline routes to (a) assess whether the chosen test locations give representative results and (b) extrapolate data with confidence to the entire pipeline network. The SMARTSURF™ is equipped with a 3 m stroke standard T-bar or CPTU, a 1 m stroke mini-T-bar and a pushed-in stationary piston sampler specifically designed to recover 2 m long cores of soft undisturbed soil. The SMARTPIPE® and SMARTSURF™ can be operated with the same launch and recovery system (LARS), allowing safe handling from any vessel of opportunity fitted with a convenient A-frame (Fig. 3). The SMARTPIPE® and the SMARTSURF™ can also be deployed from a specialist drilling vessel. The equipment is rated for up to 3500 m water depth. A site investigation strategy to obtain fast and reliable design parameters for soil-pipe interaction assessment entirely based on in-situ testing is proposed by Borel et al. (2010).

Congo Angola

∗

3

B∗∗ D F G I L P R S

Ehlers et al. (2005) ∗∗ Dutt & Ehlers (2009)

PHYSICAL PROPERTIES

Physical and geotechnical properties described in sections 3 and 4 were derived from a review of industrial or R&D data from sixteen sites. Details are given in Table 1. 3.1 Water content and submerged unit weight Deepwater sediments in the Gulf of Guinea are characterised by very high water contents (Fig. 4). At the seafloor, w is typically comprised between 150 and 250%, and decreases to 100–200% over the very first metres of penetration (typically 6–8 m). Over these upper 6–8 m the water contents are close to or over the liquid limit (giving 1 < LI < 1.2). Beyond that depth the rate of decrease becomes very slow over the depth of interest for the engineering of most deepwater structures (i.e. 20–30 m below seabed). Most of the sites exhibit water contents between 80 and 150%. The highest values are mainly (but not exclusively) representative of the deepest sites. A few deep penetration boreholes are available. They confirm a decreasing trend (Fig. 5), but water

62 © 2011 by Taylor & Francis Group, LLC

Gulf of Guinea sites considered in paper.

Figure 4. Water contents versus penetration in top 20 m (from STACOR® samples).

Figure 5. Water contents versus penetration depth (deep boreholes).

contents still present relatively high values (80–120% at about 120 m of penetration in the deepest boreholes). The site B borehole appears as an exception but it is worth noting that site B is in only 400 m water depth (Dutt & Ehlers 2009). There is a relatively large scatter in the values of water contents at a particular depth. This observation is not only valid for the global data set but also for a particular site or field, and even for a given borehole as highlighted in Figure 4. The bulk unit weights are very low, starting at 12–13 kN/m3 at seabed and reaching 13–15 kN/m3 below 6–8 m. Submerged unit weights of Gulf of Guinea soils are significantly lower than in the Gulf of Mexico where submerged unit weights become rapidly higher than 6 kN/m3 (e.g. Quiros & Little 2003). 3.2 Carbonate content Carbonate contents are generally lower than 20% of the total soil weight (Fig. 6). Variations with depth are mostly erratic even along the same profile (see for example site M). Scanning electron microscopy (SEM) observations indicate that carbonates are mainly shell debris or foraminifers randomly distributed in the clay matrix. The proportion of calcite detected in the finer fractions (<80 µm) is generally small. Infrared spectrometry measurements were performed on a sample of site F which presents a carbonate content of about 20% (Fig. 7). The absorbance peak corresponding to carbonates at a wave number of 1430 cm−1 is strongly reduced after sieving at 80 µm.

Figure 6. Carbonate contents versus penetration depth.

Calcite cannot be considered as a cementing agent of the soil matrix. Therefore, the compressibility and shear strength of Gulf of Guinea clays are not driven by the carbonate content, as is explained in sections 4.2 and 4.3. 3.3

Particular care should be taken in the measurement of the organic content of Gulf of Guinea sediments. Past experience has shown that the organic content of

63 © 2011 by Taylor & Francis Group, LLC

Organic content

Figure 7. Infrared spectrometry measurements for site F.

these soils is dramatically overestimated when using the traditional loss on ignition (LOI) method, which provides organic matter content in the range 10–20%. This overestimation is mainly due to the high fraction of kaolinite in the Gulf of Guinea clays. Kaolinite decomposes into metakaolin at 450˚ C, thereby loosing up to 13% of its weight in the form of water, which is interpreted as a loss of organic matter. A more appropriate method consists in deriving the organic matter content from the total organic carbon (TOC). The TOC represents the percentage in weight of atomic carbon involved in organic bonds, and can be quantified as the sum of the pyrolysable carbon (PC) and the residual carbon (RC) as measured by e.g. Rockeval or Poluteval techniques (Thomas et al. 2007). Several samples tested at IFP provided TOC percentages of 1–2%, in good agreement with values given by Holtvoeth (2004). The total organic carbon is not easily related to the bulk organic matter content since other atoms than carbon are also included in the organic matter: oxygen, hydrogen, sulphur, nitrogen. The organic content (percentage in weight of the organic matter) can be derived by multiplying theTOC by a conversion factor aimed at taking into account the percentage of carbon in the bulk organic matter. For this purpose, Broadbent (1953) recommended the use of a factor ranging from 1.9 to 2.5. For the Gulf of Guinea deepwater sediments, the conversion factor was assessed at IFP with the help of the Kerogenatron method, designed for pure kerogen isolation with no loss nor alteration. Analysis of specimens from various sites yielded organic matter contents in 100◦ C dried samples ranging from 3 to 6%. Although this method is rigorous, it is not suitable for routine applications due to its cost and complexity. In the light of the Kerogenatron results, a conversion factor of 2.5 appears to be reasonable, suggesting that the organic contents of Gulf of Guinea deepwater soils generally do not exceed 6–8% (Fig. 8). The dichromate oxidation method, as recommended by e.g. the British Standard BS 1377, yields organic content values typically between 3 and 10%, with the exception of site P.Although no direct comparison with

Figure 8. Organic contents versus penetration depth.

the TOC method is available, the dichromate oxidation method then appears as a reasonable alternative. 3.4

Particle size distributions (PSD) of Gulf of Guinea sediments are below 400 µm and very often below 100 µm. It has been observed that PSDs obtained from standard hydrometer testing after dispersion by sodium hexametaphosphate (HMP) are poorly representative of the actual particle sizes. SEM micrographs show an open flocculated structure consisting of series of linked clay aggregates (see later Figs. 18–19). The size of the unit particles appears to be small (a few micrometres) whereas crushing of the dry sample results in aggregates bigger than 5–10 µm (Fig. 9). These aggregates are difficult to de-flocculate by polyphosphate dispersive agents. This results in a general shift of the PSD towards coarser particles with stepping distribution curves where classes of particle sizes are missing. Several dispersing agents were systematically tested at the INPL/LEM laboratory (Nancy, France) in a Malvern Mastersize laser granulometer. Cation exchangers in the form of polysulfonate resin beads (100–500 µm in diameter) have been found to be the most efficient (Fig. 10). This technique offers the advantage to release Na+ and to adsorb the multivalent cations, which are removed from the dispersed suspension instead of remaining as soluble complexes. Ultrasounds applied to an untreated sample produce also strong dispersion. Combining ultra sounds and resin is the ideal solution when the laser granulometer is used.

64 © 2011 by Taylor & Francis Group, LLC

Particle size distribution

Figure 9. SEM photograph of a typical Gulf of Guinea clay. Figure 12. X-ray powder diffraction patterns of a typical Gulf of Guinea clay. M/I mica or illite, K kaolinite, Ph phyllosilicates, Q quartz, H halite, C carbonate, Py pyrite (from Thomas et al. 2007).

It is recognised that (a) hydrometer sedimentation overestimates the content of the finest particles, especially when these are dominant (which is the case in the present study), and (b) laser diffraction overestimates the largest particles because the intensity of the diffracted light varies with the second power of the volume, i.e. with the sixth power of the size, which means that a tenfold larger particle scatters one million stronger light. For Gulf of Guinea sediments the main effect originates from the platy shape of the analyzed particles (Fig. 9). A correct agreement between the sedimentation curve and the laser diffraction curve can be achieved using a modified Stokes law with simple disc model. Comparing the results obtained by sedimentation and by laser diffraction further yields additional information about the average aspect ratio of the particles.

Figure 10. PSD curves obtained with various dispersive agents (after Thomas et al. 2005).

3.5

Figure 11. Densitometer versus laser granulometer PSDs (after Thomas et al. 2007).

As was discussed by Thomas et al. (2007), differences in PSD are associated with different measurement techniques. Interpreting PSD curves must be done in relation to the method used because standard hydrometer (densitometer) and laser granulometer yield dramatic different results. As shown in Figure 11, the clay content, defined as the amount of particles passing the 2 µm sieve, is 20–30% with the granulometer and 60–80% with the hydrometer.

65 © 2011 by Taylor & Francis Group, LLC

Mineralogy

Figure 12 exemplifies typical X-ray diffraction powder patterns which can be routinely obtained on sediment specimens of the Gulf of Guinea. The detected crystalline phases are quartz, calcite, halite and phyllosilicates. Among the latter minerals, kaolinite appears as the only one significantly detected (2θ = 14.5◦ /7Á). No clear modulation around 7◦ can suggest the possible presence of smectites. The different clay types are better identified on oriented preparations (Fig. 13). The 2:1 clays (smectite or micas) are clearly identified around 7–8◦ / 14–16Á, illite at 10.5◦ /12Á and its first harmonic at 21◦ /6Á; kaolinite at 14.5◦ /7Á and its first harmonic at 29◦ /3.5Á. Adsorption of ethylen glycol reveals the presence of smectite (swelling clay) by the shift of the 7–8◦ /14Á peak towards 6◦ /18–20Á. Calcination at 550◦ C eliminates kaolinites and smectites and reveals the part of illite at 10–11◦ /10.5Á. The specific surface areas measured by absorption of methylene blue from aqueous solutions are between 180 and 260 m2 /g, which confirms the significant presence of smectite (Le, 2008).

Figure 13. X-ray diffraction patterns on oriented preparation sample after separation by sedimentation. LN: oriented preparation, EG: ethylene glycol treatment, 550◦ C: calcinated at 550◦ C.

Whereas properly applied X-ray diffraction allows a clear qualitative identification of the different clay species, their quantification remains a challenging task. Using the X-ray diffraction pattern is not relevant, because the height and area of a diffraction peak are determined not only by the abundance of species but also by the structure factor. Sophisticated and arduous techniques have shown the extreme difficulty in investigating the morphological and textural complexity of the clay phases, and in distinguishing between mixed illite-smectite phases, mixed kaoliniteillite phases and aggregated species. Transmission electron microscopy (TEM) showed that kaolinite particles occur in small size and strong anisotropy. The poor crystalline quality of kaolinite was confirmed by infrared spectroscopy (Thomas et al. 2007). As a semiquantitative assessment, one can assume that kaolinite is the dominant species (probably about 40–50% of the clay minerals) and that smectites are in significant proportion (>20% and up to 40%).

Figure 14. Atterberg limits versus penetration depth.

3.6 Plasticity The plasticity index (PI) of Gulf of Guinea deepwater clays is typically between 70 and 130 but can reach values as high as 150 near the seabed. These values are much higher than those frequently encountered in the “highly plastic” Gulf of Mexico clays with PI typically in the range 30–70. The plastic limit (WP) is quasi constant with depth, in the range of 50 ± 10% whereas the liquid limit (WL) typically decreases from 150–200% at the seafloor to 130–170% below 6–10 m of penetration (Fig. 14). When represented in the Casagrande diagram, the soils plot close to the A-line and classify as highly plastic clays (CH) to highly plastic silts (MH), as shown in Figure 15. 4

Figure 15. Gulf of Guinea clays in Casagrande diagram.

to define the stress history of the Gulf of Guinea soils, since no implicit assumption regarding the past overburden pressure is introduced. The YSR is determined
from the ratio of the vertical yield stress σvy to the
actual overburden pressure σv0 :

The vertical yield pressures derived from a number of standard oedometer tests performed on several typical deepwater sites are plotted versus depth in Figure 16 and are used to compute the yield stress ratio.
is higher than the vertical effective The yield stress σvy
stress σv0 . The distance between the profiles expressed

− σv0 ) is quasi constant by the difference (σ = σvy with depth, typically in the range of 15–40 kPa for most sediments. The corresponding YSR values decrease

GEOTECHNICAL PROPERTIES

4.1 Yield stress ratio The term yield stress ratio (YSR) is preferred to the commonly used term over-consolidation ratio (OCR)

66 © 2011 by Taylor & Francis Group, LLC

Figure 17. Sensitivity framework for site Z clay. Figure 16. Vertical yield pressure and YSR versus penetration depth for typical Gulf of Guinea sites.

The properties of a natural clay differ basically from its intrinsic properties due to the influence of the soil structure (fabric and/or bonding). The structure of a natural clay depends on many factors, such as depositional conditions, ageing, cementation and leaching. The evolution of the void ratio with penetration depth,
or with the effective vertical pressure σvo in a natural clay deposit, is called the sedimentation compression curve (SCC). The majority of normally consolidated natural clays have sedimentation compression curves which (after normalisation) lie in a narrow band above the ICL. The regression line through this band has been called the sedimentation compression line (SCL) by Burland. The SCL is not a fundamental line but represents the trend followed by a majority of natural sedimentary clays. Temporal variations may result in “saw-toothed” curves instead of smooth regular curves. The fact that the SCL lies to the right of the ICL implies that, for a given void ratio, a natural clay is capable of carrying an effective overburden pressure higher than the corresponding reconstituted clay. Cotecchia & Chandler (2000) have generalised Burland’s approach to sensitive clays. They have shown that the SCL is not unique (as implicitly assumed in Burland’s work) but depends on the soil sensitivity. Typically, for site Z clay, the sensitivity framework approach is illustrated in Figure 17 by plotting the following data:

from about 3 at 2–3 m of penetration to 1.1–1.3 at depth. Higher values of YSR can be found within the first 2 m below seabed, depending on the presence of a so-called “crust” (see sections 2.3 and 4.9), but this particular point is not discussed here. Yield stress ratios in excess of 1 are not the result of over-consolidation in the geological sense (i.e. no past overloading of the material). De Gennaro et al. (2005) have shown that these soils exhibit a significant

structural effect, and that the (σvy − σv0 ) difference is a quantitative measurement of the “extra-strength” due to the material structure. 4.2 Compressibility De Gennaro et al. (2005) and Le et al. (2008) have highlighted the interest of interpreting the compressibility data of Gulf of Guinea deepwater sediments in the light of the conceptual framework introduced by Burland (1990) for comparing and interpreting the compressibility of natural sedimentary clays and reconstituted clays. A normalising parameter called void index Iv is introduced to aid in correlating the compression characteristics of various clays:

– the ICL, as proposed by Burland (1990); – the ICL obtained for Gulf of Guinea deepwater sediments (Le 2008). For effective vertical stresses in excess of 10 kPa (i.e. soils below 3 m penetration), a very good fit is observed; – the SCL proposed by Burland, which corresponds to a sensitivity St of about 4 to 5; – the SCL proposed by Cotecchia & Chandler (2000) for a sensitivity of 6; – the envelope and mean value of the SCCs obtained from the water content measurements on the site Z samples.

where e∗100 and e∗1000 = intrinsic void ratios corresponding to an effective vertical stress σv
of 100 kPa and 1000 kPa respectively (determined by oedometer tests on reconstituted sample); and C∗c = intrinsic compression index. The properties of reconstituted clays are called “intrinsic” properties since they are inherent to the soil and independent of its natural state. The compressibility of reconstituted clays can be described by a reasonably unique line called the intrinsic compression line (ICL) passing by the points (Iv = 0, σv
= 100 kPa and Iv = −1, σv
= 1000 kPa). For a particular clay, e∗100 and C∗c are related to the value of the void ratio at the liquid limit eL . © 2011 by Taylor & Francis Group, LLC

Based on the trend of the sedimentation compression curves, this analysis suggests that the clay has a marked

67

structure and a sensitivity in the order of 4 to 6. It is worth mentioning that the sensitivity measured from in situ cyclic T-bar testing was found close to 6. According to the sensitivity framework, a normally consolidated soil with a post-sedimentation structure is characterised by compressibility curves that cross
the SCL at their in-situ stress (σv
= σvo ) before show
ing an abrupt increase in compressibility at σv
= σvy (yield stress). This is exactly what is observed in Fig
ure 17 for site Z. At stresses above the yield stress σvy , the compression curves plunge and reach the experimental ICL at higher stresses (about 200–300 kPa). The compressibility is maximal after the yield point and values of compression index Cc as high as 2 can be observed. In summary, the Gulf of Guinea deepwater clays (a) are normally consolidated, (b) have a marked post sedimentation structure, and (c) have a sensitivity in the range of 3 to 6, and this sensitivity can be interpreted as a quantitative measurement of the structural effect. 4.3

Figure 18. SEM micrographs of site D clay samples from 0.5 m and 14.0 m penetration.

Microstructure

The microstructure of intact samples has been investigated at ENPC/CERMES (Le 2008) by using scanning electron microscopy (SEM) and mercury intrusion porosimetry (MIP) in parallel, as suggested by Delage & Lefebvre (1984).To preserve the microstructure during dehydration, samples were frozen as quickly as possible so as to get a crypto-crystalline structure of ice with no volumetric expansion due to freezing according to the methodology proposed by Delage et al. (2006). Figure 18 shows two SEM pictures of surface (0.5 m) and deeper (14 m) samples. At both depths, a clear evidence of large voids can be observed, with diameters equal or larger than 1 µm, typical of soft soils. Voids appear to be as frequent as the solid phase, corresponding to an approximate porosity of 50%. Connecting clayey bridges are well observed, giving to the assemblage the aspect of a honeycomb microstructure. Clay minerals are easily observable with an average thickness smaller than 0.1 µm and an average diameter close to 0.5 µm. In Figure 19, the SEM photos of the 0.5 m sample submitted to vertical stresses of 50, 200, 800 and 1600 kPa emphasise the changes in microstructure during compression.The progressive collapse of the material structure, principally due to a decrease of the pore sizes, is responsible for the high compressibility of the clay. Mercury intrusion porosity (MIP) is based on the principle that a non-wetting fluid (here mercury) cannot enter a porous medium unless a pressure is applied. Assimilating pores to cylindrical capillaries, the pressure p can be related to an equivalent entrance pore radius r by Laplace’s law:

Figure 19. SEM micrographs of site D clay samples (0.5 m penetration) after compression to 50–1600 kPa (from Le 2008).

As the pressure p is increased, pores of smaller and smaller radius are filled with the intruding liquid. Results are plotted under the form of a cumulative curve giving the pore size distribution (eHg ) of the porous medium. In Figure 20, the MIP pore size distribution curves of intact samples from various penetration depths define a particular pore entrance value of about 0.2 µm. This radius is compatible with the intra-aggregate pore sizes observed through SEM. Below 0.2 µm, the curves are superimposed and this domain is known to be typical of the inter-aggregate porosity and to be poorly sensitive to macroscopic changes in void ratio. 4.4

4.4.1 CPT cone resistance and T-bar profiles Typical cone and T-bar resistance profiles are presented in Figure 21. They are representative of the large number of CPTs performed on West Africa continental slopes. Below 2 m of penetration and down to 30–40 m, the penetration resistance increases quasilinearly with depth. The gradient in net cone resistance lies between 10 and 30 kPa/m with a general tendency

where σ = surface tension of the intruding liquid; and θ = contact angle.

68 © 2011 by Taylor & Francis Group, LLC

Undrained shear strength

Figure 22. Gradient of net cone resistance versus water depth.

Figure 20. MIP pore size distribution curves for site D clay (from Le 2008).

Figure 23. T-bar and VST test results with Ntv = 11.5 (from Borel et al. 2005).

4.4.2 Undrained shear strength and sensitivity Vane shear tests (VST) provide a measurement of the in-situ peak shear strength Suv at typically 1 m or 1.5 m intervals. The net cone resistance and the vane shear strength can be related by a cone factor Nkv such as qn = Nkv . Suv . The Nkv factor is in the range of 10–15. More detailed analyses performed on a small number of sites where 30 to 40 m profiles of VST and CPT tests could be closely correlated suggest an Nkv factor increasing with depth from about 11 near the seabed to 14–15 at depth (Puech et al. 2005). The Ntv factor applicable for T-bar data has been found close to 11.5 either for in situ standard T-bar testing (Fig. 23) or for mini T-bar testing (Puech et al. 2010). For water depth in excess of 700–800 m, the average gradient in shear strength ranges from 1.0 to 1.5 kPa/m. Slightly higher values can be locally encountered at shallower sites. Ultra deep water sites investigated so far (in a water depth of about 2000 m) have gradients on the lower bound. However, gradients of the same order have been found at sites in only 1400 m of water
are depth. Normalised shear strength ratios SuVST /σvc high, with values in excess of 1 over the very first

Figure 21. CPT net cone resistance and T-bar profiles (site K).

to decrease with increasing water depth as shown in Figure 22. In areas unaffected by the presence of geohazards, resistance profiles are highly repetitive (additional examples of CPT profiles are given in Colliat & Colliard 2010). Over the top 2 m of penetration two different types of profiles may be encountered i.e. profiles linearly increasing with depth or profiles presenting a peak in resistance corresponding to the seabed crust. Comparisons are presented later in Figure 33. The origin and consequences of such peaks in geotechnical engineering are discussed in sections 4.9 and 5.

69 © 2011 by Taylor & Francis Group, LLC

Figure 24. Composite plot showing intact shear strength profiles (CPT and VST) and residual shear strength profiles (VST) compared to CPT sleeve friction and cyclic T-bar tests (cycles 10–30 shown). Figure 25. Design remoulded shear strength profiles (assuming St = 3 to 4) compared to CPT sleeve friction, VST tests (remoulded strength after 3 rotations) and cyclic BPT test results (cycles 1–10 shown).

1–2 m of penetration, decreasing to about 0.4–0.5 below 10 m. The latter is compatible with a ratio
= 0.36 mentioned by Andersen & Jostad SuDSS /σvc (2004), assuming SuDSS = 0.75 SuVST (Fig. 27). The sensitivity St is expressed as the ratio of the intact undrained shear strength Su,i to the remoulded undrained shear strength Su,r, measured using the same instrument. The measurement of clay sensitivity is a topic that attracted specific attention, in particular since installation experiences suggested that it tended to be under-estimated by the measurements provided by both laboratory tests (UU) and in situ tests (see section 6.1). Farrar et al. (2008, cited in Robertson 2009) claim that the CPT sleeve friction is often close to the remoulded undrained shear strength measured by the VST in soft NC clays. But Lunne & Andersen (2007) have shown that the CPT sleeve friction measurements do not agree with other remoulded shear strength measurements. Figures 24 and 25 present a comparison of data obtained with the VST, the CPT and full flow penetrometers (Tbar and ball) at two different sites in Angola. In Figure 24 are shown:

Figure 25 present results from another site, showing: – the CPT sleeve friction; – the remoulded shear strength as obtained with the VST but after only 3 rotations of the blades instead of 10 as per ASTM recommendations (limitations due to time constraints); – the remoulded shear strength derived from the ball tests (Nb = 12) after 10 cycles; – the design remoulded shear strength profiles based on labvane tests and past installation experiences in the area (assuming a clay sensitivity of 3 to 4). These data illustrate typical results obtained in the normally consolidated but structured clays of the Gulf of Guinea, i.e.: • The CPT sleeve friction clearly compares with the

residual (not remoulded) VST shear strength values; • The VST remoulded shear strength measurements

(when the number of blade rotations is limited to 2 or 3 as is often the case in offshore practice) plot below the residual shear strengths but provide underestimated values of the clay sensitivity (i.e. in the range of 2 to 3); • The cyclic T-bar or ball tests provide higher values of sensitivity, generally consistent and less scattered

– the cone resistance (divided by a cone factor of 13) and the peak VST shear strength; – the CPT sleeve friction and the residual shear strength obtained with the VST; – the remoulded shear strength as derived from cyclic T-bar with Nt = 11.5 (cycles 10 to 30 are shown). © 2011 by Taylor & Francis Group, LLC

70

Figure 27. Correlation between VST and DSS shear strengths.

value representative of normally consolidated conditions (K0nc ) measured in the post-yield domain is close to 0.5 (in the range of 0.45–0.55 for the small number of specimens tested). Higher values are obtained in the pre-yield domain. 4.5.3 Shear strength anisotropy Series of direct simple shear (DSS) tests and anisotropically consolidated (CAUc and CAUe ) triaxial tests were performed. The shear strengths obtained are respectively noted SuDSS , Suc and Sue . The ratio Suc /SuDSS has been found close to 1.2. The values obtained for the Sue /SuDSS ratio are more scattered, due to the difficulty in controlling extension tests on very soft soils and at very low confining pressures, but a value of about 0.8 seems appropriate on the basis of the most reliable sets of data. A fair correlation has been found between the shear strengths obtained from VST and DSS tests, of the type SuDSS # 0.75 SuVST (Fig. 27). This result may be interesting for fast track engineering when only insitu data are available. It is noted that the data imply relatively high conversion factors between penetrometer resistances and average laboratory strengths (e.g. NtDSS # 15). This may be explained by both the high rate sensitivity of Gulf of Guinea clays and the robustness of their fabric (reduced softening with first pass of penetrometer).

Figure 26. Typical results from cyclic T-bar tests in site Z clay.

than labvane measurements. Sensitivities are often in the range of 3 to 4 but can reach 6 for site Z as shown in Figure 26. These values are in good agreement with the interpretation of compressibility data in the sensitivity framework (section 4.2) and were confirmed by back-analysing suction pile installation data at the respective sites (see section 6.1). Recent (unpublished) data obtained with the mini T-bar indicate that sensitivity values in excess of 10 may be obtained within the “crust”. It is generally admitted (e.g. Randolph et al. 2005) that 10 cycles are sufficient to reach stabilisation and derive a sensitivity value from full flow penetrometers. In the structured clays of the Gulf of Guinea, Borel et al. (2010) and Puech et al. (2010) emphasise the need to perform at least 25–30 cycles to reach stabilisation (Fig. 26).

4.6 Thixotropy 4.5 Advanced laboratory testing

Thixotropy is measured as the regain in undrained shear strength with time of remoulded samples maintained at constant water content and sheared at increasing time intervals. All available results (21 tests from 8 different sites) were compiled to provide the envelope curves presented in Figure 28. These results are consistent with those presented by Andersen & Jostad (2002). They are also in agreement with the set-up behaviour of suction piles at various Gulf of Guinea sites described by Colliat & Colliard (2010). The results obtained for site Z may suggest that the higher the sensitivity the faster the undrained

4.5.1 Sample quality A large majority of samples recovered in the deep waters of the Gulf of Guinea were taken using the STACOR® giant piston corer. All results referred to in this paper were derived from high quality STACOR® samples with the Lunne et al. (1998) quality index generally rating fair to excellent. 4.5.2 Coefficient of lateral earth pressure at rest K0 The coefficient of lateral pressure K0 can be obtained in drained triaxial tests under zero radial strain. The

71 © 2011 by Taylor & Francis Group, LLC

Figure 28. Results from thixotropy tests.

shear strength is regained, but data are too scarce to draw definite conclusions. 4.7

Consolidation properties

The coefficient of vertical consolidation Cv of Gulf of Guinea clays measured in incremental oedometer tests is generally found between 1 and 10 m2 /year in the virgin compression zone. The coefficient of consolidation Cv,r corresponding to recompression in the stress

range of 0.5 σvo to σvo (often used to evaluate reconsolidation around suction emplaced piles) lies in the same order of magnitude. These values apply in the range 2 to 20 m with a slightly decreasing trend with depth. An interesting point is that similar results are obtained on oedometer tests carried out on horizontally or vertically trimmed specimen, which is consistent with SEM observations showing identical microstructure in both directions. Unfortunately the authors are not aware of experimental field data (settlement of structures, pore pressure dissipation around instrumented pile walls) which could be used to confirm the validity of these values at large (foundation) scale. Thorel et al. (2010) have performed centrifuge tests on reconstituted Gulf of Guinea clays. The Cv values calculated from the observed settlements by the Asaoka method are of about 30 m2 /year, i.e. 5 to 6 times higher than Cv values measured from oedometer tests on in situ samples of the same clay. Of particular interest is the value of Cv at mudline for pipeline applications. Laboratory testing on the extremely soft surface material is highly challenging and poorly reliable. Hill & Wintgens (2009) and White et al. (2010) report consolidation measurements performed with the SMARTPIPE® system (Fig. 3) where an instrumented pipe section was allowed to settle on the seabed. The pore pressures dissipate rapidly showing that lay-induced pore pressures under a real pipeline would dissipate within days, not weeks. A satisfactory match between field observations and Finite Element Analyses is obtained for Cv values of about 75–100 m2 /year.

Figure 29. Results of triaxial shearing tests on site M clay (from Le et al. 2008).

4.8

72 © 2011 by Taylor & Francis Group, LLC

Effective parameters from triaxial testing

The shearing behaviour is largely explained by the structure of the clays. Le (2008) and Le et al. (2008) performed two types of tests on a natural clay: (a) triaxial compression and extension tests, in which the samples were consolidated to their estimated in-situ effective stress state, then sheared in undrained condition, and (b) SHANSEP type tests, in which the samples were consolidated isotropically (or anisotropically) beyond their in-situ states of stress before being sheared in undrained condition. Typical results are presented in Figure 29. The TN4 (compression) and TN7 (extension) tests performed at in-situ stress state present a non-plastic response in the q − p
space (p
nearly constant before yield). Failure is characterised by high effective friction angles of 40◦ or even more (M > 1.6), and such results are typical of all sites investigated so far (Fig. 30). The non-plastic pre-yield response is related to the strong structure of the clay. The high effective friction angles may be explained by a “sand-like” behaviour of the material where the aggregates of fine particles play the role of grain-sized elements. Friction angles in excess of 40◦ have been found in other clays, e.g. Mexico City clays (Diaz-Rodriguez et al. 1992).

Figure 31. Interface shear test results: peak shear strengths from ring shear tests on remoulded clay compared to Cam-shear tests on undisturbed clay.

Figure 30. Composite plot of triaxial test results in the q − p
space.

smooth interfaces simulating different pipeline coatings (Kuo et al. 2010). To simulate interface conditions at shallow pipeline embedment, the stress level was very low (<10 kPa), and rates of shearing ranging from 0.3 to 30 mm/min were used (where the slowest rate was expected to allow the determination of a drained behaviour, which was apparently not achieved). The apparently quasi linear increase of the peak shear strength with applied normal stress may appear satisfactory, but both sets of results have a number of apparent inconsistencies in common:

The TN16 (compression) and TN17 (extension) tests were performed at a stress state four times higher than the in-situ stress state. The response is different and failure is characterised by smaller effective friction angles (about 30◦ with 1.15 < M < 1.25). CIU tests performed on some specimens at even higher stress states have yielded effective friction angles as low as 20–25◦ (not shown here). This sharp decrease in the mechanical properties of the clay under increasing stress level is attributed to the progressive destructuration of the material. A practical consequence is that particular care should be taken in choosing a testing methodology (i.e. “in-situ stress” versus SHANSEP test procedure), depending on the geotechnical problem considered. It also confirms the original Ladd & Foott (1974) statement that clays with a high degree of structure do not exhibit a normalised behaviour because the structure is significantly altered during consolidation to high stresses.

• Data are highly scattered in the same test series; • No clear difference can be observed between slow

(drained) and fast (undrained) tests; • Rough interfaces may give a lower strength than

smooth interfaces.

Several aspects of the behaviour of Gulf of Guinea clays are still poorly understood and deserve further investigations. Work is presently ongoing in relation with the topics briefly described below.

Kuo et al. (2010) invoke the presence of faecal pellets as a source of pore water, inducing “hydroplaning” during interface shearing. This could explain the high variability observed for both peak and residual shear strengths. However, no pellets were observed in the samples tested at IFP. Obviously, more work is needed to clarify the soil-pipeline interface friction, in particular the potential for extremely low friction values when samples are sheared on a rough pipeline coating. Further Cam-shear tests on intact Gulf of Guinea clays are in progress as part of an ongoing CITEPH project.

4.9.1 Interface friction Peak and residual frictions that can be mobilised at soil-structure interfaces are important parameters for designing anchors, gravity bases or assessing the stability of pipelines or risers. Figure 31 presents results of ring shear tests (i.e. remoulded material) performed at Institut Français du Pétrole (IFP) on rough versus smooth steel interfaces and at moderate stress levels. The rates of shearing were respectively of 0.017 and 44 mm/min. On the same graph are plotted results of Cam-shear tests on undisturbed clay samples, performed with rough and

4.9.2 Rate effects Rate effects in the highly plastic Gulf of Guinea clays are significant. Figure 32 presents the results of a CIU triaxial test where the rate of shearing was varied in a ratio of 1 to 100. The response is clearly of the isotach type, meaning that the effect of strain rate changes is permanent and that separate constant rate of shearing curves can be generated. The increase in shear strength is about 18% per log cycle of shearing rate. The yield envelope is affected but not the critical state line. Rattley et al. (2010) carried out a series of constant rate of strain DSS tests on deepwater Angola

4.9

Further work

73 © 2011 by Taylor & Francis Group, LLC

2005). Work is ongoing within CITEPH, and first modelling attempts are promising. In parallel, the RASTRA model (Pereira & De Gennaro 2009, De Gennaro & Pereira 2010) reproduces correctly the rate effects, as shown in Figure 32. Efforts are now deployed to couple S3-SKH and RASTRA into a single model. 4.9.4 Effect of gas Gas may be occasionally encountered in West Africa deepwater sediments. It may be of thermogenic origin or associated to the dissociation of hydrates. A Joint Industry Project (JIP) on the behaviour of Gulf of Guinea clays containing gas was recently completed. The work, supported by BP, ExxonMobil, Total, Fugro, Ifremer and ENPC-CERMES, had the following objectives: – to investigate mechanisms of gas generation in marine sediments; – to identify and quantify the amount of free gas by correlating free gas content and acoustic properties of marine sediments; – to develop specific laboratory procedures for testing gassy soils and quantify the effects of various gas contents on the shear strength and compressibility; – to build a database of high quality tests on three different sediments to serve as reference for further analyses and interpretation by operators and academic organisations.

Figure 32. Strain rate effects on site D clay.

Whereas the results remain confidential, the JIP opens two areas for further R&D work. Firstly, the amount of free gas and the resulting damage are well detected by variations in the acoustic properties of the sediment. The availability of a tool capable of measuring in-situ and simultaneously the compressive and shear velocity of the sediments (VP and VS ) is essential to identify and characterise gassy soils. Developing such equipment is the aim of the “GEOPS” R&D project jointly conducted by Ifremer and Fugro and supported by Total. Secondly, limitations were highlighted in existing models dedicated to the description of the behaviour of gassy sediments (e.g. Sobkowicz & Morgenstern 1984, Wheeler 1988, or Grozic et al. 2005), and possible improvements were identified. Work is under way in the framework of an ongoing CITEPH project.

clay samples, with strain rates ranging between 0.003 and 300% per hour (the latter being 100 times faster than the reference DSS testing rate). A clear trend for increasing shear strength with increasing strain rate was obtained, with ratios of 16% to 19% per log cycle. According to Rattley et al. (2010), “the viscous mechanisms controlling this rate dependency appear to be consistent for both strain rate and creep behaviour, as described during one-dimensional compression, and provide evidence of the applicability of a unique stressstrain rate relation for the offshore Angola clay which is likely to be related to the ratio of its mineralogical components”. A practical consequence is that, based on specific testing procedures, optimised foundation design could be considered by taking full advantage of the increase in available soil strength under rapid loading (e.g. squalls) presented by Gulf of Guinea clays.

4.10 About the near seabed “crust” 4.10.1 Characterisation of the crust The undrained shear strength profile of deep ocean sediments is generally characterised by very low values at seabed (typically in the range of 1–2 kPa) and a quasi-linear increase with depth (gradient lying between 1.0 and 1.5 kPa/m). In some areas around the world, where there is evidence that the sediment is historically normally consolidated (NC), the shear strengths within the upper 2 m (exceptionally 3 m) of penetration are found higher than normally expected. Values of up to 10–20 kPa can be observed. The shear

4.9.3 Modelling Based on our present knowledge of the behaviour of Gulf of Guinea clays, it can be inferred that a constitutive model should be capable to reproduce effects of non-linearity at small strains, stress history, structure (initial sensitivity and rate of destructuration) and strain rate. The Baudet & Stallebrass (2004) S3-SKH model, including effects of structure and destructuration in its formulation, offers attractive features (Baudet & Ho

© 2011 by Taylor & Francis Group, LLC

74

strengths then decrease to reach progressively the “normal” linearly increasing profile typical of NC clays. The seabed zone with a “higher than expected” undrained shear strength is often called “crust”. This term will be used hereafter, but it is emphasised that the consistency of the sediment in this so-called “crust” layer remains very soft to soft, with Su in the range of 7–15 kPa. This phenomenon can be very pronounced in the Gulf of Guinea, with a clear peak in shear strength between 0.3 m and 1.2 m below seabed and a sharp decrease below that depth to reach the “normal” linear profile around 2 m of penetration (Fig. 33). It has been suggested that this peak might be an artifact due to the weight of seabed-based modules influencing the actual in-situ shear strength, but similar “crust” layers have also been highlighted by lightweight CPT modules (Peuchen 2000). Furthermore, there is now full evidence of their existence, e.g. from direct shear strength measurements in box corers (Fig. 1). Le (2008) has shown that the peak zone was associated with local decreases in water contents. 4.10.2 Origin of the crust Crusts have been found in the seabed sediments of various oceans, without any apparent correlation with physical parameters such as water depth or seabed temperature. At a particular field or site, a crust can be locally present or absent. No evidence of crustal zones has been found so far at depths in excess of 2–3 m, suggesting that either strengthening mechanisms are recent (Holocene period) or that the crust vanishes with time or sedimentation process. It is known from geological evidence that these sediments have never seen overburden stresses in excess of the present state, meaning that the relatively high shear strengths of the crust are not the result of mechanical over-consolidation. The apparent over-consolidation ratio (AOCR) is typically of the order of 3 to 5 within the crust zone. From a geotechnical point of view, it is certainly preferable to use the term yield stress ratio (YSR) which does not refer to the past effective vertical stress (section 4.1). Based on a study done by Brausse (2001), three families of mechanisms might be considered as possible drivers of apparent over-consolidation in marine sediments: (a) strong interparticle physiochemical bonding and ion exchange which may be combined with effects of cementation and cause higher attraction between particles (through ionic and van der Waals forces), (b) inorganic cementation involving cementing agents such as silicates, carbonates, iron oxides, or alumina, and (c) bioturbation and organic cementation. No unique or convincing explanation has been proposed so far. Physico-chemical effects are favoured by Sultan et al. (2001) who propose an interpretation of the apparent over-consolidation using a microscopic model based on the theory of the diffuse double layer (Gouy 1910, Chapman 1913) which relates the mechanical behaviour of the soil to the variations in ionic concentration. More recently, Ehlers et al. (2005)

© 2011 by Taylor & Francis Group, LLC

Figure 33. CPT profiles in seabed “crust” measured at various deepwater sites offshore Angola and Nigeria compared to site F with no “crust”.

have presented X-ray radiographs of surficial cores from offshore Nigeria showing a correlation between the most intensively burrowed sections and the highest soil shear strength, thus favouring the role of biological activity (bioturbation). Research work on this fascinating issue is ongoing at Cambridge University under the direction of Prof. M. Bolton. Kuo & Bolton (2009) and Kuo et al. (2010) consider the presence of faecal pellets in the studied samples as the cause of higher shear strengths. However crusts have been observed on other sites without such presence of pellets in the samples. Obviously further effort is needed to understand the origin of the crust, the conditions for its development and the reasons for its demise. The low rate of sedimentation prevailing in the Gulf of Guinea may play a particular role. 5

DESIGN ISSUES RELATED TO SOIL-PIPELINE INTERACTIONS

The purpose of this section is to describe the geotechnical considerations related to the design of flowlines (transporting an unprocessed product), pipelines (generally for a single phase fluid after processing) and risers. For the purpose of this section, the generic term “pipelines” is used. Pipelines in deep offshore areas are generally laid on the seabed under their own weight, without burial in a trench. They may experience lateral buckling and axial “walking” due to expansion caused by their operating temperatures and pressures. Analysis

75

Figure 35. Principle of the NGI model test rig.

Figure 34. Model test rig used for 1-g physical modeling tests carried out at NGI (from Dendani & Jaeck 2007).

Significant advances in the understanding of soilpipeline interactions in soft soils have come from extensive physical modelling performed in the framework of R&D projects (e.g. the SAFEBUCK JIP, Bruton et al. 2006) or sponsored by oil companies (e.g. Dendani & Jaeck 2007). Figures 34 and 35 show an example of model test rig used for 1-g soilpipeline interaction tests in soft deepwater clays from the Gulf of Guinea. Dynamic soil-pipeline stiffness tests were carried at 3S-R, Grenoble on mixtures of kaolin and bentonite aimed at approaching the mineralogical properties of GoG clays (Orozco et al. 2007). A limitation of these experiments is the incapacity to properly reproduce the natural soil structure (see section 4). It is recognised that:

of pipe-soil interaction is required to assess these phenomena. As mentioned by e.g. Cathie et al. (2005) and White & Cathie (2010), valuable experiences and progress have been gained by the geotechnical and pipeline engineering and construction communities toward a better understanding and modelling of the soil-pipeline interactions. These experiences are related to the following aspects: – site characterisation and geotechnical investigation (see section 2.3); – modelling of soil-pipeline interactions; – analysis of pipeline inspection data to improve and calibrate the interaction models. 5.1

• Initial pipeline penetration largely conditions the

axial and lateral responses and should be accurately predicted. This penetration is not only driven by the submerged weight of the pipeline but also by installation conditions (i.e. vertical overstressing, dynamic motions). • Pipeline “walking” is controlled by the longitudinal soil-pipeline friction; axial transfer curves should account for both peak shear strength and residual shear strength at large displacements. • The lateral pipeline behaviour is complex, and the transfer curves should capture first break-out resistance, post-peak resistance, additional resistance due to berm formation, and subsequent evolutions during loading/unloading cycles.

Modelling soil-pipeline interactions

Pipelines in deep water are submitted to production (start-up and shut-down) cycles. Internal variations in temperature and pressure cause successive expansions and contractions of the line which induce both axial and lateral displacements. Large cumulated displacements can be experienced, which are commonly designated by “walking” (in longitudinal direction) or “snaking” (in lateral direction). These movements generate stresses in the pipeline itself, at the connection spools and on subsea structures. This may result in fatigue damage or lateral buckling. Extensive structural analyses are performed, using both analytical and finite element analyses, in order to demonstrate that pipeline movements cannot lead to excessive stresses or that mitigation procedures are required to control the displacements. Interaction conditions between pipeline and seabed are fundamental for the integrity evaluation of the pipelines. It is now well established that simple friction coefficients are not sufficient to capture the complexity of the longitudinal and lateral behaviour of pipelines on soft soils. Interaction curves are generally incorporated into the structural models, using an approach similar to the t-z and p-y load transfer methods used for analysing the response of foundation piles. © 2011 by Taylor & Francis Group, LLC

5.2

Pipeline penetration

The assessment of pipeline embedment is fundamental for the evaluation of the axial and lateral resistances. The initial pipeline embedment depends directly on the shear strength profile in the upper half-metre of soil, and the presence of a seabed crust (see section 4.9) has a significant impact on the resulting penetration. The consequences regarding pipeline stability can be high, and heavy expenditures might be required if artificial stabilisation is needed. The embedment under self-weight may be estimated from conventional bearing capacity equations modified for the curved shape of a pipeline

76

Figure 36. Axial soil resistance model (from Dendani & Jaeck 2007). Figure 37. Lateral soil resistance model (from Dendani & Jaeck 2007).

(Randolph & White 2008, Bruton et al. 2009). However, field observations show that the as-laid pipeline embedment is generally greater than the penetration calculated from the static pipeline submerged weight alone. During laying operations, the pipeline is submitted to dynamic motions in the touch-down zone, resulting in the force applied by the pipeline being much greater than its submerged weight. A force amplification factor Ka for dynamic motion shall then be considered to reproduce the observed penetration. For a 12

pipeline on soft clay, Cathie et al. (2005) suggest a load concentration factor equal to 2.0. For the soft deepwater Gulf of Guinea sediments, a back analysis of the observed penetrations suggests an amplification factor Ka in the range of 1.5 to 2.0. Alternatively, the pipeline penetration can be assessed from model in-situ pipe tests and measurements with the SMARTPIPE® and SMARTSURF™ tools (see section 2.3).

against model tests. The break-out resistance can be divided into a frictional component, and a passive resistance component depending on pipeline embedment and undrained shear strength:

where W
= submerged pipeline weight (kN/m); Clat = lateral bearing factor; Suav = average undrained shear strength (kPa) over depth z; and z = pipeline bottom penetration (m). Dendani & Jaeck (2007) reported that Clat = 2.3 gives a fairly good match with pipeline model test results performed on Gulf of Guinea clays. The lateral behaviour of a partially penetrated pipeline is usually brittle and involves high break-out resistance followed by lower residual resistance during sweeping. During each lateral sweep, a soil berm develops ahead of the pipeline and causes a rise in lateral resistance. A schematic lateral transfer curve used for pipeline design in deepwater Gulf of Guinea soils is given in Figure 37.

5.3 Axial resistance The axial (longitudinal) resistance for a pipeline on soft clay may be estimated using an approach similar to the total stress method (α method) used for pile capacity:

5.5

Pipeline inspection data in deep offshore Gulf of Guinea obtained in several fields operated by Total show the following main results:

where α = adhesion factor; L = arc length of pipeline embedded in soil; and su = intact undrained shear strength. The intact value of undrained shear strength should be used because set-up will take place between pipeline installation and operation. Dendani & Jaeck (2007) reported that the axial resistances recorded during the model tests are best predicted by taking α = 0.7 and α = 0.35 for the peak resistance and residual resistance, respectively. An example of axial resistance curve is given in Figure 36.

• Pipeline penetration is generally in the range 30–60

% of the pipeline diameter; • Soil berms are identified (see Fig. 38); • No excessive lateral or axial movements are

observed, suggesting that the soil-pipeline interaction models used for the design are generally acceptable. 6

5.4 Lateral resistance

DESIGN ISSUES FROM INSTALLATION EXPERIENCES

The Nkossa production barge, in a water depth of 200 m offshore Congo, marked the first step of deepwater developments in the Gulf of Guinea. It was also

The break-out lateral (horizontal) resistance is conventionally assessed using empirical methods calibrated

© 2011 by Taylor & Francis Group, LLC

Inspection data

77

Table 2. Characteristics of foundations for typical Gulf of Guinea deepwater structures. Water depth (m)

Figure 38. Pipeline on seabed from a video survey in deep offshore Gulf of Guinea.

the first production FPU anchored with suction piles in soft deepwater sediments (Colliat et al. 1997). Similar to other major offshore oil provinces, deepwater Gulf of Guinea oilfields have subsequently been developed rapidly, and currently cover water depths between 400 m and about 2000 m, in particular offshore Angola and Nigeria. The main types of foundations are suction piles and driven piles, applied for the full spectrum of production structures, i.e. FPU, FPSO, OLT, TLP and CT. Because of relatively mild environmental conditions, the Gulf of Guinea has also seen the installation of two compliant towers in Cabinda (the CT concept originally described by Des Deserts in 1992 and Reusswig & Nair in 1994 was used in Cabinda in 2005, see Will et al. 2006), and the very first FPDSO for the Azurite field in 1400 m of water offshore Congo. Within existing deepwater production structures, the deep draught SPAR and semi-submersible floaters are the only two which have not yet been utilised in the Gulf of Guinea. Table 2 summarises the main characteristics of the foundations and anchors that are used for illustration in this section. All case studies given in Table 2 are from offshore Angola (or the Cabinda enclave for BenguelaBelize), with the exception of the Akpo field, located offshore Nigeria. 6.1

1300

Girassol RTA

1300

Rosa RTA

1350

Dalia FPSO

1300

Akpo FPSO

1300

Kizomba A TLP

1200

Benguela-Belize compliant tower Kizomba MODU moorings Akpo standby moorings

400

∗

1200 1300

Suction piles 4.5 m OD Gravity piles 8.0 m OD Gravity pile 8.0 m OD Suction piles 4.9 m OD Suction piles 5.0 m OD Driven piles 2.14 m (84

) Driven piles 2.74 m (108

) SEPLA 28 m2 plate VLA 15 m2 plate

Penetr. Depth (m) 16.5 19.5 23 20.5 24 123 150 25∗ 30

Installed with a suction follower pile

design principles for calculating the friction resistance along the outside and inside walls of suction piles in different types of cohesive sediments, where specific requirements for high plasticity clays apply to deepwater West Africa sediments. The environmental conditions in the Gulf of Guinea also have some design implications, some of which are common to other deepwater areas and others being more specific, in particular the cyclic undrained shear strength of the clay and the reverse end bearing capacity of suction piles in relation to the design mooring load case. 6.1.1 Cyclic undrained shear strength No systematic study of the cyclic response of Gulf of Guinea clays has been performed so far. However a compilation of existing cyclic DSS or triaxial test data allows “average” contour diagrams to be proposed. An example is given for DSS loading in Figure 39, where the response of the Drammen clay is also shown for comparison. Gulf of Guinea clays are characterised by a strong rate effect and a slow degradation process. This good resistance to cyclic loading combined with the West Africa storm composition, which for anchoring systems involves a significant static loading component and relatively low peak dynamic loads, explains why the undrained cyclic shear strength is typically found about 5–10% larger than the static reference strength. Hence, designs are de facto based on static shear strengths.

Suction piles in taut leg moorings

Subsea field development with oil storage in an FPSO on site is the preferred solution offshore West Africa. As per 2010, a total of 20 FPSO or FPU vessels and OLT buoys are operated in the Gulf of Guinea, representing the installation of about 500 suction piles in less than 10 years. To the Authors’ knowledge, all existing FPSO vessels and OLT buoys, in water depths between 200 m and 1500 m, are anchored with suction piles in semi-taut to taut leg configuration, i.e. with the mooring line making an angle of 20◦ to 45◦ to the horizontal or seafloor and a significant vertical loading component at the anchor. Andersen & Jostad (2002 & 2004) and Andersen et al. (2005) have described the

6.1.2 Reverse end bearing capacity In TLMs with a significant vertical loading component at the anchor point, reverse end bearing (REB) is an important contribution to the ultimate capacity of a

78 © 2011 by Taylor & Francis Group, LLC

Girassol FPSO

Foundation type & size

Figure 39. Normalised diagram showing number of cycles to failure for an “average” Gulf of Guinea clay and the Drammen clay (DSS mode, OCR = 1).

suction pile in clay. In early cases (e.g. the Nkossa FPU or the Girassol FPSO, Colliat et al. 1997 & 2007), REB was not considered in the pile capacity. With the increased industry confidence and perceived reliability, REB is now taken into account in full (as is usually the case in the Gulf of Mexico) or a proportion of it only (typically about 30% in the Gulf of Guinea, on the basis of the ratio of the double amplitude cyclic load to the total mooring load, as introduced by Bureau Veritas). A regional issue that has driven a certifying authority to conservatively limit REB capacity is the lack of long term monitoring of sea states and storm events, and the difference in dynamic loads between 100-year storm and squall conditions. Specifically, the design load case for the mooring of an FPSO vessel offshore Nigeria will be governed by squall wind loading and it is considered that the peak squall load will last longer than a peak storm wave load. However, both rapid and slow pull-out load tests of model piles in the centrifuge do show that high excess pore pressures develop at the pile tip, corresponding to a significant REB capacity (Colliat et al. 2010).

The same result is obtained for the FPSO piles of the D and K sites, where the predicted self-weight penetration ranged from 8.3–11.5 m to 9.6–13.5 m when the observed values were respectively equal to 12–13 m and 15–16 m (Fig. 40). From these larger self-weight penetrations, a sensitivity of 4 to 5 is back-calculated, whereas laboratory and in-situ test results (VSTs with only 2 revolutions) available at time of design suggested a sensitivity ranging between 2.5 and 4. An even lower sensitivity of about 2 would have been obtained had the remoulded shear strength been taken directly equal to the CPT sleeve friction (see discussion in section 4.4). Therefore, it is recommended to put more emphasis on tests that better mimic large strains at the soil-pile interface, such as cyclic T-bar or ball penetration tests with sufficient number of cycles (N > 25).

6.1.3 Sensitivity estimated from self-weight penetrations During penetration, either by self-weight or when suction is applied, the friction resistance along the outside and inside pile wall is assumed to be governed by the remoulded clay strength at the soil-pile interface, i.e. by the soil sensitivity. From the back-analysis of the self-weight penetration of a number of suction piles at various deepwater West Africa sites, Colliard & Wallerand (2008) suggest a clay sensitivity larger than expected.

6.1.4 Installation behaviour Figure 40 summarises the installation behaviour of three sets of FPSO piles, with a self-weight penetration equal to 50–67% of the final embedment depth, which is typical of suction piles in soft deepwater sediments. At final penetration, the suction ranges from 70–95 kPa (at 16.5 m for site S) to 100–135 kPa (at 20.5 m and 24.0 m for site D and site K, respectively). Required suction reduces with increased pile diameter (for a given penetration) since it acts over an increased internal pile cross section. With the suction piles from

Figure 40. Typical suction versus depth curves for installation of FPSO suction piles.

79 © 2011 by Taylor & Francis Group, LLC

site D and site K having similar diameter and submerged weight, together with similar undrained shear strength profiles at both sites, the lower suction measured for the K piles suggests a higher sensitivity of the deepwater clays from this site. The suction curves given in Figure 40 are typical of the penetration resistance of suction piles with a diameter in the order of 5 m and a height-to-diameter ratio of about 4 to 5 (see Table 2). Colliat (2006) showed that similar installation behaviour is obtained for suction piles in the Gulf of Mexico when the gradient in undrained shear strength is similar (i.e. of about 1.0 to 1.5 kPa/m). However, it should be noted that the suction curves shown in Figure 40 were not always measured directly inside the pile and have not been corrected for venturi effects or hydraulic losses in the pumping system; this may be the reason for the step increase of suction immediately after self-weight penetration for the D and K piles. 6.1.5 Effect of ring stiffeners The lack of heavy lift vessels in the Gulf of Guinea initially promoted fabrication solutions in favour of reducing the suction pile weight by using internal ring stiffeners. This was the case for the Nkossa FPU or the Girassol FPSO piles (Colliat et al. 1997 & 2007). When such ring stiffeners are relatively wide and close to the pile tip, it is now well known that they have a direct influence on the soil-pile interface friction. The inside friction is reduced, in relation to an increased remoulding effect and down-dragging of a mixture of seafloor soil and water, whereas the outside friction may be increased by displacing a larger volume of the inside soil plug out of the pile during penetration. For the latter effect, Andersen & Jostad (2004) describe how the normal stresses along the external pile wall are initially taken by the pore pressures, then give increased effective stresses and friction resistance as pore pressures dissipate. Pull-out tests of model piles with and without stiffeners in the centrifuge show that the net pull-out capacity of stiffened piles is 10–35% higher than for piles without ring stiffeners (see Fig. 43 and Colliat et al. 2010).

Figure 41. Examples of installation and retrieval behaviour.

Gulf of Guinea sites is described in Colliat & Colliard (2010), with set-up periods ranging between 1 day and 3½ years. The results obtained suggest a rapid increase in friction resistance, from 35–45% in one week to about 70% in one month, but little gain between one month and 3½ years. This is in relatively good agreement with (a) the results of thixotropy tests in high plasticity clays (Andersen & Jostad 2002, Dendani 2003, Fig. 28), and (b) similar data obtained at deepwater Gulf of Mexico sites (Jeanjean 2006). The interpreted interface friction factor ranges from 0.25 to 0.35 at time of installation, and increases to 0.40–0.56 in the longer term. Despite some inevitable uncertainties related to too scarce a database and other operational issues (in particular, when the measurements of the retrieval over-pressure and submerged pile weight are inaccurate), obtaining a long term friction factor below the generally accepted design value of 0.65 for high plasticity clays is a concern (where the friction factor is equal to 1.95/St with St = 3, from Andersen & Jostad 2002). A thorough confirmation or verification would require a larger set-up database, or (preferably) the performance of specific field tests.

6.1.6 Set-up behaviour Suction piles might have to be retrieved when they are installed out of tolerances (generally within one or two days after installation), or in the event of other operational constraints such as the replacement of a mooring line (i.e. several months or years after installation). In both cases, retrieval of the suction pile by over-pressurising allows a direct measurement of the increase in friction resistance with consolidation time (set-up). One example of installation and retrieval behaviour of two FPSO piles is given in Figure 41, showing an installation suction of about 90 kPa at 16.5 m depth, and an over-pressure of 180 kPa and 165 kPa at time of retrieval one and two days later, respectively. Although limited to a database of six retrieval cases, the set-up behaviour of suction piles at three different

6.2

The case of the Girassol RTA piles experiencing an abnormally low friction resistance due to the painted outside wall (contractor’s choice for protection against corrosion) has been published in detail (Dendani & Colliat 2002, Dendani 2003, Colliat et al. 2007) and is not repeated here. But, as a direct detrimental

80 © 2011 by Taylor & Francis Group, LLC

Suction piles for riser towers

Figure 42. Installation predictions and results for large RTA piles (from Colliat et al. 2007).

consequence, the design of following RTA piles was more conservative and did not rely on passive suction. For the Girassol RTA piles, the design shear strength (determined from DSS tests) was multiplied by 0.70 for the effect of constant tension loading on the undrained clay strength (thus increasing the material factor from 1.30 to 1.86), and passive suction (REB) was considered only for the short term loading case. After installation, the reduced friction capacity of the painted piles was balanced by doubling the RTA submerged weight with ballast, thus making it a kind of “gravity friction pile”. On the other hand, the RTA pile for site R was originally designed as a gravity friction pile, with the submerged weight of the RTA equal to 1.25 times the service tension load case, in order to have the RTA working under compression loading. A similar type of foundation with counter-weights is also mentioned by Zimmermann et al. (2009) for the Greater Plutonio riser tower in Angola. Figure 42 shows a comparison of the installation results and predictions for both the site S and site R RTA piles. Being the first suction piles ever installed in soft deepwater Gulf of Guinea sediments, a rather conservative prediction was done for the site S RTA piles, but the much lower penetration resistance is mainly related to the paint effect (Colliat et al. 2007). The good installation prediction for the site R RTA pile confirms that the design model is correct. However, more field experience will be required before the industry regains confidence in REB capacity for RTA piles under tension service loading.

Figure 43. Centrifuge pull-out testing of suction piles with and without stiffeners (from Colliat et al. 2010).

This RTA experience was also studied by model pull-out tests in the centrifuge, specifically performed with similar deepwater clay samples reconstituted in the centrifuge, and with emphasis on the effect of ring stiffeners on both penetration and pull-out resistance (Thorel et al. 2010). It was observed that significant passive suction is mobilised in rapid pull-out loading of both the stiffened (S) and un-stiffened (U) piles, although it is necessary to displace the pile in excess of one metre to develop 100% of the REB capacity (Fig. 43). Similar results were obtained by Raines et al. (2005) with centrifuge tests carried out with Speswhite clay samples. 6.3

With a penetration of 150 m, the foundation piles of the Benguela-Belize compliant tower (together with those of the Tombua-Landana CT in the same area) are within the longest piles ever driven in soft deepwater sediments. The 2.74 m (108

) diameter CT piles were driven with a Menck MHU 2100 hammer, and relatively low driving blow-counts (i.e. 22–24 blows/25 cm at final depth) are reported by Will et al. (2006). On another hand, with a penetration of 123 m, the 2.14 m (84

) diameter piles of the Kizomba A TLP are within the range of pile diameters and penetrations for equivalent Gulf of Mexico TLPs (Doyle 1999).

81 © 2011 by Taylor & Francis Group, LLC

Driven piles

Despite a slight difference in pile diameter (i.e. 96

for Kizomba, versus 72

–96

in the Gulf of Mexico), a fair comparison of the Kizomba pile driving behaviour is possible since the same type of hydraulic underwater hammer (Menck MHU 500T) is used in the Gulf of Mexico. For the Kizomba TLP piles, Labbe & Perinet (2004) report a lower SRD than expected and lower driving blow-counts at final depth (i.e. 28 blows/25 cm instead of a predicted value of 45 blows/25 cm). In the Authors’ opinion, this lower than expected SRD is due to the use of the Stevens et al. (1982) method, which assumes implicitly a clay sensitivity of 2 and will then provide conservative drivability predictions in soft clays with a larger sensitivity. For the Kizomba TLP piles, a better drivability prediction would have been obtained by using an alternative SRD calculation method with a friction degradation model (e.g. Puech et al. 1990, Colliat et al. 1993, or Dutt et al. 1995). The largest difference between the Kizomba and Gulf of Mexico sites is in the self-weight penetrations of the TLP piles, with 51–55 m for Kizomba (for a prediction ranging between 34 m and 52 m), as compared to a typical range of 30–40 m for similar pile sizes in the Gulf of Mexico (Doyle 1999). At Kizomba, Labbe & Perinet (2004) also report the performance of a re-drive test, showing a rapid gain of pile shaft friction resistance with time. Interestingly, similar rapid set-up of large diameter driven piles is described by Dutt & Ehlers (2009), also comparing two case studies from two deepwater sites in West Africa and in the Gulf of Mexico. These pile driving results from various Gulf of Guinea sites suggest a sensitivity of the clay higher than considered in standard driveability prediction methods and a rapid regain of resistance with time. This is in good agreement with the installation behaviour of suction piles described above.

deepwater clay, the required UHC is typically in the order of 4–5 MN for temporary MODU moorings, as compared to about 8–9 MN for semi-permanent or permanent moorings (Colliat 2006). In a recent application for three stand-by moorings for supply boats, 15 m2 Stevmanta anchors were installed in 1300 m of water offshore Nigeria, reaching an embedment depth of 28–32 m. Published Stevmanta or Dennla application cases (Ruinen & Degenkamp 2001, Murff et al. 2005, Magne 2008) mention embedment depths of 15–25 m for 10–15 m2 VLA anchors (with the smallest values of penetration and plate area corresponding to VLAs installed into stiffer clays offshore Brazil). In comparison, the result obtained with this Nigeria example suggests a lower penetration resistance in the Gulf of Guinea sediments, in a similar way to suction piles or driven piles. 7

Originally considered as an area with relatively gentle soil conditions, the deep offshore Gulf of Guinea has proven to be unique in certain aspects, highlighted by a number of differences with the Gulf of Mexico characteristics. After more than 10 years of investigations in the Gulf of Guinea, a large database on the behaviour of these deepwater sediments is now available, and has been tentatively summarised. A soil investigation strategy adapted to the regional requirements of this remote area is proposed, with emphasis on the importance of in-situ testing. Geotechnical characterisation is based on the use of conventional CPTs and VSTs, with the need for performing cyclic full-flow penetrometer tests (T-bar or ball tests) with at least 25–30 cycles for the determination of the soil sensitivity. Specific laboratory testing procedures were developed to address some key physical and geotechnical properties of these unusually high plastic clays, e.g. for the measurement of organic content and the determination of fine particle sizes. Since the Gulf of Guinea deepwater sediments have a high degree of structure, triaxial testing of high quality samples under in-situ stresses should be preferred to the SHANSEP test procedure usually applied for Gulf of Mexico clays. This high degree of structure is also the cause for measuring an apparent overconsolidation ratio of about 1.5–2 in the top 10–20 m of sediments, but the Gulf of Guinea deepwater sediments are generally normally consolidated clays (away from topographic highs created by diapirism). Two important properties of the Gulf of Guinea deepwater sediments are confirmed by the results of installation experiences, i.e. a relatively high sensitivity (St = 4 to 6), giving a lower than expected resistance to penetration of suction piles or driven piles from friction degradation, and a rapid and important regain in friction resistance with time from thixotropy and set-up effects. Some geotechnical issues still require further studies, in particular: (a) origin and characterisation of the near seabed “crust”, (b) interface soil-pipeline friction

6.4 Vertically loaded plate anchors In 1995, conventional drag anchors were used for the moorings of the Nkossa oil FSO in a water depth of 125 m offshore Congo (Colliat et al. 1997). To the Authors’knowledge, the first application of VLA plate anchors in TLM moorings in the Gulf of Guinea was for the semi permanent mooring systems of MODUs, for the drilling of production wells at the Kizomba field offshore Angola. SEPLA anchors were installed by means of a suction follower, and the same type of SEPLA anchors has then been used at the TombuaLandana field, offshore Cabinda, for the same kind of application. On another hand, both the Vryhof Stevmanta and Bruce Dennla VLAs have been used in a number of pre-installed temporary moorings for exploration drillling MODUs. For permanent moorings, VLA plates of about 25–30 m2 in surface area are generally used, whereas smaller anchors, with areas of the order of 12–15 m2 , are used in the latter case of temporary moorings. The plate area is directly related to the required anchor UHC (Murff et al. 2005). For an estimated embedment depth of about 25 m in soft

82 © 2011 by Taylor & Francis Group, LLC

CONCLUSIONS

Prof. P. Delage (ENPC-CERMES), Prof. F. Thomas (INPL/LEM), N. Sultan and J. Meunier (Ifremer) and H. Manh Le (Fugro).

resistance, (c) effect of gas and gas hydrates in the sediments, and (d) definition of a single constitutive model to reproduce the effects of stress history, structure and strain rate. These issues are targeted by ongoing R&D studies.

REFERENCES 8 ABBREVIATIONS AOCR BPT CT CPT DSS FPDSO FPSO FPU FSO ICL MIP MODU NC OCR OLT REB RTA SCL SCC SEM SEPLA SHANSEP SRD TOC TPT TLM TLP UHC VLA VST YSR

Alhayari, S., Chen, J.H. & Maerten, B. 2000. Kuito project: Successful suction pile design and installation based on innovative soil investigation. Proc. Offshore West Africa Conference, OWA, Abidjan. Andersen, K.H. & Jostad, H.P. 2002. Shear strength along outside wall of suction anchors in clay after installation. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Kyushu. Andersen, K.H. & Jostad, H.P. 2004. Shear strength along inside of suction anchor skirt wall in clay. Proc. Offshore Technology Conference, OTC paper 16844, Houston. Andersen, K.H., Murff, J.D., Randolph, M.F., Clukey, E.C., Erbrich, C., Jostad, H.P., Hansen, B., Aubeny, C., Sharma, P. & Supachawarote, C. 2005. Suction anchors for deepwater applications. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Baudet, B.A. & Ho, E.W.L. 2005. Modelling the effects of structure in deep-ocean sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Baudet B.A. & Stallebrass S.E. (2004). A constitutive model for structured clays. Géotechnique 54 (4), 269–278. Borel, D., Puech,A., Dendani, H. & De Ruijter, M. 2002. High quality sampling for deep water geotechnical engineering: the STACOR experience. Proc. Int. Conference on Ultra Deep Engineering Technology, UDET, Brest. Borel, D., Puech, A., Dendani, H. & Colliat, J.L. 2005. Deepwater geotechnical site investigation practice in the Gulf of Guinea. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Borel, D., Puech, A. & Po, S. 2010. A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Bourillet, J.F., Damy, G., Dussud, D., Sultan, N. & Woerther, P. 2007. Behaviour of a piston corer from accelerometers and new insights on quality of the recovery. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Broadbent, F.E. 1953. The soil organic fraction. Adv. Argon, 5, 153–183. Brausse, M. 2001. Underconsolidation and apparent overconsolidation in marine sediments: a study of stress states in the northwestern Gulf of Mexico. Civil Engineering MsC Thesis, University of Rhode Island. Bruton, D., White, D., Cheuk, C., Bolton, M. & Carr, M. 2006. Pipe-soil interaction behavior during lateral buckling, including large amplitude cyclic displacement test by the Safebuck JIP. Proc. Offshore Technology Conference, OTC paper 17944, Houston. Bruton, D., White, D.J., Langford, T.L. & Hill, A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments. Proc. Offshore Technology Conference, OTC paper 20096, Houston. Buckley, D.E., MacKinnon, W.G., Cranston, R.E. & Christian, H.A. 1994. Problems with piston core sampling: Mechanical and geomechanical diagnosis. Marine Geology, 117, 95–106. Burland, J.B. 1990. On the compressibility and shear strength of natural clays. Géotechnique, 40 (3), 329–378.

Apparent Over-Consolidation Ratio Ball Penetration Test Compliant Tower Cone Penetration Test Direct Simple Shear Floating Production Storage Drilling & Off-loading Floating Production Storage & Off-loading Floating Production Unit Floating Storage & Off-loading Intrinsic Compression Line Mercury Intrusion Porosimetry Mobile Offshore Drilling Unit Normally Consolidated Over-Consolidation Ratio Oil off-Loading Terminal Reverse End Bearing Riser Tower Anchor Sedimentation Compression Line Sedimentation Compression Curve Scanning Electron Microscope Suction Embedded Plate Anchor Stress History And Normalized Soil Engineering Properties Soil Resistance to Driving Total Organic Carbon T-bar Penetration Test Taut Leg Mooring Tension Leg Platform Ultimate Holding Capacity Vertically Loaded Anchor Vane Shear Test Yield Stress Ratio

ACKNOWLEDGEMENTS The authors thank Total EP, Fugro France and IFP for the permission to publish this paper. Parts of the work published come from various R&D programmes conducted in the framework of CLAROM (CLub pour les Actions de Recherche sur les Ouvrages en Mer) with financial support from the private programme for the oil & gas E&P industry CITEPH (Programme de Concertation pour l’Innovation Technologique dans l’Exploration et la Production des Hydrocarbures). Projects were carried out with the joint participation of Acergy, Doris Engineering, Ifremer, Saipem, Technip and Total, and in collaboration with several University laboratories. The contribution of colleagues from these organizations is acknowledged. The Authors are particularly indebted to V. De Gennaro and

83 © 2011 by Taylor & Francis Group, LLC

Cathie, D.N., Jaeck, C., Ballard, J-C. & Wintgens, J-F. 2005. Pipeline geotechnics – state-of-the-art. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Chapman, D.L. 1913. A contribution to the theory of electrocapillarity. Philosophical Magazine, 25 (6), 475–481. Cochonat, P., Bourillet, J.F., Quemeneur, P. & Colliat, J.L. 1996. Deepwater sedimentary processes in the Gulf of Guinea. Proc. Int. Conference on Deepwater Site Investigation, SUT, London. Colliard, D. & Wallerand, R. 2008. Design and installation of suction piles in West Africa deepwaters. Proc. 2nd BGA Int. Conference on Foundations, ICOF 2008. Colliat, J.L. 2006. Evaluation of suction piles and plate anchors from current deepwater mooring applications. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, San Francisco. Colliat, J.L., Boisard, P., Gramet, J.C. & Sparrevik, P. 1997. Geotechnical design and installation behaviour of suction anchor piles – Comparison with drag anchors. Proc. Int. Conference on Behaviour of Off-Shore Structures, Delft. Colliat, J.L. & Colliard, D. 2010. Set-up of suction piles in deepwater Gulf of Guinea clays. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Colliat, J.L., Dendani, H., Jostad, H.P., Andersen, K.H., Thorel, L., Garnier, J. & Rault, G. 2010. Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners and set-up. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Colliat, J.L., Dendani, H. & Schroeder, K. 2007. Installation of suction piles at deepwater sites in Angola. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Colliat, J.L., Vergobbi, P. & Puech, A. 1993. Friction degradation and set-up effects in hard clays offshore Congo and Angola. Proc. Offshore Technology Conference, OTC paper 7192, Houston. Cotecchia, F. & Chandler, R.J. 2000. A general framework for the mechanical behaviour of clays. Géotechnique, 50 (4), 431–447. De Gennaro, V. & Pereira, J-M. 2010. A viscoplastic constitutive model for unsaturated geomaterials. Computers and Geotechnics (under review). De Gennaro, V., Puech, A. & Delage, P. 2005. On the compressibility of deepwater sediments of the Gulf of Guinea. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Delage, P. & Lefebvre, G. (1984). Study of the structure of a sensitive Champlain clay and of its evolution during consolidation. Canadian Geotechnical Journal, 21, 21–35. Delage, P., Marcial, D., Cui, Y.J. & Ruiz, X. (2006). Ageing effects in a compacted bentonite: a microstructure approach. Géotechnique, 56 (5), 291–304. Dendani, H. 2003. Suction anchors: Some critical aspects for their design and installation in clayey soils. Proc. Offshore Technology Conference, OTC paper 15376, Houston. Dendani, H. & Colliat, J.L. 2002. Girassol: Design analysis and installation of the suction anchors. Proc. Offshore Technology Conference, OTC paper 14209, Houston. Dendani, H. & Jaeck, C. 2007. Pipe-soil interaction in highly plastic clays. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Des Deserts, L. 1992. Architecture and application of various configurations of compliant towers. Proc. ASME Int. Conference on Offshore Mechanics and Arctic Engineering, OMAE, Houston.

Diaz-Rodriguez, J.A., Leroueil, S. & Aleman, J.D. 1992. Yielding of Mexico city clay and other natural clays. Journal of Geotechnical Engineering, ASCE, 118 (7), 981–995. Doyle, E.H. 1999. Pile installation performance for four TLP’s in the Gulf of Mexico. Proc. Offshore Technology Conference, OTC paper 10826, Houston. Dutt, R.N., Doyle, E.H., Collins, J.T. & Ganguly, P. 1995. A simple model to predict soil resistance to driving for long piles in deepwater normally consolidated clays. Proc. Offshore Technology Conference, OTC paper 7668, Houston. Dutt, R.N. & Ehlers, C.J. 2009. Set-up of large diameter driven pipe piles in deepwater normally consolidated high plasticity clays. Proc. ASME Int. Conference on Ocean, Offshore and Arctic Engineering, OMAE2009-79012, Honolulu. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origin of near-seafloor “crust zones” in deepwater. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Enjaume, P., Hamon, M., Epalanga, K. & Mackenzie, B. 2010. Angolan deepwater soil conditions: GIS technology development for sediment characterization. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Evans,T.G., Feyereisen, S. & Rheaume, G. 2002.Axial capacities of jetted well conductors in Angola. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Evans, T.G. 2010. A systematic approach to offshore engineering for multiple-project developments in geohazardous areas. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Gouy, G. 1910. Sur la constitution de la charge électrique à la surface d’un électrolyte. Journal de Physique Théorique et Appliquée, 9, 457–468. Grozic, J.L.H., Nadim, F. & Kvalstad, T.J. 2005. On the undrained shear strength of gassy clays. Computers and Geotechnics, 32 (7), 483–490. Hill, A.J., Southgate, J.G., Fish, P.R. & Thomas, S. 2010. Deepwater Angola: Geohazard mitigation. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Hill, A.J. & Wintgens, J-F. 2009. In-situ measurement of pipesoil interaction in deep water – Results of a successful offshore campaign. Proc. SUT Annual Conference, Perth. Holtvoeth, J. 2004. Terrigenous organic matter in sediments of the eastern equatorial Atlantic – Distribution, reactivity and relation to Late Quartenary climate. PhD thesis, Bremen University. Jeanjean, P. 2006. Set-up characteristics of suction anchors for soft Gulf of Mexico clays: Experience from field installation and retrieval. Proc. Offshore Technology Conference, OTC paper 18005, Houston. Kelleher, P. & Hull, T 2008. Quality assessment of marine sediments recovered with a hydraulically tethered piston corer. Proc. Offshore Technology Conference, OTC paper 19687, Houston. Kuo, M.Y-H. & Bolton, M.D. 2009. Soil characterization of deep water African clays: Is biology a source of mechanical strength?. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Osaka. Kuo, M.Y-H., Bolton, M.D., Hill, A.J. & Rattley, M. 2010. New evidence for the origin and behaviour of deep ocean ‘crusts’. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth.

84 © 2011 by Taylor & Francis Group, LLC

Labbe, J.P. & Perinet, D. 2004. The deepwater installation of the Kizomba-A SWHP anchor piles and related constraints for ultra deepwater lifting. Proc. Int. Conference on Deep Offshore Technology, DOT, New Orleans. Ladd, C.C. & Foott., R.1974. New design procedure for stability of soft clays. Proc. Journal of the Geotechnical Engineering Division, ASCE, 100 (7), 767–787. Le, M.H. 2008. Caractérisation physique et mécanique des sols marins d’offshore profond. PhD thesis, ENPCCERMES. Le, M.H., Nauroy, J.F., De Gennaro, V., Delage, P., Thanh, N., Flavigny, E., Colliat, J.L., Puech, A. & Meunier, J. 2008. Characterization of soft deepwater West Africa clays: SHANSEP testing is not recommended for sensitive structured clays. Proc. Offshore Technology Conference, OTC paper 19193, Houston. Lunne, T. & Andersen, K.H. 2007. Soft clay shear strength parameters for deepwater geotechnical design. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Lunne, T., Berre, T.V. & Strandvik, S. 1998. Sample disturbance effects in deep water soil investigations. Proc. Int. Conference on New Frontiers in Offshore Site Investigation and Foundation Behaviour, SUT, London. Magne, E. 2008. Pre-installed mooring systems for MODUs. Offshore Asia Conference, Kuala Lumpur. Montarges, R., Fay, J.B. & Le Tirant, P. 1987. Soil reconnaissance at great water depth. Proc. Int. Conference on Deep Offshore Technology, DOT, Monte Carlo. Murff, J.D., Randolph, M.F., Elkhatib, S., Kolk, H.J., Ruinen, R., Strom, P.J. & Thorne, C. 2005. Vertically loaded plate anchors for deepwater applications. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Orozco, M., Foray, P. & Nauroy, J.F. 2007, Pipe-soil horizontal dynamic stiffness in soft soils. Proc. 17th International Offshore Polar and Engineering Conference, ISOPE, Lisbon, Portugal. Osborne, J.J., Yetginer, A.G., Halliday, T. & Tjelta, T.I. 2010. The future of deepwater site investigation: seabed drilling technology? Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Pereira, J.M. & De Gennaro, V. 2009. Time dependent behaviour of fluids filled geomaterials: application to reservoir formation. Proc. 4th Biot Conference on Poromechanics, Poromechanics IV, Columbia University, New York. Peuchen, J. 2000. Deepwater cone penetration tests. Proc. Offshore Technology Conference, OTC paper 12094, Houston. Puech, A., Colliat, J.L., Nauroy, J.F. & Meunier, J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Puech, A., Orozco-Calderon, M. & Foray, P. 2010. Mini T-bar testing at shallow penetration. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Puech, A., Poulet, D. & Boisard, P. 1990. A procedure to evaluate pile drivability in the difficult soil conditions of the southern part of the Gulf of Guinea. Proc. Offshore Technology Conference, OTC paper 6237, Houston. Quiros, G.W. & Little, R 2003. Deepwater soil properties and their impact on the geotechnical program. Proc. Offshore Technology Conference, OTC paper 15262, Houston. Raines, R.D., Ugaz, O. & Garnier, J. 2005. Centrifuge modelling of suction piles in clay. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth.

Randolph, M., Cassidy, M., Gourvenec, S. & Erbrich, C.H. 2005. Challenges of offshore geotechnical engineering. Proc. Int. Conference on Soil Mechanics and Geotechnical Engineering, ICSMGE, Osaka. Randolph, M.F., Hefer, P.A., Watson, P.G. & Geise, J.M. 1998. Improved seabed strength profiling using T-bar penetrometer. Proc. Int. Conference on New Frontiers in Offshore Site Investigation and Foundation Behaviour, SUT, London. Randolph, M.F., Low, H.E. & Zhou, H. 2007. In situ testing for design of pipelines and anchoring systems. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Randolph, M.F. & White, D.J. 2008. Pipeline embedment in deep water: processes and quantitative assessment. Proc. OffshoreTechnology Conference, OTC paper 19128, Houston. Rattley, M.J., Hill, A.J., Thomas, S. & Sampurno, B. 2010. Strain rate dependent simple shear behaviour of deepwater sediments in offshore Angola. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Robertson, P.K. 2009. Interpretation of cone penetration tests – A unified approach. Canadian Geotechnical Journal, 46, 1337–1355. Ruinen, R. & Degenkamp, G. 2001. Anchor selection and installation for shallow and deepwater mooring systems. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Stavanger. Reusswig, G.H. & Nair, N.N. 1994. A compliant tower design for mild environments. Proc. ASME Int. Conference on Offshore Mechanics and Arctic Engineering, OMAE, Houston. Sobkowicz, J.C. & Morgenstern, N.R.1984. The undrained equilibrium of gassy sediments. Canadian Geotechnical Journal, 21, 439–446. Stallebrass, S.E.. & Taylor, R.N. 1997. The development and evaluation of a constitutive model for the prediction of ground movements in overconsolidated clay. Géotechnique 47 (2), 235–253. Stevens, R.S., Wiltsie, E.A. & Turton, T.H. 1982. Evaluating pile drivability for hard clay, very dense sand, and rock. Proc. Offshore Technology Conference, OTC paper 4205, Houston. Sultan, N., Cochonat, P., Cauquil, E. & Colliat, J.L. 2001. Apparent overconsolidation and failure mechanisms in marine sediments. Proc. OTRC Int. Conference Honoring Prof. W. Dunlap, Houston. Sultan, N., Marsset, B., Ker, S., Marsset, T., Voisset, M., Vernant, A.M., Bayon, B., Cauquil, E., Adamy, J., Colliat, J.L. & Drapeau, D.H. 2010. Hydrate dissolution as a potential mechanism for pockmark formation in the Niger delta. Proc. Journal of Geophysical Research, Solid Earth (in press). Sultan, N., Voisset, M., Marsset, T., Vernant, A.M., Cauquil, E., Colliat, J.L. & Curinier, V. 2007. Detection of free gas and gas hydrate based on 3D seismic data and cone penetration testing: An example from the Nigerian continental slope. Marine Geology, 240, 235–255. Thomas, F., Nauroy, J.F., Rebours, B & Meunier, J. (2005). Mineralogical characteristics of the Gulf of Guinea deep water sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Thomas, F., Puech, A., Nauroy, J.F., Palix, E. & Meunier, J. 2007. Specific identification test procedures for deepwater sediments of Gulf of Guinea. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London.

85 © 2011 by Taylor & Francis Group, LLC

Thorel, L., Dendani, H., Garnier, J., Colliat, J.L. & Rault, G. 2010. Installation process of suction anchors with and without stiffeners in Gulf of Guinea clay: centrifuge modelling. Proc. Int. Conference on Physical Modelling in Geotechnics, ICPMG, Zurich. Wheeler, S.J. 1988. A conceptual model for soils containing gas bubbles. Geotechnique, 38, 389–397. White, D.J. & Cathie, D.N. 2010. Geotechnics for subsea pipelines. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. White, D.J., Hill, A.J., Westgate, Z. & Ballard, J-C. 2010. Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Will, S.A., Mascorro, E., Robertson, W, Hussain, K. & Paulson, S. 2006. Benguela-Belize compliant tower:Tower

© 2011 by Taylor & Francis Group, LLC

design. Proc. Offshore Technology Conference, OTC paper 18068, Houston. Wong, P.C., Taylor, B.B., & Audibert, J.M.E. 2008. Differences in shear strength between Jumbo piston core and conventional rotary core samples. Proc. Offshore Technology Conference, OTC paper 19683, Houston. Young, A.G., Honganen, C.D., Silva, A.J. & Bryant, W.R. 2000. Comparison of geotechnical properties from largediameter long cores and borings in deep water Gulf of Mexico. Proc. Offshore Technology Conference, OTC paper 12089, Houston. Zimmermann, C.A., Layrisse, G., De La Cruz, D. & Gordonnat, J. 2009. Greater Plutonio riser tower installation – Studies and lessons learnt. Proc. ASME Int. Conference on Ocean, Offshore and Arctic Engineering, paper OMAE2009-79028, Honolulu.

86

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Geotechnics for subsea pipelines D.J. White Centre for Offshore Foundation Systems, University of Western Australia, Perth

D.N. Cathie Cathie Associates, Brussels, Belgium

ABSTRACT: The geotechnical analysis performed for subsea pipeline design involves challenges that are not common in conventional foundation engineering. This paper reviews recent research in pipeline geotechnics and shows examples of how this research is being applied in practice. A general theme running through this paper is the twin challenges of the changes in seabed topography and the changes in soil properties that occur through the installation and operating life of a pipeline. Results from in situ and element testing of soils that replicate the loading and disturbance imposed by pipelines are used to show that significant changes in strength are induced. Soil generally weakens during the episodes of remoulding that accompany pipeline laying, buckling, walking and storm loading, and during ploughing and trenching. The soil strength recovers during subsequent episodes of reconsolidation between storms, and between startup and shutdown events. Solutions for incorporating this behaviour into the estimation of axial and lateral pipe-soil resistance, and the assessment of trenching and ploughing operations, are discussed. A unifying theme is the relative magnitude of drained and undrained soil strengths, the evolution of these strengths through cyclic episodes, and the importance of recognising the widelyvarying rates of shearing involved in pipe-soil processes. Pipeline geotechnics can involve drained behaviour in fine-grained clayey soils – for example, during slow axial expansion of pipelines – and undrained behaviour in coarse-grained soils – for example during ploughing. Concepts from critical state soil mechanics often provide a simple framework for clarifying this behaviour.

1

INTRODUCTION

soft fine-grained soils, where the management of thermal and pressure-induced expansion is a critical design issue. The second is the stability of light large-diameter pipelines in shallow water, where primary and secondary stabilization measures represent a significant capital expenditure, and where geotechnical analysis techniques are not well established. The first scenario is relevant to pipeline design in almost all deepwater frontiers globally.The second scenario is particular relevant off the coast of Australia, where many hundreds of kilometers of gas trunkline are currently planned, to bring gas from deepwater fields to onshore LNG facilities. Section 1.2 provides a brief overview of some key pipeline design considerations that have a strong geotechnical influence. This overview is intended to provide a brief introduction to some of the most novel aspects of pipeline geotechnics, for those who are unfamiliar with this area of geotechnics. In Section 1.3 comparisons are made between pipeline geotechnics and conventional foundation engineering. Some of the most relevant aspects of soil behaviour are highlighted in Section 2, using recent experimental observations, including some from unusual forms of penetration testing. These observations provide a backdrop to the mechanisms and analysis techniques that are outlined in Sections 3, 4 and 5 for assessing

1.1 Scope of paper The purpose of this keynote paper is to set out the challenges of pipeline geotechnics and to highlight some recent developments in this area that the authors have been involved with. This is not an exhaustive treatment of the subject, but is intended to – provide an overview of areas in which intensive research has recently been published. – highlight certain novel analysis techniques for design that this research is beginning to permit. More complete introductions to pipeline geotechnics are provided by the relevant chapters of books on offshore geotechnical engineering by Randolph and Gourvenec (2010) and Dean (2010). Cathie et al. (2005) presented a more exhaustive review of research across the whole of pipeline geotechnics. The present paper is intended to be of value to pipeline engineers as well as geotechnical specialists, and includes some basic geotechnical content where we consider this useful. The topics in this paper are generally relevant to one of two design scenarios. The first is high pressure high temperature pipelines laid in deepwater on

87 © 2011 by Taylor & Francis Group, LLC

Figure 1. Some geotechnical aspects of pipeline design.

pipeline embedment, lateral pipe-soil resistance and axial pipe-soil resistance respectively. Sections 6 and 7 are focussed on the geotechnics of pipeline trench construction by ploughing and jetting respectively. Space limitations preclude discussion of in-trench pipeline stability, and the associated upheaval, backfill liquefaction and flotation issues. The paper finishes with brief conclusions. 1.2

Pipe-soil interaction processes

Many of the areas of pipeline design that have geotechnical aspects are illustrated in Figure 1. Offshore pipelines are often left on the seabed, unburied, if this does not lead to unacceptable instability under hydrodynamic loading. The interaction between the pipeline and the seabed feeds into many aspects of the pipeline design. If the pipeline must be buried, for stability or to avoid fishing gear, the shielding of the pipeline via the construction of a trench (possibly backfilled) requires geotechnical design. On-bottom pipelines are increasingly being designed to allow movement during their operation, either under hydrodynamic loading or under thermal and pressureinduced expansion. Steel catenary risers, which are extensions of pipelines that connect to surface facilities, inevitably move where they touchdown on the seabed, in response to oscillation of the floating facility. Throughout the lay process and during subsequent operating cycles, the pipeline is subjected to geotechnical forces where it is in contact with the

88 © 2011 by Taylor & Francis Group, LLC

seabed. A pipeline is a forgiving structure, being able to tolerate significant deformation and gross movements across the seabed, except at points of fixity such as end terminations. Such instabilities are unacceptable for the platforms and foundations that conventional geotechnical engineering is equipped to design. Indeed, if a pipeline was not permitted to move across the seabed under thermal loading, this would often induce unacceptable thermally-induced stresses. The forgiving flexibility of a pipeline does not, therefore, alleviate the need to quantify the pipesoil resistance forces to a sufficient accuracy that the robustness of the design is demonstrated. One of the most difficult aspects of pipeline design, which is an increasing challenge as operating temperatures and pressures rise, is the management of thermal and pressure-induced loading. Controlled on-bottom lateral buckling is an attractive design solution but one which requires the pipe-soil responses to be bracketed: both high and low geotechnical resistance can hamper a design (Bruton et al. 2007; AtkinsBoreas 2008). A second and related behaviour that arises from the thermal and pressure-induced loading is the tendency for pipes to ‘walk’ axially over cycles of startup and shutdown (Tornes et al. 2000; Carr et al. 2006). This phenomenon can be driven by the asymmetry of the heat-up and cool-down processes or by the presence of a seabed slope or end-of-line tension (which creates an asymmetry in the mobilized axial pipe-soil resistance).Accurate assessment of the axial pipe-soil resistance forces is required for robust modelling of this process.

Table 1.

Comparison of pipeline geotechnics and conventional foundation engineering (after White and Gaudin 2008). Foundation

On-bottom pipeline

Problem geometry

Known, controlled.

Design criteria for in-service behaviour

To remain fixed, movement uD.

Surrounding soil conditions

Similar to in situ state. Relatively unaffected by installation.

Soil-structure interaction

Usually minimal. Imposed loads are not strongly affected by foundation displacements. Scour and wave-induced liquefaction may require mitigation Usually available. Can assume lowest credible geotechnical capacity

Uncertain. Embedment affected by lay process and metocean conditions. Subsequent pipeline movements disturb seabed topography. May be required to displace significantly, u D, through hundreds of cycles of operation or hydrodynamic loading. Soft soil is significantly affected by installation. Remoulding, heave and reconsolidation affect the local strength. Often significant. Local pipe-soil load-displacement relationship affects overall pipeline response.

Soil-ocean interaction Single conservative design approach

Another significant design issue that is particularly relevant in the shallow waters offshore Australia, is pipeline stability under hydrodynamic loading from storm-induced currents and waves. In this situation a conservative approach is to adopt a low value of soil resistance. However, the cost of stabilization measures such as concrete coating is huge, and there is a strong incentive to refine the geotechnical analysis to remove any unnecessary conservatism in the design seabed resistance. In shallow water a pipeline may require additional – ‘secondary’ – stabilisation for hydrodynamic stability. Secondary stabilisation solutions revolve around reducing the hydrodynamic loading and increasing the available lateral resistance. An open trench provides partial shielding from hydrodynamic load. Burial of the pipe eliminates direct hydrodynamic loading (although soil liquefaction under hydrodynamic loading can destabilise a buried pipe). Geotechnical assessments must be made of the trenching process – which may be by ploughing, cutting, jetting, dredging or a combination. Other secondary stabilisation techniques include continuous rockdumping, or engineered solutions to provide local anchoring at intervals along the pipe. These solutions include flexible concrete mattresses, anchor blocks or saddles placed over the pipeline, or small piles on either side of the pipeline. The stability of these objects must also be assessed in design, taking account of the additional cyclic loading transferred to them by the unstable pipeline.

foundations and piles. This is partly because it is only recently that design codes have permitted gross pipeline movement, and so designers have not needed to explicitly assess the interaction forces as pipelines sweep across the seabed. Also, it is only recently that some of the complexities of the underlying soil behaviour have been recognized. A generally accepted framework for routine analysis has not emerged. The contrasts between pipeline geotechnics and conventional foundation engineering are summarised in Table 1 and illustrated in Figure 2. The designer’s task in the geotechnical design of a pipeline is aided by the structure’s tolerance of movements and mild deformation, but is hampered by the difficulty of assessing the geometry of the scenario and the operative soil properties. The laying of a pipeline and any subsequent lateral or axial movements disturb the topography of the seabed. The changed geometry and the altered soil properties need to be captured in calculations of the available pipe-soil resistance. Even the intact soil properties are difficult to establish at the shallow embedments relevant to pipeline geotechnics. Undisturbed sampling of soft near-surface soils is difficult and penetrometer tests at shallow embedment require particular interpretation techniques (Puech and Foray 2002, White et al. 2010a). Some soil properties and parameters such as friction angles and undrained strength ratios tend to be different at the very low stress levels relevant to pipeline geotechnics. A further complication is that interaction between the ocean and the seabed – leading to scour and liquefaction – can be significant in shallow water. The result is a tripartite interaction between the ocean, the pipeline and the seabed, which is illustrated in Figure 3. Cross-disciplinary design approaches for

1.3 Comparison with foundation engineering Geotechnical design procedures for pipelines and risers are relatively undeveloped compared to

89 © 2011 by Taylor & Francis Group, LLC

Scour and wave-induced liquefaction can dominate behaviour Often unavailable. Both upper and lower bound geotechnical capacity may adversely effect structural response

Figure 3. Tripartite interaction between the seabed, the ocean and a pipeline in shallow water during storms.

buckles to sweep laterally across the seabed – and storms – which create high hydrodynamic loading. Over the operating life of a pipeline, the surrounding soil may therefore be subjected to a large number of episodes of disturbance followed by recovery and reconsolidation. The soil strength will generally fall and rise with each episode, and the net effect may be an overall increase or decrease in the strength of the soil. Coupled with the associated changes in seabed topography, this may lead to a rise or a fall in the geotechnical restraint on the pipeline. Figure 4a illustrates the case of a buckling pipeline on a fine-grained soil (in deepwater, where hydrodynamic loading is negligible). Each startup or shutdown of the buckle causes gross monotonic remoulding of the surrounding soil, comparable to the high disturbance created during passage of a penetrometer. The startup and shutdown episodes are separated by a period of time which may or may not be sufficient for full dissipation of the excess pore pressures generated during the previous disturbance – depending, obviously, on the consolidation characteristics of the soil relative to the frequency of the startups and shutdowns. Typical pipeline designs require several hundred or even one thousand shutdown and startup events to be considered. Figure 4b illustrates the case of a pipeline on a sandy or silty soil in shallow water, where storms create high hydrodynamic loading, which in turn causes the pipeline to exert cyclic loads on the surrounding soil. A storm loading event is perhaps best considered as a pre-failure cyclic disturbance as distinct from the gross (undrained) monotonic remoulding of the previous example. In this illustration the cyclic loading leads to a weakening, associated with pore pressure buildup, which is subsequently compensated for by reconsolidation and an associated densification of the soil. Storms occur at a frequency such that full dissipation occurs between events, for the soil considered here. In design, it is necessary to consider the effect

Figure 2. Comparison of pipeline geotechnics and conventional foundation engineering (images from Jayson et al. 2008 and Fisher and Cathie 2003).

this interaction are in their infancy. There is not yet a routine basis for assessing pipeline stability under hydrodynamic action that incorporates all three interactions concurrently (Damgaard and Palmer 2001, Cheng et al. 2010).

2 2.1

RELEVANT SOIL MECHANICS Illustrations of soil behaviour near pipelines

The soil close to a pipeline is grossly disturbed as the pipe is laid. If that disturbance happens rapidly enough for excess pore pressure to be generated then the subsequent reconsolidation process generally leads to an increase in the strength and density of the soil. Subsequent events may disturb the pipeline and the surrounding soil further. These events include the startup and shutdown of the pipeline – which cause

90 © 2011 by Taylor & Francis Group, LLC

Figure 5. Changes in the undrained penetration resistance of fine-grained soils during cyclic penetrometer tests (after Gaudin and White 2009).

remoulded state, with this ratio being termed the sensitivity. Much higher sensitivities can sometimes be found, particularly in carbonate soils. In the analysis of problems involving significant disturbance, it is necessary to identify the relevant soil strength, which may lie somewhere between the intact and remoulded values. Cyclic T-bar or ball penetrometer tests allow this behaviour to be quantified. The progressive reduction in net bearing resistance through cycles of disturbance is shown for various different soils in Figure 5. The strength degrades exponentially with the number of cycles of disturbance, which allows the response to be characterized by two parameters – the sensitivity, St (or its inverse, δrem ) and a parameter related to the ductility, N95 (which is the number of cycles of disturbance after which the resistance has decayed by 95% of the difference between the intact and remoulded values). It is evident that both of these parameters vary significantly between soil types (particularly St ), but the form of the decay is similar. Analysis techniques for design can capture the decaying soil strength by adopting a value that represents the relevant level of disturbance. The general approach is to firstly convert the penetrometer ductility parameter, N95 , to an equivalent strain level (e.g. Zhou and Randolph 2009). The relevant strain level for the problem being considered is then used to deduce the operative undrained strength. Such techniques have been proposed for spudcan penetration (Erbrich 2005, Hossain and Randolph 2009). These methods utilise the type of strength degradation curves shown in Figure 5 to link the strains and operative strength around a spudcan to those around a T-bar. T-bar and spudcan penetration involves a comparable level of disturbance to monotonic pipe embedment. However, the dynamic motions that accompany pipe laying mean that a greater level of disturbance and hence a lower operative soil strength is applicable, compared to that mobilised during initial T-bar penetration. The dramatic reductions in strength evident in Figure 5 are partly due to the generation of positive

Figure 4. Illustrative histories of soil element behaviour near unstable pipelines: episodes of disturbance and recovery.

of these disturbances on the available lateral pipe-soil resistance – which controls the stability. This stability is affected by changes in the pipeline embedment as well as the changes in the strength of the soil around the pipe during disturbance. The same history of disturbance and reconsolidation shown in Figure 4b is applicable to the backfill above a trenched pipeline. These illustrations of the history of soil behaviour are clearly very idealized, and it is rare that the related changes in soil strength are tracked explicitly within a design analysis. However, it is important to recognize these effects, since they have a significant influence on the geotechnical restraint on a pipeline. These illustrations provide a convenient background to the following three examples of soil element behaviour. These examples show the changes in soil strength that can accompany episodes of disturbance and recovery, and also showcase novel testing techniques that may in the future be utilized to quantify this behaviour for design. The examples are: – episodes of undrained disturbance and reconsolidation seen by a T-bar penetrometer; – episodes of undrained disturbance and reconsolidation seen by a vertical rod penetrometer; – drained and undrained failure of different soils; – low stress friction response of fine-grained soils.

2.2 Disturbance and recovery: T-bar tests The strength of soft fine-grained seabed soils typically reduces by a factor of 2–5 from the intact to the fully

© 2011 by Taylor & Francis Group, LLC

91

Figure 6. Undrained strength through episodes of remoulding and reconsolidation (test in lightly overconsolidated kaolin in the UWA beam centrifuge) (White and Hodder 2010).

Figure 7. Critical state interpretation of episodes of remoulding and reconsolidation (White and Hodder 2010).

the attraction of setting this behaviour within an effective stress framework, but it does rely on a very crude simplification of the overall behaviour, in which the response of all the elements of soil around a penetrometer is lumped into a single representative effective stress level and specific volume. More refined models will allow this behaviour to be more accurately quantified and numerical simulations will test the validity of this simplification. The key point, however, is that the rises in soil strength during episodes of reconsolidation can eclipse the reductions in soil strength during the preceding episodes of remoulding.

excess pore pressure in these contractile materials during undrained shearing. As this positive pore pressure dissipates and the effective stress rises back to the geostatic state the material densifies and the subsequent undrained shear strength may be higher. Cyclic T-bar penetrometer tests with periods of reconsolidation between episodes of cycling show this regain in strength. Figure 6 summarises the results of a cyclic T-bar test in kaolin clay reported by White and Hodder (2010), expressing the T-bar strength at a particular depth – 2.25 m – during each cycle. After just three episodes of full remoulding and reconsolidation, the current remoulded strength was comparable to the original intact strength. These results quantify the contrasting effects of disturbance and recovery shown in Figure 4a for this soil and the particular disturbance pattern imposed by a T-bar. This behaviour is easily understood within a critical state-type framework, since this provides an explicit link between moisture content (which reduces as positive pore pressures dissipate) and undrained strength. This interpretation can be extended to a quantitative treatment, expressed in terms of the operative strength averaged over all of the soil near the penetrometer (rather than of a single soil element). An accurate back-analysis of the results shown in Figure 6 can be achieved by defining two failure lines in stressvolume space, which represent the intact and fully remoulded strengths of the soil, as proposed by White and Hodder (2010). As shown in Figure 7, this back-analysis of the T-bar resistance at a depth of 2.25 m (which corresponds to
an in situ effective vertical stress of σvo = 12 kPa) is based on the intact strength line (ISL) being reached during the initial T-bar stroke of a episode, and the effective stress point migrating towards the remoulded strength line (RSL) according to an exponential trend (i.e. the reduction in effective stress per T-bar stroke is proportional to the difference between the current effective stress and the effective stress at the remoulded state for the current specific volume). This analysis has

2.3

The example above involves only 3 episodes of reconsolidation. A significantly larger number of episodes of reconsolidation are involved in the second example. A novel vertical rod penetrometer has been used on a recent centrifuge project at UWA, with the aim of quantifying the resistance and strength of surficial material, as the soil is forced to flow past the penetrometer. This device is a cylindrical bar, oriented vertically, 4 mm in diameter. The device is embedded until the tip is typically 5 – 10 diameters below the soil surface. The bar is equipped with multiple levels of strain gauging located above the soil surface, which allow the magnitude and distribution of the pressure on the bar to be derived. In one test the bar was embedded in soft kaolin to a depth of 45 mm then cycled laterally by a distance of 20 mm at a rate of 0.3 mm/s. This rate corresponds to a dimensionless velocity of vD/cv ∼ 10 which is almost fully undrained, based on the limits demonstrated by Finnie and Randolph (1994) (albeit for a different geometry of problem). The elapsed time between the bar passing the mid-point of each lateral stroke was 100 seconds, which corresponds to a dimensionless consolidation time of T = cv t/D2 = 0.5. This value is indicative of significant (∼50%) pore pressure dissipation, based on limits provided by Randolph (2003) (again for a slightly different geometry of problem).

92 © 2011 by Taylor & Francis Group, LLC

Disturbance and recovery: vertical rod tests

Figure 8. Lateral resistance on a vertical bar penetrometer through episodes of disturbance and reconsolidation.

Figure 9. Drained and undrained resistance during penetration.

The resulting variation in the average lateral resistance on the bar penetrometer is shown in Figure 8. Initially the resistance reduces, as the remoulding damage exceeds the recovery from reconsolidation. However, a steady rise in resistance is soon evident, with the strength after several episodes exceeding the initial (intact) value. After this test the level of the soil surface around the penetrometer had lowered. This is evidence that the soil surrounding the device had densified through the episodes of reconsolidation, which is consistent with the changing strength. This example highlights again the changes in soil strength that can occur through the episodes of disturbance and recovery that can be imposed on the soil surrounding a seabed pipeline. As shown in Figure 8, the strength continued to increase through more than 25 cycles of heavy remoulding, but ultimately should reach a limiting value, when the soil has densified sufficient that there is no longer a tendency to generate positive excess pore pressure when disturbed.

plough share (Peng and Bransby 2010). In all cases the velocity is normalized by the coefficient of consolidation of the seabed and an appropriate drainage distance – generally the size of the foundation. There is some variation in the dimensionless velocities at which fully drained and fully undrained behaviour occurs, although this may be partly due to the difficulties in establishing appropriate values of the coefficient of consolidation. There are more significant differences between the relative magnitude of the drained and undrained resistance. These arise from the state of the soil – and hence its tendency to generate positive or negative pore pressure in undrained conditions – and also the particular boundary value problem. For example, for a soil with a particular undrained strength and angles of friction and dilation, the relative magnitude of the drained and undrained bearing capacity depends on both the applicable bearing factors (i.e. Nq and Nγ for drained conditions and Nc for undrained conditions) as well as the soil strength properties. The relative magnitude of the drained and undrained strengths of an interface is not affected by the geometry of the problem, so provides a more simple differentiation between the drained and undrained behaviour of a particular soil (albeit in combination with a particular interface). Figure 10 shows the steady residual resistance measured during monotonic shearing of normally consolidated kaolin clay over a rough steel surface at different velocities. The kaolin was normally consolidated to a stress of 2.5 kPa prior to shearing. These tests used a direct shear box at UWA that has been modified to operate at the low stress levels relevant to pipeline geotechnics. The four tests show a trend of increasing resistance with reducing velocity. This trend is consistent with a hyperbolic backbone curve of the same form as used in Figure 9, drawn between the fully drained and fully undrained limits. These limits correspond to an interface friction angle of 30◦ and an undrained
= 0.25, both of which are constrength ratio of su /σvc sistent with other published results for this stress level

2.4 Drained and undrained soil responses The shear strength of a given soil in a particular state depends on whether drained or undrained conditions are imposed, as well as the mode of shearing. In conditions in which drainage is permitted, the shearing can be imposed at rates that span from fully drained (i.e. in which no excess pore pressure builds up) to fully undrained (i.e. in which effectively no pore pressure dissipation occurs, despite drainage being permitted). Various authors have recently explored the continuous variation in mobilized soil strength and geotechnical resistance between drained and undrained conditions. Results summarized in Figure 9 show the variation in penetration resistance of circular surface foundations (Finnie 1993), and cone penetrometers in soft clay (Randolph and Hope 2004) and dense silt (Silva 2005). Similar relationships have been derived for the uplift resistance of buried pipelines (Bransby and Ireland 2009) and the sliding resistance of a pipeline

93 © 2011 by Taylor & Francis Group, LLC

Figure 11. High effective stress friction and a non-linear failure envelope during low stress interface tests and axial pipe-soil movement (data from Bruton et al. 2009, White et al. 2010b).

Figure 10. Interface shear resistance at varying rates, from drained to undrained.

(Bolton and Barefoot 1997; Pedersen et al. 2003; Bolton et al. 2009). The movements of an on-bottom pipeline in a given soil may span the ranges of velocities and therefore drainage conditions shown in Figure 9 and Figure 10. Thermal expansions and the associated lateral movements span from zero at ‘virtual anchor points’ to millimetres per second of axial movement near ends and buckles, and even metres per second of lateral movement during initiation of lateral buckles. A further range of velocities arises from pipeline movements driven by hydrodynamic action – which include oscillations in the touchdown zone during laying that are created by vessel motion, or oscillations in response to direct hydrodynamic action on the pipe during storms. A high velocity applies during the ploughing of a pipeline trench. In combination with the large size of a ploughshare, this leads to undrained conditions even in sands. As a consequence of the varying velocities involved in these processes, it is common for the geotechnical analysis for pipeline design in a fine-grained soil to require an assessment of the drained response. Conversely, in coarse-grained soils an assessment of the undrained behaviour can be required. Also, there are often occasions when the actual response involves partial drainage, and it is necessary to tie together drained and undrained assessments in order to predict the most likely behaviour, and the potential range of responses.

2.5

The same trend appears in low stress soil-soil and soil-interface shearing of fine-grained soils (Pedersen et al. 2003, White and Randolph 2007, Hill and Jacob 2008, Bruton et al. 2009, White et al. 2010b). Nonlinear failure envelopes that express the friction angle or limiting stress ratio as a function of effective stress can capture this variation (e.g. Figure 11). It is important to recognise that the friction angles measured at conventional geotechnical stresses may not be appropriate for the assessment of drained pipe-soil resistance. The high friction angles are also reflected in the higher undrained strength ratios found at low stresses – a link highlighted in the approximate expression for normally consolidated undrained
strength ratio su /σvc = φ/100 derived from the analysis of Wroth (1984), where φ is the friction angle in
is the consolidation stress. Changes in degrees and σvc friction angle affect both the drained and undrained strengths of soil, since the underlying behaviour is principally frictional. 2.6

As well as the conventional aspects of soil behaviour that are considered in the analysis of foundations, pipeline geotechnics is also often concerned with a greater degree of soil disturbance, and intervening periods of recovery and reconsolidation. These processes also take place at lower stress levels compared to conventional geotechnics, and can lead to significant changes in the state and therefore the strength of a soil through the operating life of a pipeline. Also, due to the relevant drainage distances and rates of movement, it is can be necessary to focus on the drained response of fine-grained soils and the undrained response of coarse-grained soils.

Low effective stress friction

A final feature of soil behaviour that is particularly relevant to pipeline geotechnics is the variation in friction angle with stress level. At the low stresses relevant to pipelines, higher friction angles are found compared to more usual geotechnical stress levels. The peak friction angle of sands increases with reducing stress level (Bolton 1986). Results from experiments performed on Earth (Fannin et al. 2005) and onboard the space shuttle (Sture et al. 1998) also show that the critical state or constant volume friction angle is higher at very low stresses.

3 3.1

PIPELINE EMBEDMENT Pipelaying mechanics

The as-laid embedment of a pipeline affects the subsequent pipe-soil resistance, as well as the thermal

94 © 2011 by Taylor & Francis Group, LLC

Summary of soil behaviour

Figure 12. Pipeline laying notation (Randolph and White 2008a).

Figure 13. Maximum stress concentration factors for pipe laying on an elastic seabed (Randolph and White 2008a).

insulation. The installation process involves soilstructure interaction, since the maximum vertical pipe-soil force during the lay process will exceed the submerged pipe weight, W
, by an amount that depends on the seabed stiffness and the geometry of the catenary created by the S-lay or J-lay arrangement. The configuration of a pipeline during laying is shown in Figure 12. A key parameter is the horizontal component of tension, T0 , which is constant through the suspended part of the pipeline and can be assessed from the pipe weight, water depth and hang-off angle. The maximum contact force (per unit length) with the seabed, Vmax , and hence the local force concentration factor, flay = Vmax /W
, is a function of the seabed stiffness, k (defined as the secant ratio of force per unit length, V, to embedment, w) in addition to the bending rigidity, EI, and T0 . The force concentration factor reduces with increasing water depth and decreasing seabed stiffness. A characteristic length, which relates to the length over which the bending stiffness moderates the catenary behaviour, is given by λ = (EI/T0 )0.5 . Parametric solutions for the static lay conditions have been presented by Randolph and White (2008a), who showed that for horizontal tension of T0 > 3λW
(which holds for most pipelines), results from analytical solutions (Lenci and Callegari 2005) and numerical analysis using OrcaFlex (Orcina 2008) all converge to unique design lines. The value of flay may be expressed approximately as (Figure 13):

Section 3.4, but firstly the significant influence of dynamic pipe movements during the laying process is highlighted. During J-lay or S-lay installation, dynamic movement of the pipe occurs within the touchdown zone, driven by the vessel motion and hydrodynamic loading of the hanging pipe. These loads induce a combination of vertical and horizontal motion of the pipeline at the seabed (Lund 2000, Cathie et al. 2005). In addition to vessel motion due to swell and waves at the sea surface, cyclic changes in pipeline tension may occur (depending on the accuracy of the tensioning system) if the offloading of the pipe is not smoothly coincident with the vessel advancement. This dynamic movement, although often of very small amplitude, leads to local softening of the seabed sediments and can push soil away to either side of the pipe alignment, creating a narrow trench in which the pipe becomes embedded. An illustration of the significant additional embedment that can occur simply due to small amplitude cyclic motions is shown in Figure 14. These results are from a centrifuge model test on lightly overconsolidated kaolin clay (Cheuk and White 2010a). A model pipe was penetrated to a normalised embedment of w/D = 0.1 (point A), when the vertical pipe-soil load was fixed constant (the normalised vertical load, V/su D reduced with pipe embedment, due to the increasing su with depth). A series of packets of horizontal oscillations were then imposed, increasing in amplitude (Figure 14a). The adopted amplitudes of motion reflect ROV observations during laying, although these are dependent on the lay geometry and metocean conditions (Westgate et al. 2010a). The aim in this experiment was to represent dynamic lay motions in an idealised manner. The pipe initially settled at a rapid rate, by an amount that far exceeds that due to the combined vertical and horizontal loading alone. The lateral soil resistance mobilised during the first two cycles, when the embedment doubles (to point B), corresponds to an equivalent friction factor of H/V < 0.25. As the embedment increases, a greater soil resistance is mobilised for a given amplitude of

3.2 Seabed disturbance during pipelaying Equation 1 is derived based on a single value of secant seabed stiffness, V/w, which would be applicable to a purely elastic seabed. The actual seabed response is generally non-linear during vertical penetration, and is stiffer during unloading since the seabed has been plastically deformed. As a result, the actual operative secant stiffness varies along the touchdown zone. Theoretical solutions for monotonic vertical pipe penetration into undrained soil are described in

© 2011 by Taylor & Francis Group, LLC

95

motion, reflecting the increasing constraint on the pipeline. The softening of the surrounding soil is evident in the reductions in lateral resistance at points C and D, with increasing disturbance. Complementary assessments of the level of soil remoulding during pipeline laying can be made using large deformation finite element analysis. A study reported by Wang et al. (2009) replicated the first stage of the model test shown in Figure 14 (with horizontal motions of +/−0.05D) using continuum finite element analysis. The soil constitutive model included softening through a reduction of the undrained strength with accumulated plastic strain, in a manner consistent with the behaviour shown in Figure 5. The resulting patterns of lateral resistance and embedment and the local soil remoulding are shown in Figure 15. Even small lateral motions of just +/−0.05D lead to a remoulded zone that extends by almost one pipe diameter to each side, and the pipe itself rests on fully remoulded soil. These results highlight the importance of assessing pipeline embedment using an appropriately degraded value of soil strength, as well as accounting for the catenary overstress via Equation 1 or some comparable approach. 3.3 As-laid pipeline survey observations Similar conclusions can be drawn from the results of ROV surveys following pipe laying. The variation in embedment along a pipeline varies for a variety of reasons including local variations in the seabed strength. However, consistent variations have also been identified due to wave height, lay rate, downtime events, and changes in lay angle. The first three of these effects influence the level of dynamic motion that the pipe is subjected to in the touchdown zone and the latter effect influences the vertical stress concentration (Westgate et al. 2010a). The resulting range of embedment can be expressed as a statistical variation, and this range can be compared with calculations performed using theoretical solutions for the catenary overstress and the seabed penetration resistance, using both intact and remoulded soil strengths. Figure 16 shows the distribution of embedment from survey measurements taken at 1 m intervals along a 13 km long pipeline (excluding short lengths of pipeline that had a far greater embedment due to downtime events). The pipeline was laid on almost uniform deepwater soil conditions, in a water depth of 1215– 1450 m (Westgate et al. 2010b). The lay process took several days, during which the sea state varied with a significant wave height of between 0.6 m and 1.7 m. The calculated pipeline embedment based on the static catenary overstress (Equation 1) coupled with intact and fully remoulded strengths are highlighted, along with estimates based on a dynamic analysis of the vertical stress concentration performed using Orcaflex (Orcina 2008) (Figure 16). The fully remoulded strength coupled with the static overstress

Figure 14. Effect of lateral motions on pipeline embedment (after Cheuk and White 2010a).

matches well with the most frequent embedment, although the agreement was not as good for some other pipelines at the same site. This result and other comparisons from a limited range of post-installation surveys, suggest that the fully remoulded soil strength leads to reasonable estimates of the average pipe embedment for average lay conditions. However, there remains significant scatter between different pipelines in the same conditions and variations in embedment along a single pipeline (Westgate et al. 2010a, Westgate et al. 2010b). An additional effect appears to be that lighter pipelines are more susceptible to dynamic lay effects – probably due to their reduced inertia – and so a greater degradation in soil strength applies.

96 © 2011 by Taylor & Francis Group, LLC

Figure 15. LDFE simulation of soil strength after dynamic pipe laying (after Wang et al. 2009).

breakout resistance since it is not immediately obvious what combination of input parameters will lead to upper and lower bound outcomes. It is also attractive to set the geotechnical components of pipeline analysis within a full probabilistic framework. This is consistent with the probabilistic structural reliability analyses that are increasingly performed during the assessment of pipeline on-bottom stability or lateral buckling. 3.4 Solutions for vertical pipe penetration Using an appropriate operative soil strength, su , the shallow embedment of a pipeline in undrained conditions can be calculated via a bearing capacity expression that reflects the appropriate geometry of failure mechanism. Experimental results presented by Dingle et al. (2008) revealed the internal soil failure patterns during pipe penetration (Figure 17a). These detailed observations, coupled with the vertical force-displacement response from other tests, have been used to validate numerical simulations (e.g. Figure 17b, Merifield et al. 2008a) and plasticity limit analyses (Randolph and White 2008b). The resulting bearing capacity, Vult , can be calculated as the superposition of components related to the soil strength at the pipe invert (via a bearing factor, Nc , that varies with embedment, pipe roughness and strength heterogeneity) and the soil buoyancy – enhanced by heave (Randolph and White 2008a, Merifield et al. 2009):

Figure 16. Distribution of as-laid embedment for a single pipeline (Westgate et al. 2010b).

The use of the remoulded soil strength over-predicts the embedment in the case of minimal pipeline motions (for example in calm weather or during lay down of the final catenary section of pipe) and under-predicts embedment during severe weather or downtime events, again based on limited field data (Westgate et al. 2009; Westgate et al. 2010b). Due to this inevitable scatter, including variability due to effects that cannot be predicted in advance of the pipelaying, assessments of pipeline embedment are subject to uncertainty. If only an upper or lower bound embedment is required for design then this uncertainty can be circumvented using conservative assumptions. However, both upper and lower bound embedments are usually critical for different design considerations. It is therefore necessary to perform lower and upper bound assessments using opposite extremes of the input parameters. These can be extended to a full probabilistic assessment, using statistical variations in input parameters, as illustrated in Section 4.4. Such an approach is desirable when assessing the subsequent

The nominally embedded cross-sectional area is denoted A
, the soil submerged unit weight is γ
and

97 © 2011 by Taylor & Francis Group, LLC

Figure 19. Undrained failure envelopes for lateral breakout in uniform soil.

4

LATERAL PIPE-SOIL INTERACTION

4.1 Theoretical failure envelopes The same combination of experimental, numerical and analytical techniques has been used to derive failure envelopes in vertical-horizontal load space, which allow the lateral breakout resistance in undrained conditions to be assessed (Dingle et al. 2008; Merifield et al. 2008a, Randolph and White 2008b). Different forms of failure mechanism apply, depending on whether tension can be sustained at the rear of the pipeline (Cheuk et al. 2008; Merifield et al. 2008b) as shown in Figure 19. For the no-tension case (and w/D < 0.5), the failure envelopes pass through the V-H origin, and there is a ‘frictional’ cut-off corresponding to a failure mechanism involving the pipe riding up at an angle shown as θ in Figure 19, with no soil deformation occurring. In drained conditions, experimental results show that the failure envelope has a similar shape to the no-tension undrained case (Zhang et al. 2002a). Centrifuge modelling results from tests on soft clay show that the breakout resistance is well predicted by the unbonded failure envelopes if the pipe is installed with some level of dynamic movement during laying (such as the simulation in Figure 14) (Cheuk and White 2010b). In this case, for very soft clay, the appropriate undrained strength appeared to be the intact value. In this particular case the effects of remoulding and reconsolidation approximately cancel out. Without dynamic movement the breakout resistance is often higher, but is highly brittle, reflecting the opening of a crack behind the pipe prior to mobilisation of a full two-sided mechanism (Dingle et al. 2008). It is thought that any dynamic movement leaves a skin of weak remoulded soil at the pipe surface. This can fail in a local mechanism behind the pipe, even if some tension is sustained, providing negligible extra resistance above the unbonded case. Modified versions of these failure envelopes can be created for other boundary conditions, such as a pipe running parallel to a slope (e.g. Morrow and Bransby

Figure 17. Failure mechanisms during vertical pipe penetration (Dingle et al. 2008, Merifield et al. 2008a).

Figure 18. Bearing factors during vertical penetration into heterogeneous and uniform soil (Chatterjee et al. 2010b).

the factor fb captures the enhancement of soil buoyancy beyond that given by Archimedes’ principle. A value of fb = 1.5 is typical. Solutions for the bearing factor, Nc , in this expression have been provided for uniform and linear soil strength profiles (Aubeny et al. 2005, Merifield et al 2008a), and with modifications to account for heave (Merifield et al. 2009, Chatterjee et al. 2010a, 2010b). A pair of bearing factor profiles derived from large deformation finite element analyses are shown in Figure 18, for two different profiles of soil strength (Chatterjee et al. 2010b).

98 © 2011 by Taylor & Francis Group, LLC

Figure 21. Effect of load path on equivalent breakout friction.

load path reaches the unbonded failure point marked A, with the equivalent friction factor marked µeq . The load path marked B represents hydrodynamic loading, when the pipe is subjected concurrently to both horizontal drag and upwards lift. In this case the equivalent friction factor at failure is higher (although the constant V path marked A provides a conservative assessment). The theoretical failure envelopes of the form shown in Figure 19 can be re-expressed as equivalent friction factors, varying with load path and load level, W
/Vult . For illustration, results have been derived from the unbonded envelopes corresponding to a smooth pipe interface, uniform soil strength and embedments of w/D = 0.25 and 0.45. Load paths varying from dV/dH = 0 to dV/dH =−2 (i.e. a lift force of twice the drag) have been used (Figure 20). This form of presentation clarifies how much the seabed response differs from the simple Coulomb friction that is often assumed. As the pipe weight reduces, the equivalent friction factor rises, particular at low load levels. The effect of load path is shown in Figure 21 which compares µeq for two ratios of dV/dH, using the pure drag case of dV/dH = 0 for normalisation. The load path has a significant influence on the equivalent friction factor at breakout. For the load path with an uplift of dV/dH = −2, the equivalent friction at failure is typically 1.5-2 times higher than pure drag at w/D = 0.25. The difference rises to >2.5 for w/D = 0.45. At low loads, however, the load path does not affect the equivalent breakout friction because the same tangential mechanism applies. This limiting friction factor at low loads that is caused by the tangential mechanism is not captured in conventional ‘friction + passive’ models for breakout resistance. The component of passive resistance in these models is uninfluenced by the vertical load level, leading to a finite breakout resistance under a vertical load of zero. This is incorrect and unconservative for stability design. The tangential failure mechanism leads to zero breakout resistance under zero vertical load for any embedment of w/D < 0.5, if tension cannot be sustained at the pipe-soil interface (regardless of the pipe roughness).

Figure 20. Equivalent friction factors from failure envelopes.

2009). For this case the failure envelope is rotated by the slope angle and changes shape slightly, depending on the ratio of soil strength to weight. Due to the curved shape of the failure envelopes, the equivalent friction factor for lateral failure in a downslope direction is not simply the value for level ground, adjusted by the slope angle according to Coulomb friction (unless failure is via the tangential mechanism). 4.2 Failure envelopes as friction factors Failure envelopes depict the limiting pipe-soil loads in a different manner to the empirical expressions that are commonly used in practice (e.g. DNV 2007a, Verley and Sotberg 1994, Verley and Lund 1995, Bruton et al. 2006). These approaches divide the soil resistance into frictional and passive resistance components that are superposed. It is then common for the resulting resistance to be expressed as an equivalent friction factor, by dividing by the vertical pipe-soil load. In this way, the geotechnical resistance is linked only to the pipe weight (and lift), and any structural analysis software need not include any soil parameters, nor incorporate explicit calculations of the pipe embedment. The theoretical failure envelopes can also be manipulated to express the pipe breakout resistance as an equivalent friction factor, µeq . For a given initial state, the equivalent friction at failure, µeq = H/V depends on the load path in H-V space. The initial state is typically well within the failure envelope (W
/Vult << 1) due to the unloading beyond the maximum overstress of the catenary, dynamic lay effects and also from any gain in soil strength after laying. A general parameter to describe the current level of unloading is the overloading ratio, R = Vult /V, where V is the current vertical load and Vult is the current vertical bearing capacity (Zhang et al. 2002a; Zhang and Erbrich 2005). In lateral buckling analysis, the load path is generally assumed to involve horizontal loading, whilst the vertical load remains constant and equal to the pipe submerged weight (i.e. dV/dH = 0). In Figure 19 this

99 © 2011 by Taylor & Francis Group, LLC

4.3

Table 2.

Coupling of geotechnical analysis and structural and hydrodynamic responses

To properly capture this strong influence of load path in an assessment of pipeline stability it is necessary to couple the failure envelopes with the imposed hydrodynamic loads. As a minimum, the load path of the critical wave condition can be combined with the failure envelopes to derive a single value of equivalent friction factor that is then used in an absolute stability check. The conservative approach of assuming a load path of dV/dH = 0 will provide a safe under-prediction of the available breakout resistance (Figure 19). Adoption of the actual load path when assessing µeq will reduce this conservatism. To provide information beyond simply the breakout resistance, the failure envelopes can be incorporated within a plasticity macroelement model, which allows the full load-displacement response to be simulated (e.g. Schotman and Stork 1987, Zhang et al. 2002b, Cocchetti et al. 2009). In essence, these models allow the pipe embedment and therefore the size and shape of the failure envelope to be updated whenever the load reaches the current failure envelope and displaces. If the load remains within the failure envelope the pipe is stable. The most simple use of this type of model in design is to track the response of a single pipe element whilst it is subjected to the most severe hydrodynamic conditions, unsupported by any neighbouring pipe (which is a conservative simplification). A more sophisticated approach is to attach multiple pipe-soil elements to a full structural model of the pipeline, allowing longitudinal load-shedding to be incorporated. This approach is described by Tian and Cassidy (2008) who incorporated the Zhang et al (2002b) model for drained pipe-soil interaction into a structural analysis of a pipeline subjected to hydrodynamic loading. This allowed full wave-pipe-soil interaction to be simulated using a more sophisticated pipe-soil model than had been previously used (an inevitable limitation being that the wave-seabed interaction depicted in Figure 3 is not captured). 4.4 Quantifying uncertainty in lateral breakout resistance: Monte Carlo analysis As discussed in Section 3.3, the as-laid embedment of a pipeline is strongly influenced by the dynamic effects in the touchdown zone during laying and the consequent reduction in soil strength. These processes depend on the laying procedures and the metocean conditions, which are not known during design. Similarly, the operative soil strength during breakout is subject to uncertainty, partly from natural variability and partly due to potential reconsolidation of the soil near the pipe after lay-induced disturbance. Despite these difficulties it is necessary to provide some assessment of the pipeline embedment and the consequent breakout resistance during design. This assessment can be performed within a probabilistic framework, with the uncertainty in each input

Parameter

Value

Pipeline diameter Submerged pipe weight, W
Specific gravity, SG Touchdown lay factor, klay Soil submerged unit weight, γ
Buoyancy factor, fb

0.5 m 1.5 kN/m 1.38 1.25 8 kN/m3 1.5

Table 3.

Probabilistic parameters for breakout assessment. Statistical values

Operative soil strength profile during laying Operative soil strength profile during breakout

Parameter

P5

P95

Gradient, ksu (kPa/m) Dimensionless gradient, κ = ksu D/sum Mudline value, sum (kPa) Dimensionless gradient, κ = ksu D/sum

1.5

3 20

2

4 0

parameter being given a statistical description. Such an approach is consistent with the reliability-based approaches being used for the structural aspects of pipeline design (e.g. Carr et al. 2004, DNV 2007a, AtkinsBoreas 2008, Rathbone et al. 2008). The recently-developed solutions for pipeline embedment and combined V-H capacity provide a robust theoretical framework within which a probabilistic assessment of the pipeline response can be made.This is illustrated by an example of a 0.5 m diameter smooth pipeline with the deterministic parameters shown in Table 2. The soil strength profiles relevant to the embedment and the breakout events are given a probabilistic description, as set out in Table 3. For simplicity, only the soil strength profiles are treated probabilistically, although it is straightforward to include other parameters that have natural variability, such as the soil unit weight. The adopted values are typical for a flowline being laid on a soft fine-grained soil in deep water. The operative soil strength profile, representing the disturbed state during laying, increases with depth and has a small mudline intercept. The operative value during breakout is uniform and higher. The upper bound (UB) and lower bound (LB) parameters for each strength profile differ by a factor of two and are assigned 5% and 95% exceedence values (i.e. the LB P5 value has a 5% probability of exceeding the actual soil strength). This ratio between UB and LB values is broadly typical of practice, but it is strongly affected by the level of natural variability and the spatial intensity and quality of site investigation data. The probabilistic soil strength parameters might be derived from in situ and ex situ testing on samples distributed along the entire pipeline length, which will then capture scatter related to the natural variability

100 © 2011 by Taylor & Francis Group, LLC

Deterministic parameters for breakout assessment.

Figure 22. Strength distributions in probabilistic analysis case.

(as well as scatter from the tests themselves). It is necessary also to include within these soil strengths some uncertainty to capture the lay process, and any reconsolidation. Having chosen P5 (LB) and P95 (UB) values of soil strength (or any other pair with a specified likelihood, representing upper and lower bounds) a probabilistic distribution must be fitted. In this case a gamma distribution has been chosen because it does not feature negative values (since negative values of strength are not possible). A gamma distribution is also free of the long positive tail that is present in a lognormal distribution (which is sometimes adopted for non-negative variables). The resulting cumulative distribution functions for each soil strength parameter are shown in Figure 22. Using these distributions coupled with the embedment and failure envelope analyses described earlier, a Monte Carlo analysis has been performed using a software program called MCWHIPLASH, developed by Advanced Geomechanics, to illustrate the influence of soil strength uncertainty. The analysis involved 10,000 realisations, which is sufficient to generate smooth distributions of the outputs. The first step of the calculation involves estimation of the as-laid embedment. The probabilistic operative strength leads to the variation in embedment shown in Figure 23. The as-laid embedment varies monotonically with the single variable – the operative strength during laying – in a non-linear manner. Although the P95 and P5 strength parameters differ by a factor of 2, the ratio between the P95 and P5 embedments is only 1.38 (Figure 23b). This reflects the non-linearity of the bearing capacity expression (Equation 2). A halving of the soil strength, for example, does not lead to a doubling of the embedment because the bearing capacity factor rises with depth. Also, the strength of the soil is not the only mechanism supporting the pipeline – the soil buoyancy has an influence and this is not given a probabilistic variation in this example. In practice, the soil unit weight has a narrow range of variability, so the soil buoyancy term generally attenuates the variability arising from the soil strength. Overall, there is a narrowing of the

Figure 23. Variation in as-laid embedment for example probabilistic analysis.

outputs relative to the inputs due to the particular form of the mechanisms that govern pipe embedment. The next step in the analysis is the calculation of breakout resistance, which is expressed here as an equivalent breakout friction factor, Hbrk /V. In this example, the load path to breakout has been assumed to be purely horizontal loading under a maintained vertical load corresponding to the submerged pipe weight (V = W
), which is the usual assumption in a lateral buckling analysis. The breakout resistance is affected by both the aslaid embedment and also the (probabilistic) strength at the time of breakout, which are assumed to be uncorrelated in this example, for simplicity. The variation in breakout friction factor with as-laid embedment is shown in Figure 24a and the variation with soil strength at breakout is shown in Figure 24b. The resulting distribution of friction factor is shown in Figure 24c. The overloading ratio, R = Vult /V, varied between 2.3 and 9.5 at breakout in this example, with a mean value of 4.9. Consequently, none of the Monte Carlo realisations involved failure via the tangential cut-off mechanism (indicated by the cap on Hbrk /V for w/D = 0.25 in Figure 20). A trend line through the centre of the clouds of results in Figure 24a and Figure 24b indicates that the breakout friction factor increases approximately linearly with both soil strength and embedment (although this is not the case for other examples that could have been adopted).

101 © 2011 by Taylor & Francis Group, LLC

Figure 25. Normalised variability of input and output parameters in example probabilistic analysis.

Figure 24. Variation in breakout friction factor for example probabilistic analysis.

For this example, the P5 and P95 values of Hbrk /V are 0.94 and 1.45 respectively, which have a ratio of

© 2011 by Taylor & Francis Group, LLC

102

1.54.As for the embedment, this variability is narrower than the range of input strengths. The cumulative distributions for the strengths and the resulting embedment and breakout friction factor are compared in Figure 25, with each variable being normalised by its P50 value. This form of presentation highlights the relative uncertainty in each parameter. As noted previously, the analysis leads to a narrower output range than input range: a given uncertainty in soil strength leads to a lower uncertainty in breakout friction factor. This is a useful outcome, given the difficulty in assessing accurately the soil strength close to the seabed. This effect can be contrasted with the response of a surface foundation (with or without short skirts). The lateral capacity of the foundation varies in proportion to the surface soil strength, so the soil strength and the foundation capacity have the same level of variability. The deterministic results based on combinations of the P5 (LB) and P95 (UB) values are also marked on Figure 24a and Figure 24b. These results highlight the value of a Monte Carlo analysis. The friction factors found using the extreme values of soil strength in combination (i.e. LB/LB and UB/UB) lie within the centre of the distributions (at the P34 and P55 percentiles respectively). These do not provide a useful indication of the potential variability. The friction factors calculated using the more unlikely assumptions of the UB/LB and LB/UB soil strength combinations give breakout values that have percentiles of P1 and P99 . The (un)likelihood of the results found using these assumptions is also strongly dependent on the particular combination of pipe and soil parameters being used, due to the non-linearity of the governing mechanisms. There is therefore no simple method for linking the probability of the inputs to the probability of the outputs, without a Monte Carlo analysis or a similar statistical technique. If it is assumed that the operative strengths during laying and breakout are correlated, it might be assumed that the LB/LB and UB/UB cases would then provide a true indication of the potential variability in Hbrk /V.

However, this is not the case. In fact, for the example shown here, with fully correlated strengths, the highest value of Hbrk /V is found for an analysis based on the P31 strength during laying and breakout. The maximum breakout resistance does not arise from either the weakest or strongest soil, but an intermediate case. Why does this surprising observation arise? Soft soil gives high embedment, and therefore a high wall of passive soil on breakout. However, the beneficial embedment is countered by the weakness of the passively loaded soil. Due to these counteracting effects coupled with the non-linear shape of the V-H failure envelope (and its non-linear growth with embedment), the extreme values of breakout resistance are often not associated with the extreme values of soil strength. This same interaction underlies the narrower range of breakout friction factors compared to the input range of soil strengths. The form of Monte Carlo analysis shown here provides a solution to this difficulty, capturing the uncertainties associated with soil strength and other parameters that influence pipe embedment and breakout. The simple example shown here includes only the variability associated with soil strength. In practice the other geotechnical inputs such as the soil unit weight can also be treated probabilistically. It is also possible, using MCWHIPLASH, to incorporate other parameters such as the pipeline bottom tension during laying, via the solution given in Equation 1. The uncertainty associated with the geotechnical parameters arises from the site variability and the quantity and quality of site characterisation data. To provide a consistent approach throughout the entire analysis process – from interpretation of the geotechnical SI data through to the pipeline structural analysis – a probabilistic approach can also be used to assess the soil strength parameters (e.g. Lumb 1966, DNV 2007b, Lacasse et al. 2007). However, these methods do require a sufficient quantity of site characterisation data, which is often not the case. The uncertainty related to the changes in soil strength during laying and recovery can be reduced through model tests that simulate these processes (White and Gaudin 2008). 4.5 Large amplitude monotonic lateral response The engineered buckles that are often used to relieve thermal and pressure-induced expansions typically involve lateral movements of several pipe diameters, as is evident in the picture within Table 1. This example is a buckle that was initiated over a sleeper, which provides a vertical upset and therefore reduces the buckling load locally, providing a reliable buckle initiation point. Only the central part of the buckle is lifted away from the seabed, and a significant length of grounded pipeline sweeps back and forth across the seabed as the buckle expands and contracts during operating cycles. Buckles are sometimes initiated on-bottom at lay route curves, and designers always need to consider the likelihood of a ‘rogue’ buckle being initiated anywhere along the route, where the

Figure 26. Deformation mechanisms during large amplitude lateral sweeping of a pipe on soft clay.

pipe rests on the seabed. It is therefore necessary to assess the lateral pipe-soil resistance over long distances of lateral movement across the seabed. Flowlines are typically designed for several hundred startup and shutdown events throughout the field life. During the early years of operation these will be more frequent. In later years, as the operating temperature and pressure reduce, the feed-in to the buckle may reduce. However, if new wells are tied-in, or if the field recovery is enhanced, then the length of feed-in into a buckle, and the consequent amplitude of movement, may increase in later life. As discussed previously, the overloading ratio of a pipeline during operation is typically greater than 2, which puts the load point on the ‘dry’ side of the unbonded breakout failure envelopes (V/Vult < ∼0.45). In this case the pipe moves upwards at failure, leading to a reduction in resistance after breakout as the pipe rises towards the ground surface. Model tests using PIV image analysis have shown that in soft clay the steady failure mechanism involves basal sliding of a berm of soil ahead of the pipe, with a thin layer of additional material being ploughed from the seabed (Figure 26a). The same mechanism has been replicated in large deformation finite element analysis (Figure 26b). The subsequent steady ‘residual’ resistance depends on the size of the berm and

103 © 2011 by Taylor & Francis Group, LLC

the strength of the soil within it (as well as the seabed strength). The size of the berm depends on the initial embedment of the pipe and its trajectory, whilst the strength within the berm depends on the remoulding it has undergone, and the sensitivity of the soil. At a given point in a large amplitude lateral sweep, these two effects can be quantified by defining an ‘effective embedment’, w
/D, which amalgamates the embedment below the original soil surface, w/D, and an additional component of embedment arising from the berm, h
berm /D (Figure 26c). The volume of the berm is calculated from the area swept by the pipe invert. By assuming that the berm has a particular aspect ratio, η = length/height, its height can be found from the area. This height is discounted by a sensitivity, St,berm , which represents the reduction in berm strength due to remoulding, to give the effective additional embedment, h
berm /D:

This normalisation approach provided excellent agreement between a set of centrifuge model tests on soft clay, simulating pipes of different weights and initial embedments (White and Dingle 2010). The tests all showed differing values of steady residual lateral resistance, ranging from Hres /V = 0.32 to 0.65 (Figure 27a, b). There was no clear influence of simulated pipe weight, V, but a scattered trend with initial embedment, (w/D)init . However, the profiles of normalised lateral resistance, H/su D, when plotted against effective embedment, match closely regardless of the vertical load (Figure 27c). All of the responses lie close to a power law relationship between H/su D and w
/D. The same behaviour is seen in large deformation finite element analysis (Wang et al. 2010). This effective embedment approach alone does not provide a predictive tool, only a robust normalisation of the response. A method for calculating the pipe trajectory is required, in order that the evolving size of the soil berm and the local embedment can be assessed. This trajectory is affected by the pipe weight, even though the resistance at a given local embedment appears not to be. This missing link has been filled by project-specific empirical models, and is currently being tackled within the SAFEBUCK JIP, via the development of a plasticity macroelement model. This model will operate in the usual macroelement manner of combining yield envelopes, a hardening law and a flow rule to describe the general force-displacement response of an element of pipe (following Zhang et al. 2002b). However, it will incorporate the influence of the changing soil strength and berm geometry on the shape of the failure envelope. This will allow the macroelement approach to be extended to large amplitude movements. In the meantime, various empirical expressions have been proposed to estimate the residual lateral resistance with reference only to the as-laid state (not the post-breakout trajectory) (e.g. Bruton et al. 2006,

104 © 2011 by Taylor & Francis Group, LLC

Figure 27. Centrifuge modelling observations of residual lateral resistance on soft clay (White and Dingle 2010).

Cardoso and Silviera 2010). Based on the six tests on soft kaolin clay shown in Figure 27, White and Dingle (2010) proposed the a simple relationship between initial embedment, (w/D)init , overloading ratio, R, and residual lateral friction factor, Hres /V:

with suggested parameters of µ = 0.3 and k = 2 (Figure 27d). The form of Equation 4 encapsulates some of the factors that influence the residual resistance: a high (w/D)init leads to a large initial berm size and a low overloading ratio, R, leads to a deeper pipe trajectory; both of which increase the passive horizontal resistance. An obvious limitation of this tentative expression is that it does not include any soil parameters, so does not explicitly capture the varying strength as soil is remoulded ahead of the pipe, or the variation in soil strength with depth. Surprisingly, this empirical expression has a different form to another one derived by a recent confidential study for the SAFEBUCK JIP (White and Cheuk 2009), and also from that published by Cardoso and Silviera (2010). In all cases these expressions are based on a particular collection of model tests, rather than any theoretical mechanism. Such contrasts provide a warning that these empirical expressions may not be as accurate when applied beyond the conditions in which they were calibrated.

Figure 28. Typical full scale modelling observations of cyclic lateral response on soft clay (White and Cheuk 2008).

4.6 Large amplitude cyclic lateral response During large amplitude cyclic lateral movements, the pipe-soil resistance remains dependent on the berm material being swept back and forth. Each time the pipe changes direction the berm is left behind, and is remobilised if the pipe approaches that point during a subsequent cycle. These aspects of behaviour are illustrated by a typical model test involving cycles of large amplitude lateral movement, between fixed displacement limits (Figure 28, White and Cheuk 2008). This test involved cycles of fixed lateral amplitude under constant simulated pipe weight. The general form of the response involves initial breakout of the pipe (point A in Figure 28) followed by a gentle increase in resistance associated with the growth of a small ‘active’ berm ahead of the pipe (B). On reversal of the sweeping direction, this response is repeated (C) and the berm generated during the previous sweep is left behind, becoming ‘dormant’. When the pipe again approaches this point during a later sweep, an increase in resistance is experienced as the dormant berm is collected (D). With repeated cycles of movement, the berms at the limits of the pipe movement grow, causing a corresponding increase in resistance. The first sweep encounters slightly higher steady resistance than later sweeps due to the larger active berm arising from the initial embedment of the

105 © 2011 by Taylor & Francis Group, LLC

Figure 29. Soil berms during large amplitude lateral movement.

pipe. An image of berms created by this type of large amplitude pipe motion are shown in Figure 29a. In the design of lateral buckles, it is important to model the constraint imposed by the soil berms, in order to provide an adequate assessment of fatigue. If the pipe-soil resistance is assumed to be constant,

with the berms ignored, then a buckle will progressively lengthen through cycles of expansion and contraction, which reduces the peak bending stresses near the crown (Cardoso et al. 2006, Bruton et al. 2007). Soil berms inhibit this lengthening, causing the high stresses generated during buckle initiation to be locked-in. Although the restraint provided by the berms also attenuates the amplitude of the pipe motion within each cycle – and therefore the cyclic stresses – the overall effect on fatigue is usually harmful, due to the higher mean stresses that are locked-in (Bruton et al. 2007). Models for the cyclic large-amplitude lateral behaviour can be based on the accumulation and deposition of berm material, as shown schematically in Figure 29b. The current berm size can be used as a hardening parameter that governs the passive resistance, rather like in Equation 4. For undrained conditions, the rate that the berm grows with lateral pipe movement is equal to the depth of soil scraped away by the pipe, from conservation of volume (White and Cheuk 2008). Re-consolidation of the soil that has been remoulded and transported ahead of the pipe will increase the berm resistance. The cyclic resistance is not only affected by the changing geometry. Depending on the soil type, pore pressure dissipation may occur during lateral sweeping, and is also likely to occur between startup and shutdown events. This leads to reconsolidation of the disturbed soil within the berm, and also swelling of the unloaded seabed that is exposed by the scraping action of the pipe. For design, the variation in lateral resistance can be estimated from a geotechnical analysis that considers in detail the mechanisms described above. Then, to incorporate the results into the structural analysis of the pipeline, they can be converted into more simple relationships. For example, the results of a geotechnical analysis can be converted into equivalent friction factors (H/W
) for the residual and berm resistance as a function of cycle number (Bruton et al. 2009). Over the design life of a lateral buckle, the depth of the trench created by the sweeping action of the pipe can be significant. As the trench deepens, soil debris may collect in the base, raising the residual resistance. Soil may even flow over the crown of the pipe. The seabed topography near the crown of an on-bottom buckle that has been operating for several years is shown in Figure 30 (Cardoso and Silviera 2010). The pyramidal berms on each side of the pipe are similar in shape to the model test observations in Figure 29. The pipe itself is almost hidden from view, being partly buried under softer soil that has accumulated within the trench. The situation evident in Figure 30 is analogous to the cyclic rod penetrometer results shown in Figure 8. Repeated remoulding and reconsolidation of the seabed will tend to increase the soil strength and consequently increase the constraint on the pipe. The accumulation of pipe embedment during a set of full-scale and centrifuge model tests is shown

Figure 30. Topography of a soft clay seabed after several years of operation of a lateral buckle (Cardoso and Silviera 2010).

Figure 31. Accumulation of pipeline embedment during cyclic large-amplitude sweeping: SAFEBUCK model test database (White and Cheuk 2009).

in Figure 31. These results are from a database of large-amplitude cyclic lateral tests collated by the SAFEBUCK JIP. These tests involved soft finegrained model seabeds. Short elements of pipe were modelled, exerting a constant bearing pressure in the range V/D = 0.5– 8 kPa. Fixed-amplitude lateral cycles of 1–10 pipe diameters were imposed. The pipes embedded at rates

106 © 2011 by Taylor & Francis Group, LLC

of 1–50% of the pipe diameter per cycle and there was no tendency for a stable embedment to be reached. It is worth noting that these tests involved up to only 60 cycles of movement, whereas most pipeline systems are designed for an order of magnitude more startup and shutdown events. An effect not accounted for in these tests was the longitudinal flexural stiffness of the pipe. In all tests the simulated pipe weight was held constant. In practice, any vertical movement of the pipe down into the trench can lead to a reduction in the vertical pipe-soil contact force, with load being transferred longitudinally to adjacent sections of pipe. This in turn will attenuate the rate at which the pipe embeds. This soil-structure interaction is difficult to capture in design without a pipe-soil interaction model that simulates the trajectory of the pipe. The macroelement plasticity models (e.g. Zhang et al. 2002b; Tian and Cassidy 2008) have this capability, but have not yet been extended to capture large deformation effects related to soil berms and remoulding. In cases where the pipe settlement has been considered important to capture, some recent projects have used a cycle-by-cycle approach in which the trench geometry is updated after each startup and shutdown, based on a geotechnical analysis of the pipe trajectory that is performed independent of the structural analysis (Bruton et al. 2009). This approach allows the changing vertical pipe-soil contact force to be captured through a structural analysis of many operating cycles, albeit in a manner that requires significant user intervention. Given the settlement implied by extrapolation of Figure 31 to the typical design number of operating cycles, it is likely that some longitudinal load shedding will occur during the design life of most on-bottom lateral buckles. Centrifuge and large-scale model testing is currently a widely-used technique to aid the assessment of cyclic large-amplitude lateral pipe-soil resistance (White and Gaudin 2008; Langford et al. 2007). In the long term, it is envisaged that analysis techniques for this behaviour will become more well-established, and model testing will not be required as often. However, given the complexity of the mechanisms shown here, and the difficulty of predicting the behaviour based on conventional soil parameters alone, it is currently common to perform project-specific model tests to assess appropriate cyclic pipe-soil model parameters. 5 AXIAL PIPE-SOIL INTERACTION 5.1 Effective stress and total stress models The axial resistance between an on-bottom pipeline and the seabed affects the feed-in response towards lateral buckles and also controls the end expansions. In addition, it influences the ‘walking’ behaviour through cycles of startup and shutdown (Carr et al. 2006). Low axial friction can be particularly problematic, and the relationship between pipeline walking rate and axial pipe-soil resistance is generally non-linear. The rate of

walking per cycle can increase rapidly as the available axial resistance decreases (Bruton et al. 2007). Compared to the lateral behaviour, the axial pipesoil response involves a more simple geometry, with failure being constrained to occur at, or close to, the pipe-soil surface. By analogy with pile design, both α (total stress) and β (effective stress) approaches can be considered. The β-method expression for the ultimate axial resistance per unit length, T, is:

where µ is the pipe-soil friction coefficient, which can be alternatively expressed in terms of a pipe-soil friction angle, δ, where µ = tan δ. The parameter ζ is a factor that accounts for the enhancement of the normal pipe-soil contact force due to a ‘wedging’ action (White and Randolph 2007). The α-method expression is

where αsu is the shear stress acting on the pipe surface at failure and DθD
is the pipe-soil contact length around the pipe perimeter, with θD
being the angle subtended by radii to the limits of the pipe-soil contact. The adhesion factor, α, captures the roughness of the surface (i.e. the relative strength of pipe-soil and soil-soil shearing). Also, if su is taken as the in situ soil strength then α also encompasses any differences between that strength and the strength of the soil at the pipe surface, due to downdrag, and the processes of remoulding and reconsolidation since the pipe was laid. The argument in support of β-methods is rooted – perhaps idealistically – in the fundamental concept that soil strength is controlled by effective stress friction. In addition, pipeline movements can often be so slow that drained conditions prevail even in fine-grained soils. However, excess pore pressures are generated at even modest speeds in some clays. For this situation, the β-method can be adjusted by using an excess pore pressure ratio, ru = u/σn,av , based on the average excess pore pressure (u) and total stress (σn,av ) around the pipe surface: Equation 5 is then multiplied by (1 − ru ). If the response is fully undrained it is instead tempting to use an analysis based on the relevant undrained shear strength. Estimation of the reconsolidated su beneath the pipe might be thought easier than assessing ru , although the two types of calculation are essentially interchangeable and based on the same principles. As a pipe penetrates the seabed, the surficial soil is dragged downwards. As a consequence, the pipe is bearing not on soil from below the surface, but on soil that was previously at or very close to the surface – as illustrated by the experimental data (Dingle et al. 2008) and numerical results (Zhou et al. 2008) shown in Figure 32. Also, the excess pore pressure generated during the undrained penetration process will take time to dissipate. Solutions for the dissipation of excess pore pressure around a pipe based on an elastic soil

107 © 2011 by Taylor & Francis Group, LLC

Figure 32. Seabed distortion after vertical pipe penetration (Dingle et al. 2008 (left) vs. Zhou et al. 2008 (right)).

Figure 34. Interface shearing: kaolin clay and Storæbelt clay till.

Figure 33. Post-laying pore pressure dissipation around pipe perimeter (Gourvenec and White 2010).

model are presented by Krost et al. (2010) and Gourvenec and White (2010). Figure 33 shows the variation in normalised excess pore pressure, averaged around pipe, with dimensionless time. These dissipation curves allow the level of consolidation at the pipe-soil interface prior to any in-service movements to be assessed – which is analogous to the ‘set-up’ of a driven pile. The pipe weight often applies a bearing pressure that exceeds the apparent pre-consolidation pressure of the seabed, so the soil strength will generally increase during dissipation of the excess pore pressure created by the laying of the pipe. A final complication is that if the response is undrained, then the state of the soil beneath the pipe will inevitably change through the hundreds of operating cycles, and therefore hundreds of episodes of failure followed by reconsolidation. The tendency for excess pore pressure to be generated during axial movement will be reduced, as described below. 5.2

Interface shearing and drainage

The response of a fine-grained soil during interface shear varies between fully-drained and fully undrained across a range of velocities that typically spans two or three orders of magnitude. It is useful to consider firstly the response in tests where the apparatus forces the failure to occur on a particular plane, rather than the case of a curved pipe surface on the seabed. The interface direct shear box tests from Figure 10 are shown again in Figure 34, and supplemented by results from shearing of a concrete – clay till interface at varying rates, reported by Steenfelt (1993). This study

related to the sliding capacity of shallow foundations supporting the Storebælt Crossing in Denmark, but the results provide useful insights relevant to axial pipe-soil interaction. The responses of both the kaolin clay and also the Storebælt clay till are very similar, in terms of both the limiting drained and undrained resistances and also the velocity range over which this transition occurs. A hyperbolic relationship provides a convenient fit to the response. However, these results are not generally applicable to all soils. Other proprietary tests show the drained-undrained transition at a velocity that is two orders of magnitude different, reflecting the particular characteristics of the soil. It is not immediately obvious how to normalise the drained-undrained transition for shearing on a plane. The soil permeability, k, is a relevant parameter, since it controls the rate of pore water flow and excess pore pressure dissipation, but it can be argued that the stiffness may not be: the volume of pore water to be expelled may depend principally on the shear zone thickness (and its potential to dilate or contract), rather than the bulk stiffness of the soil. This argument is pursued by Palmer (1999) in his analysis of ploughing and is discussed in Section 6. If the soil stiffness is not relevant then the coefficient of consolidation is not required within the normalisation, unlike the cases shown in Figure 9. It remains to be established whether the appropriate dimensionless group to normalise the velocities in Figure 34 is vL/cv (with L being some relevant drainage distance) or v/k. In practice, it can be more efficient to perform multiple interface shear tests at varying velocities (and at the low stress relevant to pipelines) in order to identify this response directly, rather than assessing it indirectly from a normalised theoretical curve via measurements of cv and k. The form of response shown in Figure 34 can be used to assess the relevant axial resistance for the initial axial movement of a pipeline. Over a longer period, and after a series of movements, this relationship no longer applies. The apparent friction coefficient during undrained and partially drained sliding is lower than the drained case due to the generation of positive

108 © 2011 by Taylor & Francis Group, LLC

pore pressure when shearing is initiated. Subsequent dissipation of this pore pressure leads to a rise in effective stress. This is accompanied by contraction of the interface zone, which can be observed as settlement of the platen in a direct shear test. An example of this behaviour is shown in Figure 35. These are results from shearing of normally consolidated kaolin clay on a steel interface.The shearing rate was close to the fully undrained limit and a sequence of 2 cycles of +/−5 mm is shown. No consolidation period was permitted between cycles. The initial apparent friction (based on total stresses) is τ/σn = 0.28. However, over the following two cycles the resistance rises (Figure 35a) whilst the sample settles (Figure 35b). The rate of settlement is approximately constant through the two cycles, with a slight tendency for more rapid settlement following each reversal. Meanwhile, the absolute shearing resistance, when plotted against the cumulative horizontal movement, shows a continuous rise, interspersed with reductions to zero at each reversal point (Figure 35c). After only 35 mm of movement, or an elapsed time of approximately 3 hours, the apparent friction coefficient has risen from 0.28 to 0.46 (Figure 35d). This response can be illustrated schematically in stress – specific volume space (Figure 36a). For the type of test shown in Figure 35, the stress path initially heads to the critical state line in an undrained manner (path OU). If sliding continues, then the state moves along the critical state line (CSL) following the increase in effective stress that accompanies the dissipation of excess pore pressure (UP). Ultimately the test path and the strength (or apparent friction) reaches the same state as a drained test would (point D) (Figure 36b). The drained test follows the path OD. At point D the effective stress equals the applied total stress. The idealised strength response in Figure 36b ignores the possible influence of a changing mobilised friction angle, and we are also not differentiating between a critical state line approached after monotonic shearing, and one at a denser state approached through cyclic shearing. 5.3 Episodic interface shearing and consolidation The three test paths shown in Figure 36 cover three possibilities for an initial episode of sliding. For the fully undrained case, after sliding halts, the consolidation will be accompanied by contraction following an unload-reload (κ) line (U-R, Figure 36a) – exactly analogous to the reconsolidation stages shown in Figure 7. Following this reconsolidation, a smaller excess pore pressure and hence a higher undrained strength will be mobilised in a subsequent fast sliding event – just as shown in Figure 6. These idealised responses of a single soil element through episodes of undrained sliding and reconsolidation can be used to estimate the long-term cyclic axial pipe-soil resistance during movements at undrained rates. Eventually, the process of undrained shearing

109 © 2011 by Taylor & Francis Group, LLC

Figure 35. Settlement and apparent friction behaviour during interface shearing of kaolin clay.

followed by reconsolidation will take the soil state to the point on the critical state line at the applied total normal stress. In this case there is no longer a tendency for excess pore pressure to be created during shearing, and the apparent friction will be the drained value regardless of the shearing rate.

Figure 36. Critical state interpretation of drained, undrained an partially-drained interface shear tests.

5.4 Application to axial pipeline resistance These observations from shear box testing are not directly applicable to pipeline behaviour due to the potential for failure to occur on multiple planes at different radii from the pipe surface. At first sight, Figure 35 and Figure 36 suggest that once a drained failure condition has been reached then the apparent sliding friction will always be the drained value irrespective of the rate of movement, since the soil state at the interface has reached the critical state line, and has no tendency to contract or dilate. However, in a shear box test the failure plane is prescribed by the boundary conditions of the apparatus, so only this single plane needs to be hardened. In contrast, failure can occur at different radii close to the surface of a pipeline. If the pipe surface is rough then both pipe-soil shearing and soil-soil shearing have comparable resistance, and failure will occur at whichever circumferential surface offers the least resistance. We illustrate how this affects the hardening behaviour by considering a pipe that undergoes intermittent fully undrained movements, with intervening consolidation periods. For simplicity we are setting aside the effect of partial drainage during shearing. The shear stresses created within the soil by the axial traction at the pipe-soil interface decay approximately with the inverse of the radial position. Given that a shear plane is typically a fraction of a millimetre in thickness in fine-grained soils, there are many potential shear planes carrying essentially the same shear

© 2011 by Taylor & Francis Group, LLC

Figure 37. Schematic illustration of successive undrained failures on adjacent surfaces during axial pipeline movement.

stress. If the soil is assumed to be uniform, then an initial axial movement, fast enough to be undrained, will cause failure within the shear zone immediately adjacent to the pipe (zone A in Figure 37a, b). The mobilised axial resistance will be controlled by the effective stress during undrained failure of this element, which will follow the stress path OA in Figure 38. After a subsequent period of pore pressure dissipation, shear zone A will have contracted (Figure 37c) and hardened to point A
, under the stress of σn imposed by the pipe (Figure 38). However, although this element is now able to offer increased shearing resistance during a subsequent cycle, zone B has not been hardened, and will form the ‘weak link’ during the next undrained axial movement (Figure 37d). This process can continue with each successive cycle leading to hardening of another zone of soil – or, as the shear zones progressively harden, to the remobilisation of a previously-sheared zone. Ultimately all zones will reach the steady state (SS) point on the CSL

110

interpret them within the framework described here, to support assessments of axial pipe-soil resistance. 6 TRENCH CONSTRUCTION BY PLOUGHING 6.1

Figure 38. Critical state interpretation of failures on adjacent planes during axial pipeline movement (see Figure 37).

(Figure 38). Hereafter, the mobilised soil strength and hence the axial pipe-soil resistance will be the drained value, regardless of the rate of movement. However, this process of hardening the multiple potential failure zones means that a considerably greater number of cycles and episodes of consolidation are required, compared to in a shear box test. Further research involving model testing, supported by episodic interface shear tests, will shed light on these mechanisms. What we have presented here is clearly idealised, but it captures the most important features of cyclic axial pipe-soil behaviour in finegrained soils, namely: 1. The significant difference between the undrained and drained sliding resistance of ‘virgin’interfaces; 2. The potential for partial drainage within the shear zone during a long slow sliding event; 3. The consolidation and hardening that follows undrained sliding episodes in soft soils (on the wet side of the CSL); 4. The effect of the multiple potential failure surfaces at, or close to, a pipe-soil interface. The critical state framework provides a convenient basis for interpreting and analysing these events. Ultimately, given sufficient cycles of movement and reconsolidation, any shear box or pipe-soil interface should reach a state where there is no further tendency for the shear zone to contract and excess pore pressure to be generated. At this point a sliding resistance equal to the high drained value would be reliably achieved for all rates of movement. However, it remains unclear how many cycles (or what cumulative distance of shearing) is required to reach this state in a given soil. The form of interface shear box test used to illustrate this behaviour in an individual soil element is rarely conducted on fine-grained soils, precisely because of the ill-defined drainage boundary conditions, and the possibility of partially-drained behaviour occurring. In most normal geotechnical applications such behaviour would be unwanted. However, here it illustrates the influence on interface friction of partial drainage from a shear band concurrent with sliding. Given the insights shown here, it can be worthwhile to conduct project-specific interface shear tests and

Pipeline ploughs are used to excavate soil from a trench to permit a pipeline to be lowered below the seabed (Palmer et al. 1979; Reece and Grinsted, 1986; Cathie and Wintgens, 2001) and for burying cables (Jordon and Cathie, 2004). The soil mechanics of ploughing involves large soil deformations, mobilization of peak and residual shear strengths simultaneously in different zones around the share, partially drained failure, strain rate effects on soil strength, and interface shearing. In this section, some of these themes are developed with knowledge gained recently. Plough-soil mechanics is complex and field soil conditions are rarely uniform.The operational environment towing a plough with a long catenary is analogous to a stick-slip resistance on the end of a large spring hanging from a heaving vessel. Trying to keep the plough on track is difficult, let alone making clear field observations and measurements of performance. These conditions can lead trenching contractors to be satisfied with approximate correlations of plough performance in terms of trench depth, tow force applied and progress rate. However, this is unsatisfactory, at least scientifically, and it is certain that improvements can be made to the current public domain methods for pipeline plough assessment, which are summarised by Cathie and Wintgens (2001). A large pipeline plough and the key loads acting on it are shown in Figure 39. The plough advances due to a tow force (T) being applied which overcomes the soil resistance acting on the engaged parts. In normal situations these are the base of the skid and the base and upper surface of the share. Figure 40 is a close-up of a typical share of a pipeline plough for producing a V-shaped trench. Referring to Figure 39, in steady-state ploughing a force and moment balance exists between all the forces acting on the plough. In actual operations the line of action of the base share forces (Shv , Shh ) in particular is continuously changing. The line of action moves towards the tip as the plough pitches forward – for example if the operators raise the skids by rotating the skid linkage counter clockwise in Figure 39 – and towards the rear if the reverse happens. Forward or aft pitch also occurs if the downward component of the soil resistance on the top of the share changes, or if the normal component of soil resistance on the base of the share changes (e.g. if harder or softer soils are encountered). Ploughs appear to move forward in a series of accelerations and decelerations, or start-stops. This is due to the tow wire catenary behaviour, and the time dependent soil shear strength mechanisms (brittle failure, partial drainage, and thixotropy or setup).

111 © 2011 by Taylor & Francis Group, LLC

Mechanics of ploughing

Figure 39. General arrangement of a large pipeline plough showing actions and resistances.

Figure 40. Pipeline plough share (mouldboards that displace the excavated soil to the side of the trench are retracted).

The equations of motion and the catenary response can be coupled with a soil resistance model quite easily in a finite difference algorithm but the unknowns are largely associated with the soil resistance models and the simulation is of academic value except in very deep water where the catenary is particularly long and flexible.

6.2

Existing ploughing resistance model for sand

A ploughing resistance model for sand was described by Cathie and Wintgens (2001) based on some basic theory and calibrated to a limited data set of full size ploughing:

where F is the horizontal component of the ploughing resistance, W is the plough weight (increased by the roller loads i.e. the weight of pipe supported), Cw is a friction coefficient, γ
is the soil submerged unit weight, and Cs is a coefficient, similar to a passive pressure coefficient. The first term is, of course, the frictional resistance of the plough and is assumed to be independent of speed. The dynamic resistance component of Equation 7 (Fdynamic = Cd vD2 ) principally arises from the potential of the soil to dilate during shear, and the change in pore pressure (normally suction) and effective stress that this induces, offset by any drainage that may occur. Palmer (1999) has shown that, for a simplified representation of a triangular plough share, the velocity-dependent component of force, Fdynamic , is a function of the forward velocity, v, the soil dilation potential, S and the soil permeability, k:

The function f is linear until pore water cavitation begins to occur in some area around the share. Similar relationships have been developed by van Os and van Leussen (1987), and van Rhee and Steeghs (1991). The former validated the theory with plane strain tests in 2D while the latter performed tests with a plough geometry (i.e. a triangular trench) and performed multipasses to simulate several plough shares at different depth settings. The soil dilation potential, S, or volumetric strain to critical state (Van Leussen and Nieuwenhuis, 1984) may be defined as

112 © 2011 by Taylor & Francis Group, LLC

where ec is the critical void ratio and e is the in situ void ratio. Considering that the value of S can only vary up to about 0.4, while the permeability k can vary by at least an order of magnitude in sands of broadly similar grain size, it is clear that the dynamic resistance to ploughing will be dominated by the soil permeability and variations in this parameter. It is likely that much of the scatter that is observed in progress rates when ploughing in sand is due to small local variations in grain size distribution and therefore permeability. Therefore, Cd , the dynamic force coefficient, will be a function of (S/k), increasing gently with density and increasing strongly with reducing permeability. The term in D3 in Equation 8 arises from a term in D for the magnitude of the suction pore pressure and D2 arising from the cross-sectional area of the trench to be cut. This ignores the actual three-dimensional shape of the share and takes no account of the actual deformation mechanism experienced by the soil around the share. The share can be considered as a horizontally advancing ‘penetrometer’increasing in size up to a certain point. When the share tip first influences the soil, the cavity formed may be considered “deep”, i.e. unaffected by the presence of the soil surface. At this point, the suction pressures will be related to the dimensions of the tip and not the depth below seabed. As the share advances the cavity increases in size and breaks out on the seabed so a depth effect (“shallow” failure) is correct. Considerable forward movement (and therefore time – several seconds) is associated with transitioning from the “deep” to the “shallow” failure at any cross-section, during which time drainage can occur. These 3D aspects of the share geometry are one reason why Cathie and Wintgens considered that the dynamic resistance actually increases with something nearer to D2 (Equation 7) rather than D3 (Equation 8). 6.3 Components of ploughing resistance in sand Further insights into the ploughing resistance formulation are provided by more recent research discussed below. 6.3.1 Static ploughing resistance Model testing in dry sand at 1g performed by Lauder (2010) has provided some useful understanding of drained ploughing resistance and the soil flow mechanism around the share and mouldboards (Figure 41). There can be a large accumulation of spoil above the share. This adds weight to the share (thus increasing share friction), adds weight as a surcharge to the soil being sheared by the share, and adds resistance due to the pressure of the spoil on the mouldboards. The flow mechanisms seen in model testing are confirmed by the pattern of scour marks on full-size plough mouldboards. Figure 42 provides a front view of ploughing showing schematically the area of undisturbed soil in the trench, the disturbed soil (for sand,

Figure 41. Model plough showing accumulation of soil above the share (Lauder, 2010).

this material is likely to be close to the critical state) and the spoil heaps. The effect of any build-up of spoil in the body of the plough was taken account of implicitly in calibrating the lumped Cs and Cw terms of the model of Cathie and Wintgens (2001) (Equation 7). However, one could argue that the drained resistance is made up of several separate components: 1. Share base friction or adhesion due to the downward component of soil resistance on the share face; 2. Shearing resistance of the undisturbed soil ahead of the share (passive pressure) augmented by the weight of the surcharge; 3. Shearing resistance of the disturbed soil acting on the mouldboards. The passive resistance component of Equation 7 (Fpassive = Cs γ
D3 ) is the basic component of share resistance and the vertical component also creates additional friction on the base of a share (leading to the first term in Equation 7). This can be assessed approximately using 3D finite element analysis. Results are presented in Figure 43 for a range of peak friction angles representing loose to very dense sand. A Drucker-Prager soil model with a cap was used in ABAQUS. It can be seen that the passive resistance does not increase with D3 and is more correlated with D2 or the trench cross-sectional area, A. The effect of surcharge is not included in this analysis and the soilsteel friction angle was taken as φ – 5◦ which may be rather high for the denser soils. Peng and Bransby (2010) performed 2D FEA modeling and concluded that the static resistance parameter Cs was linearly related to Kp tanφ or K2p for the 2D model, varying with depth as D2 . This study did not include the effect of surcharge. Bransby et al (2005) and Lauder et al (2008) report 1 g model test results that appear to broadly confirm the static resistances suggested by Cathie and Wintgens (2001) but with a margin of uncertainty due to the lower effective stress level and therefore possibly higher operational friction angle in the 1g tests. The formulation of the drained ploughing resistance therefore still needs to be resolved finally. The dimensionless parameter group would be expected to be Fpassive /γ
D3 but this does not appear to be the case.

113 © 2011 by Taylor & Francis Group, LLC

Figure 42. View of soil moving operation with a pipeline plough.

Figure 43. Drained resistance of share only as a function of depth and angle of friction.

The effect of surcharge in a two-dimensional idealisation has been considered by Hettiaratchi and Reece (1974, 1975) and is a function of surcharge pressure (q) and trench depth (D). For a 3D diamond shaped share, the surcharge pressure is (as a first approximation) a function of the trench cross-sectional area divided by the trench width (see Figure 42). Therefore, the surcharge pressure is a function of depth (q = f(γ’D)) and the effect on soil passive resistance a function of D2 . Surcharge pressure also increases the vertical load on the share and therefore increases friction. This component would be related to the volume of soil retained above the share and so both D2 and the length of the share, i.e. D3 for a diamond-shaped share. In respect of the component of resistance arising from mouldboard pressure, this should be related to the surcharge depth as discussed above, and therefore the resistance would be a function of D2 . Finally, work must be done to lift the soil from the trench to the spoil heap level. For an increment of forward movement, the volume of soil to be raised is a function of trench area. Assuming that the average distance it must be raised is related to D the contribution to resistance is unsurprisingly related to γ
D3 . Clearly, a combination of these terms will control experimental or field data. It is essential to get an improved understanding of these components of

Figure 44. Suction pressures during soil cutting (He et al. 2005) (low and negative to hydrostatic: blue, green, yellow, red).

static resistance and their relative magnitudes before measured data can be properly interpreted to establish the dynamic resistance components. 6.3.2 Dynamic ploughing resistance in sand As noted in Section 6.2, rate-dependent effects in submerged sands are largely due to the suction created by dilation potential during undrained shearing. This subject has been studied extensively in respect of cutting sands during dredging (Van Leussen and Nieuwenhuis, 1984; van Os and van Leussen,1987; Miedema, 1987; Miedema, 2005). Experimental and numerical analysis (e.g. van Os and van Leussen, 1987; He et al, 2005) demonstrates that the normalized cutting speed v/k is a key variable controlling this resistance (Figure 44). All of the research into the cutting of sand for dredging, which is centred on groups based in Delft, has used v/k for normalisation rather than vD/cv , which is more commonly considered for normalizing penetrometer and foundation speed effects in geotechnical engineering. The answer may lie in the localisation during shearing of sand, limiting the dilation to a narrow plane. The resulting pore pressure response is essentially a seepage problem, rather than one of consolidation. Interestingly, there are exceptions within

114 © 2011 by Taylor & Francis Group, LLC

the interface properties: sud = f(S,v/k, φ, δ). Dynamic resistance would then be related to D2 considering the effect of this operative undrained strength on the passive resistance (Hettiaratchi and Reece, 1974) as suggested by Cathie and Wintgens (2001). It is possible that a large component of the dynamic resistance arises from the increased normal stresses on the base of the share due to suction. The use of an operative undrained strength would facilitate an integrated interpretation for ploughing in sands and clays and enable the resistance to be evaluated in terms of a normalised ploughing speed (which would be consistent with the framework shown in Figure 9, which is used elsewhere in pipeline geotechnics). 6.4 Figure 45. Pore water pressures around share during ploughing (low and negative to hydrostatic: black, blue, green, yellow, red; the upper figure is at the share tip, the lower figure is further back along share).

geotechnical engineering: Elsworth and Lee (2005) analyse excess pore pressure generation during cone penetration using only v/k, treating the problem as seepage not consolidation. Cathie Associates has performed 3D finite element modeling of the plough share problem to investigate the dynamic component of ploughing resistance. A Mohr-Coulomb cap soil model with limited dilation was used within ABAQUS. Differing from the He et al. (2005) work where cutting rates are faster, our FEA suggests that maximum suction pressures are relatively localized around the share, although of course the zone would grow for higher speeds or lower permeability. Ahead of the share, the suction pressures are lower due to the increased mean stress resulting from the advancing share. Figure 45 also shows clearly the effect of the free surface higher up the share where suction pressures drain much more quickly. Partial dissipation of excess pore water pressures is limited near the tip but quite substantial further back along the share, particularly near the top. The length of the share is a factor affecting the dynamic ploughing resistance just as it is for static resistance. A further conclusion arising from the 3D FEA work was that the average plastic volumetric strain around the share (a measure of the dilation potential) was not very sensitive to the soil density (or dilation angle) even if very localized high dilations occurred. Finally, the dynamic ploughing resistance also contains a contribution from the dynamic shearing of the accumulated spoil against the mouldboards. In the view of the authors, a promising approach to assess dynamic resistance is to consider a operative undrained strength, sud , that would be a function of the average pore pressures generated around the share (capped in dilatant conditions by cavitation) and

115 © 2011 by Taylor & Francis Group, LLC

Ploughing resistance in clay

Cathie and Wintgens (2001) also proposed an empirical model for ploughing resistance in clays:

where Fw is the adhesion of the underside of the skids and share to the soil during ploughing, Cc is a coefficient similar to a bearing capacity factor, su is the soil undrained shear strength, D is the trench depth, v is the plough speed and Cd is a coefficient relating the strength of the soil at normal shear strain testing rates (su ) to the strength of the soil at ploughing rates of strain. Recent numerical studies by Cathie Associates have highlighted the sensitivity of Cc to the working interface resistance between the share and the soil. Near the tip the resistance is probably close to the intact su . Further along the share the strength almost certainly drops to a residual value, possibly corresponding to the remoulded shear strength. The working interface resistance has a strong impact on the values of Cc computed. Probably insufficient attention has been given to experience and research on tillage tools in the agricultural sector. For example, Karmakar (2005) investigated the behaviour of a tool shown in Figure 46 which has some similarities to a pipeline plough share. He found that brittle failure dominated the response of the soil above the share. While his work was in unsaturated materials, ploughing experience confirms that, in stiff clays, the soil is excavated as lumps and our FEA work also demonstrated the importance of brittle tensile failure for all but soft clays. The value of Cc indicated in Cathie and Wintgens (2001) did not take account of brittle failure and considered that it was largely independent of su . This appears to result in over-predictions of the ploughing resistance in very stiff to hard clays. We assume this is because the clays actually shear in a brittle manner with the operational strength being somewhere between the intact, residual or (submerged) tensile strengths. While for clays there is still plenty of scope for theoretical work on ploughing resistance, it receives

Figure 46. Crack propagation in brittle unsaturated clays (Karmakar, 2005).

Figure 48. Examples of jet trenchers. Figure 47. Schematic of pipeline lowering using jet trenching (Vanden Berghe et al. 2008).

less attention in industry because ploughing most clays is much easier than ploughing sands.

7 TRENCH CONSTRUCTION BY JETTING 7.1

Mechanics of jet trenching

7.2

1. Sediment entrainment (erosion and fluidization of coarse-grained soils) or soil cutting (fine-grained soils); 2. Sediment transport; 3. Trench collapse; 4. Sediment deposition.

Jet trenching involves the use of water jets to erode and fluidise sands, or to cut clays, in order to permit lowering and burial of cables (Jordon and Cathie, 2004) and pipelines as shown in Figure 47 (Vanden Berghe et al, 2008). Jetting systems utilize the power of water jets to erode or cut the soil, and to transport or fluidise it to allow the cable or pipeline to be lowered into the trench (Figure 47). Jetting systems range in power from about 250 kW to 2 MW and jet swords generally comprise quite large numbers of nozzles facing forwards, downwards and rearwards depending on the manufacturer’s design. Jetting pressures are typically in the range 2–10 bar with flow rates between 100–3000 m3 /hr.

116 © 2011 by Taylor & Francis Group, LLC

Jetting in sands

The main physical processes involved in jetting of coarse-grained soils are:

The water jets erode coarse-grained soil due to their high energy and carry the suspended particles away from the forward face, transporting them rearward. Sediment erosion is a well understood and for the typical flow rates used by jet trenchers, erosion and fluidization is normally achieved. However, if there is insufficient flow compared to the trench volume and progress rate of the trencher, the fluidization will be incomplete and the jetting swords will experience resistance from the soil still in place. A nominal measure of the soil volume to be excavated is defined by the excavation rate ER :

Figure 49. Relationship between fluidization ratio and trencher speed.

where v is the trencher progress rate (m/hour), DSw is the embedded sword depth and WSw is the sword width. The ratio of the water pumped into the soil and the volume to be excavated at a given speed can be called the fluidization ratio (i.e. jetting system flow rate/ excavation rate). This is also a measure of the resulting water/solids ratio (or the solids concentration, as is considered in the materials transport industry). It is useful for any trencher to develop charts of fluidization ratio for various speeds, sword width and sword depth to provide a rapid method of assessing likely progress rates that could be anticipated. An example is shown Figure 49. Fluidisation ratio is one indicator of the likely maximum rate of progress of a jet trencher and must be sufficient to enable the soil to be transported at the flow rates available. Fluidisation ratios above 5 appear to be sufficient but this also depends on the soil grain size and the configuration of the jets. Following erosion and fluidization of the sand grains, the sand must be transported behind the trencher before it is deposited in the lower flow regime behind the swords. The process is depicted in Figure 50 based on experimental work at the University of Taiwan on behalf of CTC Marine Projects. Both the power and orientation of the forward jets, and the power of any rear facing jets affects the sediment transport and to some extent defines the length of the turbulent region through which the pipeline or cable is lowered. Lateral inflow of soil is also observed in model tests, which merges with the turbulent flow behind the swords to increase the solids content and reduce the energy in the turbulent region. Inflow is less in fine, dense sands which can support a near vertical wall for longer than in coarse, loose sands. Eventually, sediment is deposited in a hindered settling regime at a rate that is largely dependent on the mean grain size. Fine sands settle much more slowly

than coarse sands and gravels. Therefore, the fluidized zone is longer and burial often better in fine soils (see Figure 50 and Figure 51). A mathematical model of the jet trenching process has been developed by Vanden Berghe et al (2008) and Peng and Capart (2008) under contract to CTC Marine Projects. The model is based on the fundamental physical processes that occur and has been calibrated by a series of 1g model tests. Both the tests and the mathematical modelling confirm the dependence of the progress rate on jetting power and sand density, and capture the main phenomena of sidewall collapse, sedimentation rate and overspill. Overspill (or loss of sediment from the trench) occurs due to dispersion of the sediment in the turbulent flow. The effect of cross-currents, which may be quite severe in practice, has not yet been considered. The model has also been used to optimize sword design and to assess the value of rearward facing jets to lengthen the fluidized region aft of the trencher. Note that the burial depth of the cable or pipeline also depends also on the weight, stiffness and residual tension in the cable or pipeline. Lower weight, higher stiffness or higher tension all lead to a longer span length and therefore reduced burial depth (Figure 51). A quantitative treatment of this interaction is provided by Vanden Berghe et al (2010). The unit weight (or specific gravity) of the pipeline or cable during installation must be greater than that of the fluidized soil. Moreover, there must be sufficient weight margin to ensure that the lowering of the product is not affected by the turbulence and upward flow of water behind the trencher. A minimum specific gravity is believed to be about 1.8 (but without conclusive experimental support). 7.3

Jetting of fine-grained soils is essentially a cutting process rather than an erosion process. The jets must cut and break up the material ahead of the sword. After that the essential mechanisms are similar except the clay is not fully disaggregated but remains in lumps which must be transported and maintained in suspension if a pipeline or cable is to be lowered. Conventional wisdom is that clay lumps should be removed (educted) from the trench by a dredging tool located on the trencher if good lowering is to be achieved. A water jet from a static nozzle impinging on a bed of clay creates a circular depression or cylindrical cavity in the clay often up to 3 times the nozzle diameter (Machin and Allan 2010). The depth of the cavity will depend on the jet pressure, the undrained shear strength and other properties of the soil, the offset between the bed and the nozzle (stand-off distance), and the time the jet is allowed to act. Machin et al (2001) show that the depth of cut consists of a quasi-instantaneous cut depth formed in a time span of 0.1–0.5 ms. After this, time-dependent erosion of the already-formed cut takes place. Under some circumstances, and particularly in fissured clays or clays

117 © 2011 by Taylor & Francis Group, LLC

Jetting in clay

Figure 50. Mechanisms observed during jet trenching (Vanden Berghe et al. 2008).

Figure 52. Pressure distribution of a submerged water jet acting on a target (Kondo et al. 1974). Figure 51. Lowering of a pipeline or cable in fluidized zone behind a jet trencher.

with silt or sand seams, hydraulic fracture mechanisms also operate. A basic model for clay cutting using jets requires an understanding of the mechanics of a submerged jet. If the center-line velocity of a submerged jet at the nozzle is v0 , the fluid velocity decays with distance due to spreading of the jet and entrainment of the surrounding water. The velocity, v, at the centreline of the flow at a stand-off distance x from the nozzle can be estimated by:

with the deformation of the first liquid to strike the solid; 2. A lower, quasi-steady pressure as the jet begins to flow outwards; this is associated with the pressure required to deflect the liquid sideways over the surface; 3. Shear stresses caused by liquid moving over the surface at high speed from the centre of impact.

where C is a dimensionless hydrodynamic drag coefficient (C ≈ 6.2), and d is the jet diameter. Within the initial region, up to about 6d, there is no loss in velocity. When a jet impinges on a surface, the necessary deflection of the jet fluid results in a pressure applied to the surface as illustrated in Figure 52 (Kondo et al. 1974). A solid target impacted by a liquid jet is subjected to the following pressures: 1. A strong initial compression pulse with a duration on the order of a microsecond; this is associated

118 © 2011 by Taylor & Francis Group, LLC

Considering the normal impact of a steady jet of inviscid, incompressible fluid against a rigid surface, and assuming the stand-off distance is small enough, the pressure distribution may be considered as uniform in the core area of the jet. The pressure applied to the surface is known as the stagnation pressure, p:

where ρw is the mass density of the water jet. A simple model for assessing the required jet pressure to cut clay soil can be developed by considering that the normal pressure of the jet, p, acting on the soil (Figure 52) creates a bearing failure. The reality is probably much more complex with both normal and shear stresses applied to the soil, local spalling occurring, localized drainage of the surface, erosion and variations in water pressure (hammering).

Bearing capacity failure will occur if:

where Nc is a bearing capacity factor (typically Nc = 6), or more simply the condition that p > qc , the cone tip resistance, can be used. Equations 11-13 can be combined to relate the pressure at the nozzle (p0 ) required to cut a soil of a given shear strength as a function of stand-off distance, considering that Equation 12 also applies to the pressure and velocity at the nozzle (following Bernoulli). As shown also by Machin and Allan (2010), the resulting relationship between standoff distance and the nozzle pressure to cause bearing failure is:

As a first approximation, x may also be considered the depth of cut that can be made for a jet located at the surface of the soil. The relationship is shown on Figure 53. With a jetting pressure of 20 bar from a 20 mm OD nozzle, in a soil of 150 kPa shear strength, a cut depth of about 0.2 m can be anticipated from each jet excluding any stand-off distance. From a practical perspective, considering a jetting sword of the type shown in Figure 48 with multiple jets, it is unlikely that zero stand-off distance can ever be achieved for all jets simultaneously. A minimum average stand-off might be 0.1 m if the trencher is being driven hard into the trench face. Assuming a minimum acceptable cut depth for each jet (of 20 mm OD) of 0.2 m gives x/d of 15. A typical 5 bar low pressure trenching system could be expected to operate acceptably in shear strengths of less than about 20 kPa. This agrees with general experience in the trenching industry. Low pressure (say 5 bar) systems which work effectively in sand can also create a trench in very soft to soft clays. As the bearing capacity failure occurs and the hole is deepened, the actual failure mechanism becomes more complex. The bearing capacity increases with depth, and the flow is no longer as shown on Figure 52 but is constrained inside the hole. The mechanism of entrainment of water leading to the reduction in jet velocity is quite different. However, Equation 14 is considered to provide a good indication of the ability of a static jet to cut the soil and experimental work does seem to confirm that this approach provides a conservative assessment (Machin and Allan 2010). Jet trenching involves the jets traversing rather than remaining static. Atmatzidis and Ferrin (1983) investigated the influence of time and traversing speed in the laboratory with a 1 mm diameter nozzle in clean sand, silty sand, silt and clay. The depth of jet penetration into the soil target and jet effectiveness was measured as a function of exposure time, the degree of saturation of the soil, the dry density of the soil and the traversing velocity of the jet over the soil target. Their research showed that the time required for the jet to approach maximum penetration in the soil target was about 15–20 seconds for a driving pressure

Figure 53. Required jet pressure for cutting.

Figure 54. Cutting depth as a function of jet trenching speed.

of 1000 psi (69 bar), being less for coarse-grained soil than fine-grained. It was found that the jet penetration depth, x, in a given soil is related exponentially to the corresponding time from initial impact. Machin and Allan (2010) indicate that it takes several seconds to reach the full cavity depth in clays. Atmatsidis and Ferrin (1983) also show that the penetration depth, x, could be related exponentially to the jet’s traversing velocity, v:

119 © 2011 by Taylor & Francis Group, LLC

where xmax is the limiting penetration for a stationary jet after infinite time, and ζ is an empirical constant (with units of velocity) depending on jet and material properties. Combining the static and traversing equations provides an indication of the likely depth of cut that a specific jetting system could provide, if the value of ζ were known. Machin and Allan (2010) suggest that a quasiinstantaneous cavity depth is achieved for translation speeds exceeding 0.1–0.5 m/s (360–1800 m/hr) but slower speeds are required for maximum penetration. Figure 54 shows a typical result for cutting stiff clay (su = 100 kPa) considering that the quasiinstantaneous cavity depth is achieved at a speed of 500 m/hr (assumed).

soil deformations to be observed. Other insights have emerged through finite element analysis – with notable advances being the use of coupled methods in sand to capture undrained and partially-drained ploughing behaviour, and the use of large deformation techniques to capture gross remoulding of fine-grained soils during large lateral movements.

Successful lowering of a cable or pipeline also depends on the orientation of the jets so that the soil is actually cut into blocks which can be educted or transported away from the zone where the product must sink. As far as the authors are aware, the first discussion of this issue in the public domain is given in Machin and Allan (2010). 8

CONCLUSIONS ACKNOWLEDGEMENTS

This paper has reviewed various aspects of pipeline geotechnics, by reference to recent and emerging research activity from both academia and industry. The design challenges in pipeline geotechnics differ somewhat from conventional foundation engineering. Two particular challenges are the changes in both the seabed topography and also the soil properties that can occur throughout the installation and operating life of a pipeline: these are themes that run throughout this paper. Results from novel forms of repetitive in situ testing have been used to illustrate the response of soil to the forms of loading and disturbance that are induced by a pipeline. In soft fine-grained soils, these tests illustrate the balance between the reduction in strength from remoulding and the recovery that accompanies subsequent reconsolidation. Concepts from critical state soil mechanics provide a useful framework for capturing this behaviour. Solutions for incorporating this behaviour into the estimation of axial and lateral pipe-soil resistance, and the assessment of trenching and ploughing operations, are discussed. A common theme is the relative magnitude of drained and undrained soil strengths, the evolution of these strengths, and the importance of recognising the widely-varying rates of shearing involved in pipe-soil processes.The slow rates at which pipes move under thermal loading and the high rates at which trenching machines are driven mean that pipeline geotechnics often involves a drained response in fine-grained soils and undrained behaviour in sands. The conventional laboratory tests used to characterise these soils often require modification in order to extract the properties that are relevant for pipe-soil interaction. Some of the concepts discussed in this paper represent merely a snapshot of an evolving understanding, which will no doubt advance in the coming few years. For example, the changes in soil strength through episodes of disturbance and recovery, and through cycles of partially-drained interface shearing, have only recently been observed in experiments. The underlying mechanisms are not fully established and quantitative calculation methods for design use are in their infancy. New modelling technologies have been recently applied to pipeline geotechnics. The mechanisms of pipe-soil penetration and breakout behaviour have been quantified through sophisticated centrifuge model tests, which allow complex load sequences to be imposed on a model pipe and detailed internal

The work described here forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established at the University of Western Australia in 1997 under the Australian Research Council’s Special Research Centres Program. COFS is now supported by Centre of Excellence funding from the State Government of Western Australia. The first author is supported by an Australian Research Council Future Fellowship (grant FT0991816) The assistance from a number of colleagues at UWA and Cathie Associates during the preparation of this paper is acknowledged. In particular, Fauzan Sahdi kindly provided the data in Section 2.3 from his PhD studies. The interface shear box tests in Section 5.2 were performed by Nat McNab assisted by Binaya Bhattarai. The MCWHIPLASH software (Section 4.4) was written by the first author with David Bonjean of Advanced Geomechanics, Perth. Helpful comments provided by George Zhang, also of Advanced Geomechanics, who reviewed a draft of this paper are also acknowledged. Some of the research described in this report has been guided by the SAFEBUCK Joint Industry Project, which is coordinated by David Bruton of AtkinsBoreas. The support of the SAFEBUCK participants is gratefully acknowledged. REFERENCES AtkinsBoreas. 2008. SAFEBUCK JIP: Safe design of pipelines with lateral buckling; design guideline. Report BR02050/C, AtkinsBoreas 252pp. Atmatzidis, D.K. and Ferrin, F.R. 1983. Laboratory investigation of soil cutting with a water jet, 2nd US Water Jet Conference, Rollo, Missouri, University of Missouri, 101–110. Aubeny, C.P., Shi, H., and Murff, J.D. 2005. Collapse loads for a cylinder embedded in trench in cohesive soil. ASCE Int. J. Geomechanics 5(4):320–325. Bolton, M.D. 1986. The strength and dilatancy of sands. Géotechnique 36(1):65–78. Bolton, M.D. and Barefoot, A.J. 1997. The variation of critical pipeline trench back-fill properties. Proc. of Conference on Risk-Based and Limit State Design and Operation of Pipelines, Aberdeen. Bolton, M.D., Ganesan, S.A. and White, D.J. 2009. SAFEBUCK Phase II: Axial pipe-soil resistance: summary report. Cambridge University Technical Services, Report for Boreas Consultants (SAFEBUCK JIP), ref. SC-CUTS-0705-R01. 54pp. Bransby, M.F. Yun, G.J. Morrow, D.R. and Brunning, P. 2005. The performance of pipeline ploughs in layered

120 © 2011 by Taylor & Francis Group, LLC

soils, International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, September. Bransby, M.F. and Ireland, J. 2009. Rate effects during pipeline upheaval buckling in sand. Proc. ICE Geotechnical Engineering 162: 247–256. Bruton, D.A.S., White, D.J., Cheuk, C.Y., Bolton, M.D. and Carr, M.C. 2006. Pipe-soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Proc. Offshore Technology Conference, Houston. OTC17944. Bruton, D., Carr, M. and White, D.J. 2007. The influence of pipe-soil interaction on lateral buckling and walking of pipelines: the SAFEBUCK JIP. Proc. 6th Int. Conf. on Offshore Site Investigation and Geotechnics, London. 133–150. Bruton, D., White D.J., Langford, T.L. and Hill, A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments Proc. Offshore Technology Conference, Houston, USA. Paper OTC20096. Cardoso, C.O., da Costa, A.M. and Solano, R.F. 2006. HP-HT pipeline cyclic behaviour considering soil berms effect. Proc. 25th Int. Conf. on Offshore Mechanics and Arctic Engineering, Hamburg, Germany. Cardoso, C.O. and Silviera, R.M.S. 2010. Pipe-soil interaction behavior for pipelines under large displacements on clay soils – a model for lateral residual friction factor. Offshore Technology Conference, Houston. Paper OTC20767. Carr, M., Matheson, I., Peek, R., Saunders, P. and George, N. 2004. Penguins flowline lateral buckle formation analysis and verification. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, OMAE2004-51202. Carr, M.C., Sinclair, F., and Bruton, D.A.S. 2006. Pipeline walking – understanding the field layout challenges, and analytical solutions developed for the SAFEBUCK JIP. Proc. Offshore Technology Conf., Houston, Paper OTC17945. Cathie, D.N. and Wintgens, J.F. 2001. Pipeline trenching using plows: performance and geotechnical hazards. Proc. Offshore Technology Conf., Houston. Paper OTC13145 Cathie, D.N. Jaeck, C.. Ballard, J.C and Wintgens J.F. 2005. Pipeline Geotechnics: State of the Art. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG) Perth, Australia. 95–114. Chatterjee, S., Randolph, M.F., White, D.J. and Wang, D. 2010a. Large deformation finite element analysis of vertical penetration of pipelines into the seabed. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth. Chatterjee, S., Randolph, M.F. and White, D.J. 2010b. The effects of strain rate and strain softening on the vertical penetration resistance of pipelines on an undrained seabed. Submitted for publication. Cheng, L., White, D.J., Palmer, A.C., Jas, E., Czajko, A, Fogliani, N, Fricke, R. and An, H. 2010. A new facility for research into the stability of pipelines on unstable seabeds. Offshore Pipeline Technology Conference, Amsterdam. Cheuk, C.Y., White, D.J. and Dingle, H.R.C. 2008. Upper bound plasticity analysis of a partially-embedded pipe under combined vertical and horizontal loading. Soils and Foundations 48(1):137–148. Cheuk, C.Y. and White, D.J. 2010a. Modelling the dynamic embedment of seabed pipelines. Géotechnique. Accepted August 2009, in press. Cheuk, C.Y. and White, D.J. 2010b. The influence of breakout conditions on the lateral resistance of on-bottom pipelines. Canadian Geotechnical Journal, in review. Cocchetti, G., di Prisco, C., Galli, A. and Nova, R. 2009. Soil-pipeline interaction along unstable slopes: a coupled

three-dimensional approach. Part 1: Theoretical formulation. Canadian Geotechnical Journal 46:1289–1304. Damgaard, J.S. and Palmer, A.C. 2001. Pipeline stability on a mobile and liquefied seabed: a discussion of magnitudes and engineering implications. Proc., 20th Int. Conf. on Offshore Mechanics and Arctic Engineering, Rio de Janeiro. Dean, E.T.R. 2010. Offshore Geotechnical Engineering. Thomas Telford. Dingle, H.R.C., White, D.J., and Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout on soft clay. Canadian Geotechnical Journal, 45(5):636–652. DNV 2007a. Recommended Practice DNV RP-F109, Onbottom Stability Design of Submarine Pipelines. (with amendments, April 2009). Det Norske Veritas. DNV. 2007b. Recommended Practice DNV-RP-C207, Statistical representation of soil data. Det Norske Veritas. Elsworth, D. and Lee, D.S. 2005. Permeability determination from on-the-fly piezocone sounding.ASCE J. Geotech and Geoenvironmental Engng., 131(5):643–653. Erbrich, C.T. 2005. Australian frontiers – spudcans on the edge. Proc. Int. Conf. on Frontiers in Offshore Geotechnics, Perth. 49–74. Fannin, R.J., Eliadorani,A. and Wilkinson, J.M.T. 2005. Shear strength of cohesionless soil at low stresses. Géotechnique 55, No. 6, 467–478. Finnie, I.M.S. 1993. The performance of shallow foundations in calcareous soil, PhD thesis, University of Western Australia. Finnie, I.M.S., and Randolph, M.F. 1994. Punch-through and liquefaction induced failure of shallow foundations on calcareous sediments. Proc., Int. Conf. on Behaviour of Offshore Structures, BOSS’94, Boston, 217–230. Fisher, R. and Cathie, D. 2003. Optimisation of gravity based design for subsea applications. Proc. International Conference on Foundations, ICOF, Dundee. 283–296. Gaudin, C. and White, D.J. 2009. New centrifuge modelling techniques for investigating seabed pipeline behaviour. Proc. XVIIth Int. Conf. on Soil Mechanics and Geotechnical Engineering. Alexandria. Gourvenec, S.M. and White, D.J. 2010. Elastic solutions for consolidation around seabed pipelines. Proc. Offshore Technology Conference, Houston. Paper 20554. He, J. Vasblom, W.J. Miedema, S.A. 2005. FEM analyses of cutting of anisotropic densely compacted and saturated sand, Western Dredging Association Conference, WEDAXXV and TAMU37, New Orleans, USA, June and www.dredgingengineering.com/dredging Hettiaratchi, D.R.P. and Reece, A.R. 1974. The calculation of passive soil resistance, Géotechnique, 24(3):289–310. Hettiaratchi, D.R.P. and Reece, A.R. 1975. Boundary wedges in two-dimensional passive soil failure, Géotechnique, 25(2):197–220. Hill, A.J. and Jacob, H. (2008) In-Situ Measurement of PipeSoil Interaction in Deep Water. Proc. Offshore Technology Conference, Houston, USA. Paper OTC 19528. Hossain, M.S. and Randolph, M.F. 2009. Effect of strain rate and strain softening on the penetration resistance of spudcan foundations on clay. ASCE Int. J. Geomech. 9(3):122–132. Jayson, D., Delaporte, P., Albert, J-P, Prevost, M.E., Bruton, D. and Sinclair, F. 2008. Greater Plutonio project – Subsea flowline design and performance. Offshore Pipeline Technology Conf., Amsterdam. Jordon P, and Cathie, D.N. 2004. Developments in cable protection by burial. SubOptic 2004. Karmakar , S. 2005. Numerical modelling of soil flow and pressure distribution on a simple tillage tool using

121 © 2011 by Taylor & Francis Group, LLC

computational fluid dynamics, PhD thesis, University of Saskatchewan. Kondo, M. Fujii, K. and Syoji, H. 1974. On the destruction of mortar specimens by submerged water jets, 2nd International Symposium on Jet Cutting Technology, Cambridge, paper B5, 69–88. Krost K., Gourvenec, S.M. and White, D.J. 2010. Consolidation around partially-embedded submarine pipelines. Géotechnique, Accepted December 2008, in press. Lacasse, S., Gutteromsen, T., Nadim, F., Rahim, A., Lunne, T. 2007. Use of statistical methods for selecting design soil parameters. Proc. 6th Int. Offshore Site Investigation and Geotechnics Conference, London, UK, 449–460. Langford, T.E., Dyvik, R. and Cleave, R. 2007. Offshore pipeline and riser geotechnical model testing: practice and interpretation. Proc. Conf. on Offshore Mech. and Arctic Eng., San Diego. Paper OMAE2007-29458. Lauder, K. 2010. Predicting pipeline plough performance by scale model testing, Presentation to CTC, April, University of Dundee. Lauder, K. Bransby, F. Brown, M. Cathie, D. Morgan, N, Pyrah, J. and Steward, J. 2008. Experimental testing of the performance of pipeline ploughs, 18th Int. Offshore and Polar Engineering Conf. ISOPE-2008, Vancouver, Canada, 212–217. Lenci, S. and Callegari., M. 2005. Simple analytical models for the J-lay problem. Acta Mechanica, 178:23–39. Lumb, P. 1966. The variability of natural soils. Canadian Geotechnical Journal. 3:74–97. Lund, K.M. 2000. Effect of increase in pipeline soil penetration from installation. Proc. of ETCE/OMAE 2000 Joint Conference; Energy of the New Millennium OMAE2000/ PIPE-5047. Machin, J.B. and Allan, P.J.A. 2010. State-of-the-art jet trenching analysis in stiff clays, 2nd International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, November. Machin J.B. Messina, F.D. Mangal, J.K. Girard, J. and Finch, M. 2001. Recent research on stiff clay jetting, Offshore Technology Conference,Houston, Paper OTC 13139. Merifield, R., White, D.J. and Randolph, M.F. 2008a.Analysis of the undrained breakout resistance of partially embedded pipelines. Géotechnique, 58(6)461–470. Merifield, R.S, White, D.J. and Randolph, M.F. 2008b. The effect of pipe-soil interface conditions on undrained breakout resistance of partially-embedded pipelines. Proc. Int. Conf. on Advances in Computer Meth. and Analysis in Geomech. Goa, India. Merifield, R., White, D.J. and Randolph, M.F. 2009. The effect of surface heave on the response of partiallyembedded pipelines on clay. ASCE J. Geotechnical and Geoenvironmental Engineering, 135(6):819–829. Miedema, S.A. 1987. Calculation of the cutting forces when cutting water saturated sand, PhD dissertation, Delft University of Technology. Miedema, S.A. 2005. The cutting of water saturated sand, the final solution, WEDAXXV and TAMU37, New Orleans, USA, June, and www.dredgingengineering.com/dredging. Morrow, D.R. and Bransby, M.F. 2009. The influence of slope on the stability of pipelines subjected to horizontal and vertical loading on clay seabeds. Proc. Conf. on Offshore Mechanics and Arctic Engineering. OMAE2009-79050. Orcina. 2008. OrcaFlex software, Orcina Ltd, Ulverston, UK Palmer, A.C., Kenny, J.P., Perera, M.R. and Reece, A.R. 1979. Design and operation of an underwater pipeline trenching plough. Géotechnique 29(3):305–322. Palmer, A.C. 1999. Speed effects in cutting and ploughing. Géotechnique 49(3):285–294.

Pedersen, R.C., Olsen, R.E. and Rausch, A.F. 2003. Shear and interface strength of clay at very low effective stresses. ASTM Geotechnical Testing J., 26(1):71–783. Peng W, and Bransby M.F. 2010. Numerical modelling of soil around offshore pipeline plough shares, 2nd International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, November. Peng A.T.H and Capart H. 2008 Underwater sand bed erosion and internal jump formation by travelling plane jets, Journal of Fluid Mechanics, 595:1–43. Puech, A. and Foray, P. 2002. Refined model for interpreting shallow penetration CPTs in sands. Proc. Offshore Technol. Conf., Houston, Paper OTC14275. Randolph, M.F. 2003. Science and empiricism in pile foundation design. Géotechnique. 53(10): 847–875. Randolph, M.F. and Hope, S.N. 2004. Effect of cone velocity on cone resistance and excess pore pressure. Proc. Conf. on Engineering Practice and Performance of Soft Deposits, Osaka, 147–152. Randolph, M.F. and White, D.J. 2008a. Pipeline embedment in deep water: processes and quantitative assessment. Proc. Offshore Technology Conference, Houston, USA. Paper OTC19128-PP. Randolph, M.F. and White, D.J. 2008b. Upper bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique, 58(4):297–301. Randolph M.F. and Gourvenec S.M. 2010. Offshore Geotechnical Engineering. Taylor and Francis. Rathbone, A. Hakim, M. A. Cumming, G. and Tørnes, K., 2008. Reliability of lateral buckling formation from planned and unplanned buckle sites. Proc. 27th International Conference on Offshore Mechanics and Arctic Engineering, OMAE2008-57300, Estoril, Portugal. Reece, A.R. and Grinsted, T.W. 1986. Soil mechanics of submarine ploughs, Offshore Technology Conference, OTC 5341. Schotman, G.J.M. and Stork, F.G. 1987. Pipe-soil interaction: a model for laterally loaded pipelines in clay. Proc. Offshore Technology Conference, Houston, OTC5588. Silva, M.F. 2005. Numerical and physical models of rate effects in soil penetration. PhD thesis, University of Cambridge. Steenfelt, J.S. 1993. Sliding resistance for foundations on clay till. Proc. Wroth Memorial Conference, Predictive Soil Mechanics. Thomas Telford. 664–684. Sture, S., Costes, N. C., Batiste, S.N., Lankton, M.R., AlShibli, K.A., Jeremic, B., Swanson, R.A. and Frank, M. 1998. Mechanics of granular materials at low effective stresses. ASCE J. Aerospace Engng 11(3): 67–72. Tian,Y., and Cassidy, M.J. 2008. Explicit and implicit integration algorithms for an elastoplastic pipe-soil interaction macroelement model. Proc. 27th International Conference on Offshore Mechanics and Arctic Engineering, OMAE2008-57237, Estoril, Portugal. Tornes, K., Jury, J. and Ose, B. 2000. Axial creeping of high temperature flowlines caused by soil racheting. Proc. Conf. on Offshore Mechanics and Arctic Engineering, OMAE-PIPE5055. Verley, R. and Lund, K.M. 1995. A soil resistance model for pipelines placed on clay soils. Proc. International Conference on Offshore Mechanics and Arctic Engineering (OMAE), Copenhagen, Denmark, 5: 225–232. Verley, R.L.P. and Sotberg, T. 1994. A soil resistance model for pipelines placed on sandy soils. ASME J. Offshore Mechanics and Arctic Engineering, 116(3):145–153. Van Leussen, W. and Nieuwenhuis, J.D. 1984. Soil mechanics aspects of dredging, Geotechnique 34(3): 359–381.

122 © 2011 by Taylor & Francis Group, LLC

Van Os, A.G. and van Leussen, W. 1987. Basic research on cutting forces in saturated sand, Journal of Geotechnical Engineering, 113 (12):1501–1516. Van Rhee, C. and Steeghs, H.J.M.G. (1991) Multi-blade ploughs in saturated sand; model cutting tests. Dredging and Port Constructions, June, 37–39. Vanden Berghe, J.-F. Pyrah, J. Gooding, S. and Capart, H. 2010. Development of a jet trenching model in sand, 2nd International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, November. Vanden Berghe, J.-F. Capart, H. and Su, J.C.C. 2008. Jet induced trenching operations: mechanisms involved. Offshore Technology Conference, Houston. Paper OTC 19441. Wang, D., White, D.J. and Randolph, M.F. 2009. Numerical simulations of dynamic embedment during pipelaying on soft clay. Proc. Conf. on Offshore Mechanics and Arctic Engineering, Honolulu. Paper OMAE2009-79199. Wang, D., White, D.J. and Randolph, M.F. 2010. Large deformation finite element analysis of pipe penetration and large-amplitude lateral displacement. Canadian Geotechnical Journal. Accepted April 2009, in press. Westgate, Z., White, D.J. and Randolph, M.F. 2009. Video observations of dynamic embedment during pipelaying on soft clay. Proc. Conf. on Offshore Mechanics and Arctic Engineering, Honolulu. Paper OMAE2009-79814. Westgate, Z.J., Randolph, M.F., White, D.J. and Li, S. 2010a. The influence of seastate on as-laid pipeline embedment: a case study. Applied Ocean Research, Available online, 4 February 2010. doi:10.1016/j.apor.2009.12.004. Westgate, Z.W., White, D.J., Randolph, M.F. and Brunning, P. 2010b. Pipeline laying and embedment in soft fine-grained soils: Field observations and numerical simulations. Proc. OffshoreTechnology Conference, Houston. Paper 20407. White, D.J. and Randolph, M.F. 2007. Seabed characterisation and models for pipeline-soil interaction Proc. 17th Int. Offshore and Polar Engng. Conference, Lisbon, Portugal. CD 11pp. White, D.J. and Gaudin, C. 2008. Simulation of seabed pipe-soil interaction using geotechnical centrifuge modelling. Proc. 1st Asia-Pacific Deep Offshore Technology Conference, Perth, Dec 2008. White, D.J. and Cheuk, C.Y. 2008. Modelling the soil resistance on seabed pipelines during large cycles of lateral movement. Marine Structures 21(1):59–79.

White, D.J. and Cheuk, C.Y. 2009. SAFEBUCK JIP: Pipe-soil interaction models for lateral buckling design: Phase IIA data review. Report to Boreas Consultants (SAFEBUCK JIP), UWA report GEO 09497r1. 185pp. White, D.J. and Dingle, H.R.C. 2010. The mechanism of steady ‘friction’ between seabed pipelines and clay soils. Géotechnique. Accepted April 2009, in press. White, D.J. and Hodder, M. 2010. A simple model for the effect on soil strength of remoulding and reconsolidation. Canadian Geotechnical Journal. Accepted November 2009, in press. White, D.J., Gaudin, C., Boylan, N and Zhou, H. 2010a. Interpretation of T-bar penetrometer tests at shallow embedment and in very soft soils. Canadian Geotechnical Journal, 47(2):218–229. White, D.J., Hill, A.J., Westgate, Z. and Ballard, J-C 2010b. Observations of pipe-soil response from the first deepwater deployment of the SMARTPIPE. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth. Wroth, C.P. 1984. The interpretation of in-situ tests. 24th. Rankine Lecture, Géotechnique, 34(4):449–489. Zhang, J., Stewart, D. P., and Randolph, M. F. 2002a. Modelling of shallowly embedded offshore pipelines in calcareous sand. ASCE Journal of Geotechnical and Geoenvironmental Engineering, 128(5):363–371. Zhang, J., Stewart, D. P., and Randolph, M. F. 2002b. Kinematic hardening model for pipeline-soil interaction under various loading conditions. ASCE International Journal of Geomechanics, 2(4):419–446. Zhang, J., and Erbrich, C.T. 2005. Stability design of untrenched pipelines – geotechnical aspects. Proc. International Symposium on Frontiers in Offshore Geo-technics (ISFOG) Perth, Australia. 623–628. Zhou, H., White, D.J. and Randolph, M.F. 2008. Physical and numerical simulation of shallow penetration of a cylindrical object into soft clay. Proc. GeoCongress 2008, New Orleans,ASCE Special Geotechnical Publication No. 179, 108–117. Zhou, H. and Randolph, M. F. 2009. Penetration resistance ofcylindrical and spherical penetrometers in rate-dependent and strain-softening clay. Géotechnique. 59(2)79–86.

123 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Axial and lateral pile design in carbonate soils C.T. Erbrich, M.P. O’Neill, P. Clancy & M.F. Randolph Advanced Geomechanics, Perth, Australia

ABSTRACT: Design of piles in carbonate soils remains one of the most challenging geotechnical problems, fraught with traps for the unwary. Whilst carbonate soils generally behave in accordance with the same underlying rules of soil mechanics that are applicable to all soils, they often exhibit characteristics at the extremes of this continuum of behaviour. Carbonate soils themselves straddle a wide range of different material types, varying from soft uncemented silts and muds to dense well cemented calcarenites and limestones. We also find dense uncemented carbonate soils and loose and compressible cemented soft rocks (calcarenites, etc.). All combinations of these are encountered offshore Australia. This paper presents new work that addresses two key aspects of pile design in this array of carbonate material types: axial design of drilled and grouted piles in compressible cemented carbonate soils and lateral design of piles (drilled and grouted or driven) in uncemented carbonate soils, varying from soft muds to dense sands. The new work on axial pile design builds on the extensive work undertaken in the 1980s following the problems with the North Rankin A platform, and which was subsequently put to good use in the design of the nearby Goodwyn A platform. However, more recent developments have identified a need to develop a more generalised approach that can be applied at sites with a variety of characteristics. In addition, more detailed review of some of the data collected during the original work has revealed important new mechanisms that are of considerable significance but have hitherto been neglected. The new work on lateral pile design follows a different tack from any of the existing methodologies used for assessing lateral pile response in such soils, which are essentially of a purely empirical form, with minimal theoretical basis. A new method with a robust theoretical basis is presented for assessing the lateral pile response in carbonate soils, which particularly accounts for their well known strong susceptibility to degradation (liquefaction) when subject to cyclic loading. 1

INTRODUCTION

The axial and lateral response of piles is analysed almost universally in the offshore industry using load transfer approaches. Interaction between the pile and the soil is treated through non-linear springs distributed down the pile shaft, with each horizontal layer of soil treated independently. The load transfer curve thus represents the integrated response of the soil layer extending outwards from the pile. The early development of load transfer curves followed an empirical approach, derived from the results of instrumented pile load tests, with limited guidance on how to scale the curves for different pile diameters or soil stiffness. More rational frameworks have since been established, aided by modern numerical analysis, but there is still an underlying need for experimental data for aspects such as the effects of cyclic loading. The key ingredients of load transfer curves may be summarised as follows:

loading) or net lateral pressure (force per unit projected area, for lateral load transfer). 3. Post-peak softening response, particularly for axial load transfer but also for lateral response in cemented soils. 4. Allowance for the effects of cyclic loading.

1. A pre-failure portion, with initial gradient linked to the shear modulus of the soil and a non-linear transition to a limiting load mobilised at an appropriate local displacement. 2. Limiting loads under monotonic loading, generally expressed as an equivalent shaft friction (axial

125 © 2011 by Taylor & Francis Group, LLC

As in much offshore design, the last aspect above provides the greatest challenge. The most common approach has been implicit, reducing the limiting load transfer for cyclic conditions and assessing the pile response by means of an equivalent monotonic analysis, rather than attempt to carry out cycle by cycle analyses where degradation of the load transfer response is modelled explicitly. In many soil types, post-peak softening under monotonic loading is relatively minor (the API guidelines suggest a maximum reduction of 30% under axial loading) and the implicit approach for cyclic loading has proved sufficient. However, post-peak softening for piles grouted into cemented carbonate soils can be dramatic, and the effect on the overall pile capacity must be addressed in detail. This is also true in respect of lateral loading of piles driven or grouted into cemented sediments, where cracks in the surrounding sediments can extend to the free surface. In addition, pile tests at laboratory and field scale in carbonate sediments have shown that the

response under axial cyclic loading is complex, and with potential for accumulating damage to lead to progressive failure of the entire pile. A brief background is provided here with respect to the load transfer algorithms that have been developed over the last two or three decades for pile response, and in particular their application to carbonate sediments. 1.1 Axial load transfer Load transfer analysis of the axial response of piles stems from the 1960s (Coyle and Reese 1966). The initial stiffness of the pre-failure load transfer response may be linked directly to the shear modulus of the soil (Randolph and Wroth 1978), and the approach can be extended to allow for non-linear soil response, by integrating the stress-strain response radially outwards from the pile (Kraft et al. 1981). For a typical hyperbolic stress-strain response, the resulting load transfer curve turns out to match closely an inverted parabola, with an initial gradient that is about double the secant gradient at full mobilisation of the ultimate friction. This form of curve is also very similar to the generic shape recommended in the API guidelines (API RP2A 2000). In siliceous clays and sands, ratios of shear modulus to limiting shaft friction result in a load transfer curve where full mobilisation occurs at a local pile displacement of 0.25 to 2% of the pile diameter, with a typical value of around 1% (Jeanjean et al. 2010). In cemented material, it is appropriate to adopt a somewhat stiffer load transfer response, partly because of higher ratios of shear modulus to limiting shaft friction, but also because this is conservative from the point of view of increased stress concentration in the upper part of the pile, hence greater damage during cyclic loading. The load transfer software, RATZ, was developed originally in the mid 1980s (Randolph 1986), and at that stage was calibrated to match the response of laboratory rod shear and constant normal stiffness (CNS) tests, and subsequently field grouted pile tests (Randolph 1988, Randolph & Jewell 1989, Randolph et al. 1996). The program uses an explicit approach, similar to the commercial finite difference code, FLAC, to follow the non-linear response of the pile under either monotonic or cyclic loading. The current version of the software is based on Excel for convenience of data input and viewing results, although the numerical computations are carried out in a Fortran-based DLL (Randolph 2003). Details of the load transfer algorithm are described later, in relation to the CYCLOPS software, which was later developed from RATZ. The load transfer algorithm includes a non-linear pre-failure response, post-failure strain-softening, cyclic degradation calculated through a damage algorithm based on accumulated irreversible relative pile-soil displacement, and a low cyclic residual friction within the current range of plastic pile-soil displacement followed by (partial) recovery for displacements outside that zone (see Figure 1). As discussed later, the low cyclic residual

Figure 1. Schematic of load transfer algorithm in RATZ.

(CFG in Figure 1) is a key feature observed in grouted pile and CNS test responses. An important distinguishing feature of the load transfer analysis in RATZ is the provision for cycle-bycycle analysis of a complete storm loading sequence acting on the pile. This allows simulation of gradual ratcheting, and transfer of load progressively down the pile as the more heavily loaded soil in the upper part of the compressing pile accumulates displacement and softens. 1.2

126 © 2011 by Taylor & Francis Group, LLC

Lateral load transfer

Forms of load transfer curves for lateral pile response are recommended for siliceous clays and sands in the API and ISO guidelines (API 2000, ISO 2007). These are mostly based on approaches developed in the 1970s by Reese, Matlock and others. At the time of the strut strengthening developed for first generation platforms in Australia’s Bass Strait (Wiltsie et al. 1988), Exxon undertook an extensive suite of centrifuge model and small field tests in order to develop lateral load transfer curves specifically for carbonate sands (Wesselink et al. 1988). The above work led to a generic load transfer curve for monotonic loading, where the lateral load per unit length, P (or net pressure, p = P/D), was expressed as a power law function of the cone resistance, qc , and the normalised displacement, y/d. A normalised version of the load transfer formulation was later proposed, following a more extensive series of centrifuge model tests (Dyson and Randolph 2001), expressed as:

where γ
is the effective unit weight of the soil. Best fit values of the parameters, R, n and m, were established as 2.7, 0.72 and 0.58 respectively. A similar expression was proposed by Novello (1999), given by

The inclusion of a depth term, z/D, in this relationship means that, in general, the ratio, p/qc , increases more rapidly with depth than using Eq. (1). For example, if the cone resistance were to vary proportionally with depth, Eq. (2) would result in the lateral load for a given displacement, y, also being proportional to depth, rather than varying with depth to the power of 0.72 using Eq. (1). In other respects though, the two formulations are quite similar. Carbonate sediments tend to have high friction angles, in the region of 40◦ . However, comparing the above relationships with the API load transfer formulations for siliceous sand, the resulting response tends to be bracketed by the API curves for 20◦ and 35◦ , at least for displacements less than 0.1D. At larger displacements, the carbonate soil formulations give lateral resistance in excess of that for the 35◦ siliceous sand (Wesselink et al. 1988). The forms of load transfer curve given by Equations (1) and (2) are not bounded as y increases, and also have an infinite initial gradient (although a finite value of around 4G0 may be adopted for numerical implementation, where G0 is the small strain shear modulus). For practical values of deflection, with y << D, the net lateral pressure on the pile remains as a small fraction of the cone resistance. Thus for qc = 10 MPa, and γ
D = 20 kPa, Eq. (1) gives a net pressure, p, of 1.86 MPa for y = 0.2D, and remains below 0.5qc even for y ∼ D. The effects of cyclic loading were explicitly considered by Novello (1999), through the introduction of a multiplier on the computed monotonic resistance:

where U* is the pore pressure ratio, varying from zero to 1.0 for a fully liquefied soil, and K0 is the coefficient of earth pressure at rest. For cemented carbonate soils, commonly encountered at shallow depth in the form of cap rock, attention must be paid to the brittle nature of the stress-strain response, and the potential for cracked fragments of cemented material to fail towards the free surface. A strain-softening p-y model to simulate this was proposed by Abbs (1983). However, a more rational approach was later developed by Erbrich (2004) for application in the carbonate deposits on the North-West Shelf of Australia. The basis of the Erbrich model is the concept of wedges of ‘chipped’ material forming near the surface. This concept was combined with the kinematic mechanism of Muff and Hamilton (1993) in order to evaluate the net resistance once a chip had occurred, and to develop a criterion for the maximum depth to which chips, as opposed to deformation of intact rock (which offers significantly higher capacity than chipped material), would occur. The ‘chipper’ model proposed by Erbrich (2004) was calibrated by means of 3D finite element analyses, and also through comparisons with centrifuge

model tests. It was found to yield greater lateral pile capacity than the Abbs (1983) model, but provided a sound physical basis on which many anchor piles have been designed offshore Australia and in other regions where cemented carbonate sediments occur. 2 2.1

Introduction

After experiencing problems with the driven piles at North Rankin A, which largely free-fell to their design penetration with minimal driving, consideration once again turned to drilled and grouted piles as the preferred foundation system in carbonate soils and soft rocks (calcarenites). However, the experience with the free-falling North Rankin piles, the results of various small scale model tests and the results of direct shear CNS tests suggested that a very stiff and brittle response might be expected for a grouted insert in these calcarenites. A major project was therefore commenced which was intended to measure the performance of grouted pile sections of varying sizes (up to 2 m diameter) in presumed similar soils to those at North Rankin A (Randolph, et al. 1996). A site was identified at Overland Corner (OC), South Australia where cemented carbonate material with large void ratios (i.e. apparently similar characteristics to the North West Shelf material) was found. The results from this test program formed the basis of design for the drilled and grouted piles adopted for the nearby Goodwyn A platform which is located in very similar soils to those found at NRA. This remains the largest structure built to date in Australian waters with drilled and grouted piles. The RATZ model was the principal design tool used in the GWA design, with the key parameters tuned to give the best fit to the OC grouted section tests (GSTs). CNS direct shear tests were undertaken for both the GWA and OC soils, but these were treated only as qualitative information and were not used directly in the design of GWA or the back-analysis of the OC grouted sections. More recently it has been necessary to design large diameter drilled and grouted piles at a number of other sites offshore Australia. For these cases, CNS direct shear tests suggested potentially quite different behaviour to that observed at GWA and OC for both monotonic and cyclic loading. This seemed conceptually in keeping with the developing understanding of the underlying soil mechanics of these soils (i.e. the effect of differing density, cementation and stress level on the foundation performance), but raised a challenge as to how to best deal with this variability in the design process. It was felt that simple reliance on the calibrated GWA/OC design parameters could not be justified. Instead the axial pile response was developed by directly matching the modelled response to the CNS test response. It was considered that this approach should be robust and inherently conservative, since it was thought that scale effects associated with shearing an interface in a small scale CNS test would (usually)

127 © 2011 by Taylor & Francis Group, LLC

PART 1 – AXIAL PILE DESIGN

may vary along the length. Interaction with the soil is characterised by discreet non-linear ‘t-z’ load transfer springs. The t-z curve is a function that relates the local shaft friction at the pile-soil interface to the current local pile displacement, accounting for the previous displacement history. The load transfer function takes full account of cyclic loading, and allows for accumulation of displacement and degradation of shaft friction. 2.2.1 Form of t-z curve The general form of the t-z load transfer curve adopted in CYCLOPS is presented schematically on Figure 3. The parameters defining the shape of the t-z curve include: Figure 2. Comparison between CNS test and RATZ.

lead to a more brittle response than obtained at the grout-soil interface for a drilled and grouted pile in the field; the original OC calibration also appeared to lend support to this notion. Once this CNS test calibration process was started in earnest, it soon became apparent that a number of the characteristic features commonly observed in a CNS test (and indeed also apparent in the GSTs) could not be reproduced with the existing RATZ model. An example illustrating these limitations is presented on Figure 2, which presents both the raw CNS test data and the ‘best fit’ achieved with the RATZ model. A key problem is that the RATZ model exhibits a ‘binary’ response during cyclic loading; while cycling within the limit of previously applied cyclic displacement amplitudes only a very low ‘cyclic residual’ friction is mobilised, but once the applied cyclic displacement is increased to and beyond the previous limit it instantaneously recovers to the full post-cyclic strength. This is clearly very different from the actual test data and seems to suggest both potentially conservative and unconservative characteristics. The former because the size of the zone where only the cyclic residual can be mobilised is overestimated, which may lead to increased redistribution of the axial pile load down a pile during cyclic loading, but the latter because a much larger pile displacement will be required to mobilise the post-cyclic strength than predicted by RATZ. It was far from obvious whether these effects would cancel out or whether or not an overall conservative or unconservative bias would be found. It was therefore considered essential to incorporate these effects into any future work. The CYCLOPS model was thus born, which is an enhancement of the RATZ model to enable much improved simulation of the observed behaviour in CNS tests and GSTs. The CYCLOPS model will be described in the next section. 2.2

Model description (CYCLOPS)

CYCLOPS was developed from the computer program RATZ (Randolph, 2003). Both RATZ and CYCLOPS treat the pile as an elastic bar with properties that

– The elastic threshold (ξ). – The peak shaft friction (τpeak ). – The displacement to τpeak (wpeak ), defined in part by the maximum shear modulus (Gmax ). – The initial strain softening and strain softening parameters (i-η and η respectively). – The residual shaft friction (τres ). – The displacement from τpeak to τres (wres ). – The cyclic residual shaft friction (τcyc- res ). – The bias parameter (b), defined in turn by the initial and final bias parameters (bi and bf respectively). – The bias parameter exponent (m). – The cyclic transition width (wcyc ). The parameters ξ, Gmax , τpeak , i-η, η, τres and wres define the shape of the ‘backbone’ monotonic failure curve, indicated by the light solid line on Figure 3. A demonstration of the influence of wres on the monotonic response is presented on Figure 4, which shows monotonic post-τpeak failure curves for a uniform set of parameters and for various wres values ranging between 0.1 m and 1.5 m. Similar sets of curves are presented on Figure 5 for various η values ranging between 0.1 and 0.9, and on Figure 6 for various τres /τpeak values ranging between 0.1 and 0.9. A point worth noting is that a specific post-τpeak monotonic response described by given values of wres , η and τres /τpeak can be approximately matched by various alternative combinations of these parameters over a limited displacement range. This point is demonstrated on Figure 7, which shows the postτpeak monotonic response for a variety of cases, all of which give a similar response over the initial 0.2 m of displacement. It can be seen that this similitude of response is obtained where low values of wres are combined with high values of τres /τpeak at one extreme, and high values of wres are combined with low values of τres /τpeak at the other extreme. The parameters bi , bf , m and to a lesser extent τcyc-res define the relative shape of the maximum bias stress (τwmax ) curve, indicated by the dashed line on Figure 3. For any given total plastic displacement, the maximum bias stress curve essentially ‘mirrors’ the monotonic failure curve according to the following relationship:

128 © 2011 by Taylor & Francis Group, LLC

Figure 3. Details of CYCLOPS t-z load transfer curve.

Figure 4. CYCLOPS – Displacement to residual.

Figure 5. CYCLOPS – Strain softening parameter.

This is demonstrated on Figure 3, which shows a nominal load path that would be followed by a two-way displacement controlled cyclic CNS test. This example test comprises 3 cycles of loading to ±wtest (where wtest > wpeak ) followed by ‘positive’ monotonic loading. During the first cycle the load path follows the monotonic curve through τpeak down to τfail-1 . The sample is then displaced −2wtest and then back to the origin. An important aspect to note at this point is development of the cyclic ‘gap zone’ which, following completion of the first cycle, exists between −wtest and +wtest and represents the maximum extent of the plastic pile-soil relative displacement. The second cycle commences and the sample is loaded back to +wtest . Relative to the previous

cycle when the sample was at this displacement, the total plastic displacement that the sample has been subjected to is approximately 4wtest , which is equivalent to a positive displacement along the ‘backbone’ monotonic curve of (approximately) 4wtest . Hence, the strength on the ‘backbone’ monotonic curve has reduced from τfail-1 to τfail-2 . At the edge of the gap zone the shear stress mobilised is therefore now equal to τwmax-2 in accordance with Equation (4). After another full cycle down to a total negative displacement of −wtest followed by reloading back to +wtest the maximum bias stress reduces to τwmax-3 , and similarly after another full cycle the maximum bias stress at the end of the gap zone is τwmax-4 . As the sample is then monotonically loaded past +wtest

129 © 2011 by Taylor & Francis Group, LLC

Figure 6. CYCLOPS – Residual shaft friction. Figure 8. CNS test comparison – RATZ and CYCLOPS.

2.2.2 Enhancements specific to CYCLOPS The CYCLOPS t-z algorithm incorporates a number of enhancements compared to the RATZ t-z model, namely: – Inclusion of a variable bias parameter (defined by bi and bf ). – Inclusion of an initial strain softening parameter (i-η). – Modification of the shape of the t-z curve within and immediately outside of the ‘gap zone’ formed during cyclic loading. The following sections discuss in detail the reasons for and implementation of the enhancements outlined above. Figure 7. CYCLOPS – Matching of various parameter sets.

and beyond the gap zone, the shear stress path gradually transitions from the maximum bias stress curve (i.e. τwmax-4 ) to the ‘backbone’monotonic failure curve (i.e. τfail-4 ) over a displacement equal to the cyclic transition width (wcyc ). The displacement controlled cyclic CNS test example described above illustrates an important aspect of the CYCLOPS t-z model, in that the amount of post-τpeak plastic displacement that occurs within the gap zone during cycling corresponds to an equivalent amount of plastic displacement along the monotonic τfail curve, and therefore essentially defines the magnitude of τfail at any point during cycling. It should be noted, however, that in the example test description above, it was assumed that all of the post-τpeak displacement was plastic. CYCLOPS actually assumes that a small component of the displacement is elastic (as determined by the maximum shear modulus Gmax ), and this elastic component is not included in the ‘total’ accumulation of plastic displacement that is used to calculate τfail .

2.2.3 Variable bias parameter As briefly described in Section 2.2.1, the bias parameter (b) defines the maximum bias shear stress (τwmax ) at the edge of the gap zone during cycling, primarily as a function of the monotonic failure stress (τfail ). The importance of the bias parameter can be demonstrated by examining a typical displacement controlled cyclic CNS test response. Such a response is shown plotted as shear stress (τ) versus horizontal displacement on Figure 2 and as the maximum shear stress attained during each cycle (τmax ) versus cycle number on Figure 8. The test comprised 25 cycles of displacement controlled loading at ±5 mm followed by monotonic loading to +12 mm, then to −5 mm and then back to the origin. The two points of interest to this discussion are:

130 © 2011 by Taylor & Francis Group, LLC

– The rate at which τmax reduces with increasing cycle number. – The recovery of τ during the post-cyclic monotonic loading phase (i.e. Cycle 26 on Figure 8) as the sample is displaced beyond the ±5 mm gap zone. Figure 8 also shows the calibrated CNS test response as determined using RATZ. In RATZ, b has an implied value of 1.0, such that the maximum bias

Figure 10. CNS test and CYCLOPS comparison – Example 1. Figure 9. Variable bias parameter definition.

shear stress curve illustrated on Figure 3 is identical to the corresponding ‘backbone’ monotonic failure curve.As indicated on Figure 8, there is no increase in τ as the sample is displaced beyond the ±5 mm gap zone at cycle 26. In attempting to fit RATZ to this test, one can decide to set τres equal to the fully recovered postcyclic test value as adopted for the response shown on Figure 8, but the drawback of this is that the τmax values obtained in each cycle are significantly (and non-conservatively) greater than the actual test values. Alternatively, one could decide to set τres equal to the value of τmax at the end of cycling. This would result in a better fit to the τmax versus cycle number response, but would fail to capture the post-cyclic recovery in τ to the final τres . Hence, for these reasons it was considered essential to introduce a user-specified bias b into the CYCLOPS model. However, further detailed evaluation of various CNS tests results demonstrated that b not only varied from sample to sample but also often decreased in value over the course of each test. Therefore, CYCLOPS incorporates a non-constant b, which is specified in terms of initial and final values (bi and bf respectively). The manner in which CYCLOPS calculates the value of b at any given stage of cycling can be expressed as:

The parameter Ngap is the number of load cycles where the axial displacement is sufficient to reach either side of the edge of the gap zone, while m is a simple exponent. Figure 9 shows the relationship between b (normalised with respect to bi and bf ) and Ngap for various values of m. The benefits of adopting a non-uniform b are illustrated by two CYCLOPS calibrated responses to the previously considered CNS test. Results from the first CYCLOPS fit (Example 1; uniform b) are presented on Figure 10 (τ versus horizontal displacement) and

Figure 11. CNS test and CYCLOPS comparison – Example 2.

Figure 8 (τmax versus cycle number). Identical parameters to those adopted for the RATZ fit were used, with the exception of i-η (which is not included in RATZ), wcyc (implied value of 0 mm in RATZ) and bi = bf which were both set to 0.21. The resulting CYCLOPS response shows significantly improved agreement with the observed test behaviour compared to RATZ, particularly with regards to τmax during the latter stages of cycling and the post-cyclic recovery to τres . However, it can be seen that during early stages of cycling CYCLOPS still underpredicts τmax , which is somewhat overconservative, and implies that a value of b greater than 0.21 should be adopted for the first few cycles. The second CYCLOPS fit presented on Figure 11 and Figure 8 (Example 2; variable b) adopted identical t-z parameters to Example 1 but with bi and bf set to 0.68 and 0.21 respectively. The resulting predicted response shows generally excellent agreement with the test response, highlighting the benefit of adopting a non-uniform b. 2.2.4 Initial strain softening parameter The ‘standard’ strain softening parameter (η) is an exponent that controls the rate at which the postτpeak monotonic failure stress (τfail ) decreases with

131 © 2011 by Taylor & Francis Group, LLC

Figure 12. Initial post-peak strain softening parameter.

Figure 13. Shear stress within gap zone.

increasing displacement, as demonstrated on Figure 5. Calcarenites typical of those encountered on the North West Shelf generally exhibit a fairly brittle post-τpeak response, with η values usually less than 0.5. For these relatively low η values, the negative gradient of the τfail versus displacement curve immediately following τpeak can be very large (in absolute terms), and it has been found previously that this large gradient can result in a slight overprediction of the degradation in shaft friction resulting from cyclic loading. In order to avoid this problem an additional parameter, termed the initial strain softening parameter (i-η), was incorporated into the CYCLOPS monotonic τfail algorithm which enables the immediate post-τpeak ductility of the τfail response curve to be increased slightly. The influence of i-η on τfail is illustrated on Figure 12, which plots τ/τpeak versus the post-τpeak displacement normalised by wres , assuming η = 0.4 and for various values of i-η. It should be noted that for a relatively large i-η value the CYCLOPS response matches the ‘baseline’ RATZ response (i.e. as shown by the iη = 1 × 106 curve on Figure 12). As i-η decreases, the post-τpeak displacement at which the CYCLOPS response curve rejoins the baseline (i-η = 1 × 106 ) response curve increases. This is demonstrated on Figure 12, where for i-η values of 1000 and 100 the CYCLOPS response rejoins the baseline response at approximately 0.8% and 7% of the displacement to residual (wres ) respectively. 2.2.5 Shear stress within gap zone As noted earlier the RATZ model gives a poor fit to the stress-displacement response within the cyclic ‘gap’ zone. A new formulation was therefore implemented in CYCLOPS to address this limitation. This modification is illustrated on Figure 13 which shows curves representing τ/τfail versus w/wcyc , where w is the displacement increment relative to the edge of the gap zone. Note that a negative normalised displacement represents behaviour inside the gap zone (i.e. a w/wcyc of zero defines the gap zone edge).

Example curves are presented for b values ranging between 0.05 and 0.95. 2.2.6 Shear stress outside gap zone As discussed earlier RATZ assumes a hardwired value of b equal to 1.0, implying that the maximum bias shear stress (τw max ) developed at the edge of the gap zone is equal to the current monotonic failure stress (τfail ). However, the available CNS test and GST data do not support this assumption since it is generally found that the value of τ at the gap zone edge is less than τfail . A certain displacement must be applied beyond the edge of the gap zone for τ to recover from the value at the gap zone edge to τfail . To address this limitation the CYCLOPS model allows the user to specify the displacement over which τ recovers from τw max (as defined by Equation 4) at the edge of the gap zone to τfail . This transfer displacement is termed the cyclic transition width (wcyc ). Further examination of the available CNS test and GST data indicate that the magnitude of wcyc is correlated with the size of the gap zone; an increase in size of the gap zone is matched by an increase in wcyc but at a decreasing rate, such that wcyc eventually appears to approach a limiting maximum value. Hence the CYCLOPS model incorporates a user specified maximum cyclic transition width (wcyc-max ). The value of wcyc at any stage is then determined according to the size of the gap zone (wgap ) and the exponential relationship presented on Figure 14, which plots wgap against wcyc , all normalised by wcyc-max . The implication of this approach is that wcyc /wgap decreases from close to unity (as implied by the dashed line on Figure 14) for wgap /wcyc-max near zero to approximately 0.2 for wgap /wcyc-max = 5. The shape of the shear stress-displacement curve assumed in the CYCLOPS model immediately outside the gap zone (i.e. w/wcyc = 0 to 1.0) is illustrated on Figure 15 for various values of b ranging between 0.05 and 0.95.

132 © 2011 by Taylor & Francis Group, LLC

Figure 16. OC GST 400S2 – CYCLOPS comparison.

Figure 14. Cyclic transition width.

Figure 17. OC GST 400L – CYCLOPS comparison.

Figure 15. Shear stress within and outside gap zone.

2.3 CYCLOPS model applied to GSTs The discussion presented to date has focused on the observed response in CNS tests. However, the same characteristic features are also clearly evident from GST data. Typical CYCLOPS fits to some of the OC GST results (Randolph et al, 1996) are presented on Figure 16 and Figure 17. Similar response characteristics have also been observed elsewhere, with the recent full scale pile test results from Haberfield et al. (2010) being a notable example. Despite the ability of CYCLOPS to provide a very good match to the OC GSTs, problems arose when attempting to find a common set of parameters that could realistically model both the OC CNS tests and the monotonic and cyclic GSTs. These problems were of two basic forms, namely: – Use of the optimal set of parameters derived from the monotonic GST responses resulted in premature

133 © 2011 by Taylor & Francis Group, LLC

failure when the grouted sections were subject to cyclic loading. – The optimal monotonic and cyclic parameters derived from the CNS tests were unable to predict the actual capacity of the grouted sections, giving conservatively low results. This originates from the fact that the observed CNS test τpeak values were low compared to the GST results. Although it is possible that this represents a systematic difference between the τpeak that can be achieved in a CNS test and that which can be achieved for a GST, we believe that this is unlikely to be the case. It is suspected that there may have been some problems in the sampling or testing of the OC cores which have contributed to the observed difference, but this is subject to confirmation. In summary, despite the much more realistic modelling that is possible with CYCLOPS compared to RATZ, the available correlation with GST data still gave a conservatively low prediction of measured GST response.This is of course safe for design purposes and hence is a robust position to occupy into the future. 2.4

Scale effects

In order to try to assess whether the underprediction of the GST response using CYCLOPS might be

due to some kind of scale effect between small scale CNS tests and full scale piles, or due to some other cause, a programme of numerical analyses was conducted using FLAC (Itasca, 2005). In these analyses an axisymmetric continuum model for the pile and soil was adopted, rather than the t-z approach used in CYCLOPS. These analyses focused on comparing responses under monotonic loading for CNS tests and GSTs and specifically addressed the potential role of borehole roughness on the GST response. A particular focus of this study was whether or not a full scale pile might exhibit a thicker interface failure plane (ie. shear band) than obtained in a CNS test, which could in turn lead to a ‘stretching’ of the required displacement to mobilise the residual friction. 2.4.1 Constitutive model The standard ‘Strain Softening Plasticity Model’ provided within FLAC was used for these analyses. Prior to the onset of plastic yielding the soil response was modelled as linear elastic. The soil cementation was modelled through a cohesive strength, which degraded progressively to zero with increasing plastic strain. Following breakdown of the cementation, the soil strength reverted to a frictional material with the strength defined by a friction angle; the friction angle was also variable starting from zero prior to any plastic yielding and then increasing to a maximum value as the true cohesion degraded to zero; this ‘peak friction’ angle reflected the fact that whilst the true cementation might be gone, the soil sample was still likely to be compact and interlocked at this stage. Further plastic shearing led to a gradual reduction of the friction angle until a ‘critical state’ was achieved at larger strain, and thereafter a constant value was assumed. Another key feature of these materials is contraction when subject to plastic shearing. This was incorporated into the soil model through the adoption of a negative dilation angle; this varied from a high initial value, corresponding to the maximum rate of contraction, and thereafter gradually increased (ie. became less negative) at increasing plastic shear strains. At the ‘critical state’ a zero dilation angle was imposed. A ‘cap’ yield surface was not included since the axial pile friction in these soil types is governed by normal stresses well below the cap yield surface and hence it is the shear induced contraction, combined with the effects of cementation degradation, that should be captured for a realistic model. Whilst cap models can simulate the shear induced contraction well, most are inadequate for dealing with the brittle cementation component. 2.4.2 CNS test simulation The shear stress-displacement response predicted in the FLAC CNS test simulation, assuming a normal stiffness applicable to a 2 m diameter pile, is presented on Figure 18. The initial brittle post-peak response reflects the rapid breakdown in cementation while the gradual reduction thereafter is due to the shearing induced compaction.

134 © 2011 by Taylor & Francis Group, LLC

Figure 18. CNS test simulation – 2 m diameter pile.

2.4.3 Mesh details for GST simulation Following a program of initial analyses, the key components identified for capturing realistic GST behaviour were identified as: – Meshes needed to be vertically aligned to ensure that the dominant and known shear band orientation (ie. vertical) could be captured accurately and within a single element thickness. Borehole roughness was introduced by assigning stiff and strong ‘pile’ properties to some of the elements that were adjacent to the pile but which would ordinarily be assigned soil properties (for a ‘smooth’ pile). At certain locations on the grout-soil interface, where the shear band is forced to propagate from one radial distance to another (invariably where a distinct change in grouted section diameter occurs), a bearing type failure is initiated at the ‘lip’ of the grouted section outstand, and this would, by definition, be smeared over several soil elements in front of the ‘lip’ and would involve imposition of high normal stresses. Strain softening and localisation into a shear band is not a critical feature of such a failure mode and hence the vertical alignment of the mesh is not believed to have a significant adverse influence on these bearing type failure modes. – Very thin soil elements were used immediately adjacent to the grouted section interface since this was essential in order to capture an appropriate degree of borehole roughness. This also offered the advantage that several elements would always be involved in any local bearing type failures that developed around individual hard-points. The minimum element thickness adopted in our analysis was around 4.5 mm at the nominal grout-soil interface, increasing steadily to the radial boundary of the mesh. For the five elements in closest proximity to the groutsoil interface the average element thickness was around 6 mm to 6.5 mm, which was very comparable to the element thickness used in the CNS test simulation (5.8 mm). This similarity of thickness is vital in order to avoid spurious mesh scale effects,

Figure 20. Load displacement response – ‘rough’ and ‘smooth’ 2 m GST.

imposed on the top of the mesh to represent a burial depth of around 45 m. – The grouted section itself was modelled as an elastic solid with a resultant EA that proved to be around double that of the actual OC test sections. A Poisson’s ratio of 0.3 was also assumed.

Figure 19. Grout-soil interface details – 2 m diameter GST analyses.

since the minimum shear band thickness is limited by the element thickness which ultimately controls the displacement rate from peak to residual friction. Analyses were undertaken with a variety of different roughness assumptions – details of two of the considered cases for the 2 m diameter GST are presented on Figure 19. The meshes used for these analyses had the following key features: – The modelled GST length to diameter ratio varied from about 1.5 to 6, which covers the range for most of the actual OC GSTs. No end bearing was included below the grouted section. – The outer radial boundary of the mesh, which was fixed against radial movement, was defined at a radius of 60 times the grouted section radius. – Soil was included above and below the grouted section in order to avoid local effects, such as unrealistic stress concentrations at the base of the grouted section and under-confined mesh distortion at the top of the grouted section. – The in situ effective horizontal stress at the mid depth of the mesh was the same as assumed in the CNS test simulation. A vertical surcharge was

2.4.4 GST analysis results The shear stress-displacement responses predicted in the 2 m diameter GST simulations are presented on Figure 20 for a variety of different roughness conditions. It may be noted that for all cases the peak frictions obtained from the rough GSTs are significantly less than the simulated CNS test peak. Some of the variability in peak friction from the GSTs is due to the slightly different diameters for the differing roughness, but the enhancement for the ‘rougher’ piles was generally significantly greater than the change in diameter would imply. The reason for the lower average peak friction in the GSTs compared to that obtained in the CNS test simulation was found to be due to progressive failure; at peak mobilisation of the GST a sufficiently high strain had developed over a short section at the top and a large section at the bottom of the grouted section such that the initial cohesion had degraded to zero. A purely frictional response therefore applied over these zones, which gave a significantly smaller interface friction than implied by the initial soil cohesion. The residual frictions exhibited a different trend with higher values obtained in all of the GST cases compared to the CNS test simulation. An evaluation of the various results revealed that this was due to several reasons:

135 © 2011 by Taylor & Francis Group, LLC

– Even at low roughness, the horizontal effective stress averaged along the GST was found to be higher than the normal stress acting at the residual condition in the CNS test simulation.

When considering results from this kind of analysis due caution should be exercised since such analyses are notoriously difficult to perform reliably. Our examination of the GST simulation results did not reveal any obviously faulty features, except (in a few cases) a slightly enhanced residual friction at large head displacements resulting from some over constraint in the system. Overall, we believe that these analyses do provide some evidence for an enhanced residual friction on a full scale grouted section compared to a CNS test, although we recommend that this conclusion should be treated as provisional in the absence of additional confirmatory data. In addition it may only be applicable in certain cases such as for soils subject to shear induced contractancy, as modelled here. Other questions may be asked as to the modelling assumptions made for the shear band. An alternative strategy, adopted in some cases (usually clay soils), is to include some strain rate dependency of the soil strength, which in turn controls the minimum shear band thickness that develops. However, for the cemented soil under consideration here, it is not readily apparent why any significant viscous strain rate effects might arise. Within a continuum framework, it is not apparent that any other assumptions could be sensibly made, in the absence of empirical data. However, it is possible that a more fundamental assessment using a discrete element method (DEM) approach might potentially reveal some different types of failure mechanism (at the micro-mechanical scale) in these kinds of soil, which could suggest a scalability of shear band thickness that is not resolvable in the current work. In any event, in order to further advance our understanding of this complex matter either additional testing at the OC test locations or more full scale GST or pile load tests combined with comprehensive programs of CNS tests from suitable sites elsewhere, will be required.

– Calculation of the ‘equivalent’ CNS spring acting along the GSTs revealed that while this was generally in accordance with the CNS test simulation, strong end effects were apparent which gave rise to the enhanced average horizontal effective stress. – For the rougher interface cases, highly concentrated pressure bulbs were mobilised just above many of the significant outstands on the grout-soil interface. Further detailed inspection revealed that each pressure bulb was associated with a transition of the sliding failure plane from a larger to a smaller diameter (i.e. to become closer to the nominal diameter of the pile). These pressure bulbs were therefore associated with local bearing failures around the outstand rather than pure frictional sliding failures. At these pressure bulbs, highly concentrated horizontal effective stresses were mobilised, which offset the contraction induced reduction in horizontal effective stress that occurred in the CNS test simulation, in turn enhancing the frictional resistance. As can be seen by comparing Figure 18 and Figure 20, the initial stiffness of the simulated CNS test response was found to be very much greater than obtained from the GST simulations. This is readily understood when considering the different configurations of the CNS and GST test responses; the former only required shearing of a single 5.8 mm thick element while the applied shear stresses at the grout-soil interface of a GST propagate throughout the mesh and hence elastic deformations accumulate over a wide area. From the same figures, it can also be seen that none of the GST analyses revealed any difference in the displacement required to transition from the peak to residual friction between a CNS and full scale GST. Hence despite the imposed interface roughness in the GST simulations, the shear band thickness in these analyses was controlled by the minimum element thickness in all cases; the increasing roughness cases did nothing to enhance the shear band thickness, and hence there was no mechanism to enhance the displacement to residual. 2.4.5 Conclusion from FLAC analyses Unfortunately the results of the FLAC analyses did not reveal a definitive explanation as to why the response in a CNS test might differ significantly from that of a full scale GST. In most cases the response of the simulated CNS tests were found to be a close proxy of the numerical GST analysis. However, the analyses did hint at the potential for some differences, both due to end effects (although these might not be important for an actual pile) and due to the influence of a ‘roughened’ interface. Specifically, increasing roughness appeared to enhance the residual friction compared to an equivalent CNS test. None of the analysis results suggested that the displacement required to transition from the peak to residual friction should be any different between a CNS test and a full scale GST.

2.5

The usual process by which the design storm history is incorporated into the cyclic axial capacity analyses for drilled and grouted piles offshore Australia is as follows: 1. A structural model of the platform incorporating all dead, live and environmental loads is used to determine the total pile head axial load occurring at the peak of the maximum design wave (i.e. maximum overturning moment) and at the trough of the same wave (i.e. minimum overturning moment) for the given design condition. 2. The structural model of the jacket is then subjected to waves of various height (H) and period (T) corresponding to the design lifecycle and storm wave H-T histograms. Processing of the model output results in a pile head cyclic axial load ‘transfer function’ which relates H and T to the pile head axial load at each wave peak and trough, thereby defining the loads applicable to each cycle. 3. The pile head cyclic axial load transfer functions, together with the number of cycles of each wave

136 © 2011 by Taylor & Francis Group, LLC

Influence of storm order

type as specified in the lifecycle and storm histograms, are input into the RATZ or CYCLOPS software in some order. Historically, the lifecycle and storm wave H-T histograms were generally only defined as occurrence matrices rather than specific time histories. Therefore, the lifecycle and storm occurrence matrices were arranged to produce various ‘ideal’ time histories, including ‘increasing’ (i.e. smallest wave to largest wave), ‘decreasing’(i.e. largest wave to smallest wave) and ‘increasing-decreasing’ (i.e. smallest waves at the beginning and end, with the largest wave occurring at the storm midpoint). Generally it was found that the ‘decreasing’ arrangement would result in the highest reduction in axial pile capacity as a result of cyclic storm loading. 2.5.1 ‘Actual’ storm load time history Recently ‘actual’ wave-by-wave storm time histories have been developed for pile capacity analyses. An example of an actual storm profile, in terms of pile head axial load at the wave peak and wave trough, is presented on Figure 21. Surprisingly, it was found that the cyclic degradation inflicted on a pile subjected to an actual storm was greater than that for a pile subjected to any of the idealised storm arrangements that had been used in the past. The reasons for this were not immediately apparent but on further investigation it was found that this was related to: – The arrangement of the larger waves within the profile. – The degree to which each of the larger waves can cause zones along the pile to go post-τpeak . – The degree to which packets of smaller waves, which occur after each of the larger waves, are able to cause further damage to the post-peak zones. Fundamentally, it is a basic characteristic of both the RATZ and CYCLOPS models that the degradation process is governed by the amount of plastic degradation that occurs for soil elements that are loaded at some point during the storm such that their monotonic (peak) strength is exceeded. Cycling at loads that never reach τpeak but which exceed the elastic threshold shaft friction ξτpeak can also cause degradation, but at a much slower rate than for any element that goes post-τpeak at some stage. As an example of the shaft friction degradation process, first consider a pile subjected to an idealised decreasing storm loading profile, where the largest load is applied first and the smallest last. The process is demonstrated schematically in a step-by-step fashion on Figure 22 and described as follows: – At the start of the storm the largest wave causes the soil along the upper portion of the pile to reach its peak shaft friction and to commence along the strain-softening part of the t-z curve (i.e. the post-τpeak zone on Figure 22a). This is due to the flexibility of the pile which results in much larger pile-soil interface movements at the top of the pile

137 © 2011 by Taylor & Francis Group, LLC

Figure 21. Pile head axial load profile – actual storm.

–

–

–

–

compared to the bottom. Hence, from this point on, these soil elements operate in the post-τpeak domain, as illustrated by the resulting failure shaft friction (τfail ) profile on Figure 22b. Subsequent large waves of decreasing height will cause some additional soil elements to transfer from pre-τpeak to post-τpeak behaviour, which enables the post-τpeak zone to propagate further down the pile (Figure 22c). In addition to this, further damage is inflicted on the initial post-τpeak zone, leading to a steady reduction in the available shaft friction (i.e. τfail ) over this region (Figure 22d). It is important to note at this point that since all the large cycles are lumped together at the beginning of the storm, the total amount of post-τpeak degradation that occurs in the first few cycles is limited. Hence, as the storm progresses and the modelled wave height reduces, the ability of subsequent waves to extend the post-τpeak zone rapidly decreases. Following the first few large cycles, the subsequent very large number of medium to small cycles are only able to continue to degrade the elements that were forced post-τpeak in the first few cycles (Figure 22e). Once full degradation of the post-τpeak zone to the residual shaft friction (τres ) has occurred, which is generally after only a relatively small proportion of the total number of smaller waves in a storm, the large number of remaining small waves are unable to cause any further damage (Figure 22f). Hence, at the end of the storm, the post-τpeak zone has generally transformed almost completely into a zone where τfail = τres . However, the initial postτpeak zone caused during the first large cycle is only extended to a limited degree by the subsequent cycles.

Now consider the same pile subjected to the same storm loads but with the load profile ordered such that the larger waves are spread out over the storm’s duration and separated by packets of smaller waves. An example of such an ‘arranged’ profile is presented on

Figure 23. Pile head axial load profile – arranged storm.

–

–

Figure 22. Pile axial capacity degradation – ‘decreasing’ storm.

–

Figure 23, which has been developed from the ‘actual’ storm history presented on Figure 21. The arranged profile comprises five ‘load packets’ with each packet comprising one large wave followed by a series of smaller waves ordered according to decreasing wave peak load. The process is demonstrated schematically in a stepby-step fashion on Figure 24 and described as follows:

–

– After application of the first large wave (Figure 24a), a post-τpeak zone will form along the upper portion of the pile (Figure 24b). This is the same as for the first cycle in the decreasing storm. – A subsequent packet of smaller waves will then work on the post-τpeak zone (Figure 24c). This

138 © 2011 by Taylor & Francis Group, LLC

post-τpeak zone is not extended, but most or all of it will be transformed into a τfail = τres zone (Figure 24d). Note that once a soil element goes post-τpeak , even very small subsequent loads may propagate the degradation, because any stress that exceeds the (very small) cyclic residual friction (τcyc-res ) leads to plastic strain and hence an accumulation of plastic damage. The small waves are used much more effectively (in terms of causing damage) than in the descending storm analysis, since one large wave followed by a moderate packet of small waves may lead to a τres zone only a little smaller than caused for the entire descending storm. When the large wave from the second load packet is applied (Figure 24e), the post-τpeak zone caused by the first large cycle is unable to offer much resistance to this load, since it has been fully degraded to the τres value. Hence, this load is distributed much further down the pile than would have occurred if this were only the second wave in a descending storm history. This therefore maximises the size of the new post-τpeak zone caused by this second large cycle (Figure 24f). A subsequent package of small waves then works on this new post-τpeak zone (Figure 24g) until it has degraded to τres (Figure 24h). With subsequent ‘packets’ of a relatively large damage triggering wave followed by a set of smaller waves the process continually repeats itself, with each larger wave propagating the post-τpeak zone further down the pile and each packet of smaller waves degrading the strength in this zone towards τres .

In summary, it can be seen that by separating the larger damaging waves by sets of smaller waves, cyclic damage is maximised. In the descending storm only a relatively few of the small to medium cycles make any contribution, whereas in an ‘optimally’designed storm (for maximising damage) all the small to medium cycles will contribute to the damage process.

this model for practical design purposes a significant number of CNS tests need to be performed. For each CNS test a best fit calibration is developed, which requires considerable skill and care, since as mentioned in Section 2.2.1, superficially similar monotonic responses over a limited strain range can result in a very different cyclic response. A variety of stress and displacement controlled CNS tests are required at each tested location in order to properly anchor the calibrated CYCLOPS response. Any excessive brittleness observed in the monotonic CNS test stress strain response also needs to be treated with extreme caution since this may never be apparent in a field situation. A well calibrated view on sensible ranges for the peak skin friction is required to ensure that an unconservative parameter selection is not made. Once the CNS test calibrations are complete the values of the various parameters adopted in each CYCLOPS fit are then plotted as a function of depth in order to ascertain appropriate trends. Design lines can be fitted through the calibrated data as desired. Our experience with this procedure so far has been good, with generally sensible ranges of parameter variability found from these calibration exercises. As more experience is obtained using the CYCLOPS model over time it is hoped that some of the required parameters may generally be found to lie within tight bounds, which may reduce the number of detailed calibrations that are required at each site. Notwithstanding, it is likely that design of drilled and grouted piles under complex cyclic loading conditions will remain a challenging task for many years to come. 2.7

Figure 24. Pile axial capacity degradation – ‘arranged’ storm.

2.6 Assessment of parameters for design The CYCLOPS model outlined in this paper can accurately model measured CNS test and GST responses, but this ability comes at a cost in that many different parameters need to be defined. In order to use

Results of an example analysis are presented here to demonstrate the application of CYCLOPS to a typical full size pile, and are compared to the equivalent results obtained using the standard RATZ algorithm. A steel pile has been adopted for this example with a 2.2 m outer diameter, a constant 90 mm wall thickness and it is assumed to be grouted in a 2.5 m diameter hole. The soil parameters used in the analysis are presented in Table 1. The assumed storm load history has been applied in various ways to illustrate how important this can be: the full actual storm (Figure 21), a ‘5 packet’ storm (Figure 23), a standard decreasing storm and an increasing storm. The pile head load-displacement response predicted using CYCLOPS for the ‘actual’ storm is presented on Figure 25, along with the undegraded monotonic response. It should be noted that the somewhat spikey appearance of the cyclic response, particularly for isolated large cycles, is due to the fact that only the peaks (minimum and maximum) of each cycle are saved in the analysis for later post processing. The t-z curves extracted from the actual storm analysis at a couple of nominal depths (9.8 m and 39.8 m) are presented on Figure 26. The extent of cyclic degradation caused by the storm is readily apparent from

139 © 2011 by Taylor & Francis Group, LLC

Example analysis

Table 1. analysis.

CYCLOPS & RATZ parameters for example

Parameter

CYCLOPS

RATZ

τpeak τres /τpeak Gmax ξ wres η i-η τcyc /τpeak bi bf m  wcyc-max

440 kPa 0.3 1232 MPa 0.4 0.2 m 0.4 300 0.01 0.5 0.2 5 0.1 m

440 kPa 0.3 1232 MPa 0.4 0.2 m 0.4 N/A 0.01 N/A N/A N/A N/A

Figure 25. CYCLOPS pile head response – actual storm.

these figures, leading to an 18.4% reduction in the post-storm monotonic capacity. From the t-z curves it is apparent that the response is affected by cyclic degradation, even at 39.8 m below the mudline. This is further illustrated on Figure 27, which presents the peak mobilisable friction at the end of the storm for each of the different representations of the storm. It can be seen that the actual storm causes the greatest degradation, and is slightly worse than the ‘5 packet’ storm. The increasing storm causes little damage (0.8%), but it can still be observed from Figure 27 that the peak mobilisable friction is less than the prescribed peak friction over the upper 20 m. However, this reduction is principally due to monotonic progressive failure not cyclic loading. The standard decreasing storm that has generally been used for design purposes in the past is shown to be significantly less damaging (10.2%) than either the actual storm or the ‘5 packet’ storm. The key feature evident from Figure 27 is the propagation of damage to increasing depth with increasing disorder in the storm, as anticipated by the process described in Section 2.5. The pile head load-displacement response for the equivalent RATZ analysis of the actual storm is presented on Figure 28. The cyclic degradation (15.0%) is only a little less than obtained with CYCLOPS for this particular case. The RATZ generated t-z curves from the actual storm are shown on Figure 29. By comparison with Figure 26, the more stylised t-z response obtained from RATZ at a depth of 9.8 m is readily apparent. Also it can be seen that cyclic degradation did not propagate as far as 39.8 m in the RATZ model although this did occur with CYCLOPS. The peak mobilisable friction at the end of the storm is presented for the various RATZ analyses on Figure 30. Interestingly, the cyclic degradation obtained with RATZ for an increasing storm was slightly higher, albeit still very small (1.4%) compared to CYCLOPS. However, for all the other storm representations, CYCLOPS gave slightly higher cyclic degradation than RATZ. In summary it can be seen that for the example presented here CYCLOPS gave slightly lower

Figure 26. CYCLOPS t-z response – actual storm.

post-storm capacities than RATZ. However, we have found much larger differences in other practical cases, particularly where complex stratigraphy is encountered. The general rule to date is that CYCLOPS invariably yields lower post-cyclic capacities than obtained with RATZ, all other things being equal. The importance of modelling actual storm histories rather than idealised representations is also clearly demonstrated by both the RATZ and CYCLOPS analysis results. 3 PART 2 – LATERAL PILE DESIGN 3.1

A new model for determining the lateral response of piles in uncemented carbonate soils when subject to undrained cyclic loading is presented herein. The new

140 © 2011 by Taylor & Francis Group, LLC

Introduction

Figure 29. RATZ t-z response – actual storm. Figure 27. CYCLOPS post-cyclic failure shaft friction profiles.

Figure 28. RATZ pile head response – actual storm.

method retains the basic p-y curve format used as standard for design of laterally loaded piles in the offshore industry but builds on the design approaches used for shallow foundations and suction piles. Design of such structures is now generally performed using a ‘cyclic strength approach’. For these kinds of structures the standard practice is to define a ‘cyclic strength’, which is some reduced proportion of the monotonic strength, and is a function of the number of applied cycles, the degree of cyclic loading (2-way, 1-way, etc.) and the magnitude of strain that is deemed acceptable for the soil to sustain.Appropriate factors of safety are applied to these cyclic strengths and the combined bearing and sliding capacity is then determined in order to define

141 © 2011 by Taylor & Francis Group, LLC

Figure 30. RATZ post-cyclic failure shaft friction profiles.

the maximum loads that may be safely applied to the foundation, with only limited in-service movements. Unfortunately, modelling of laterally loaded pile behaviour is more complex than design of shallow foundations or suction piles since the ultimate geotechnical capacity is generally not the governing design criteria. For a typical long and flexible pile, an increasing applied pile load will simply mobilise the soil over a greater depth and hence it is the pile bending

moments and pile head displacements that are usually the critical design parameters. To assess these it is therefore necessary to determine the appropriate load displacement response at each depth level in the soil. In addition, whilst shallow foundations and suction piles are usually pure load controlled systems (i.e. applied loads do not generally redistribute in a global sense) the upper part of a laterally loaded pile when subject to cyclic loads will exhibit a partially displacement controlled response, leading to the potential to redistribute soil reaction pressures down the pile as cycling progresses. In order to capture these features we believe that it is first necessary to define the appropriate monotonic p-y response and then to perform a cycle-by cycle analysis to capture the progression of the response over time. 3.2

Model development

Development of the new p-y model comprised four basic stages: – Stage 1: Finite element type analyses were performed to establish the p-y response of a pilesoil slice element for a range of generic stress strain curves representative of uncemented carbonate soils.These were developed for various numbers of applied load cycles. Analyses were also performed to assess the modification required to allow for near-surface effects. – Stage 2: The results from Stage 1 were used to develop an appropriate p-y curve format that can be used for any specific soil strength assumption and required number of cycles. – Stage 3: Using the results from Stage 2 along with an examination of the characteristic response of raw simple shear test data, an algorithm was developed to allow a cycle by cycle analysis to be advanced. – Stage 4: The p-y curve format and the cycle-bycycle algorithm were implemented into a new Excel program, which has been named pCyCOS. To aid in comprehension of the new model, the basic algorithm for the new p-y model is first presented followed by the detailed analyses that were performed to calibrate the various required parameters. 3.3 Envelope p-y response The envelope p-y response is the first basic building block of the p-y model and defines the p-y response of a horizontal slice through the pile-soil system when subjected to load controlled cycling for various numbers of cycles. For any element of soil with a given normalised
monotonic undrained shear strength (su /σvo , where su
is the monotonic undrained shear strength and σvo is the initial vertical effective stress), an S-N curve may be defined, which is a representation of the normalised
cyclic stress ratio (S = τcyc /σvo ) and the number of cycles (N) to mobilise a nominal shear strain level (γ).

142 © 2011 by Taylor & Francis Group, LLC

Figure 31. Envelope p-y response – basic concepts.

By drawing vertical slices through the S-N curve it is possible to define stress strain curves for any given value of N. This is illustrated on Figure 31a and Figure 31b. By implementing these stress strain curves into a finite element type model of a lateral pile element we can then develop the appropriate p-y curve that applies for the specified number of cycles, N, as illustrated on Figure 31c. It may be noted that a normalised pressure


response is defined from this assessment (p/σvo where p is the soil pressure applied to the pile element).
the process described For the specified su /σvo above is repeated for different values of N and the results compiled, as shown on Figure 31d. This latter plot essentially defines the response of a lateral soil element when subject to load controlled cycling. For example, considering a pile element that imposes a constant peak cyclic pressure p1 on the soil, the lateral displacement for that pile element would be equal to y1 after the first cycle, y2 after the second cycle, y3 after the third cycle and so on (Figure 31d). The entire process outlined above is repeated for
various values of su /σvo , as required, in order to cover the range of monotonic strengths expected in the field. We have conservatively adopted pure 2-way cyclic loading conditions in the model to date, and hence 2-way S-N data has been used to define the necessary stress strain curves for developing the envelope p-y curves. The model may be readily changed for any other degree of cycling, but would require additional assessment of the various input parameters.

3.4 Cycle by cycle p-y response The envelope p-y curves that define the pile-soil behaviour under pure load controlled cyclic conditions have been defined above. If the problem under consideration were purely load controlled, then the final model would require no further sophistication than to add together all of the load controlled element responses for any given number of cycles, N, in order to give the total response for the entire system after N cycles. However, as noted earlier, the (partially) displacement controlled nature of the pile-soil system means that pile-soil elements near the top will shed load to elements further down, as the upper elements, which are subject to larger cyclic displacement amplitudes, ‘soften’ to a much greater degree than the lower elements. A cycle by cycle model is therefore required to capture this behaviour. 3.4.1 First cycle The lateral soil response used for the first cycle is simply the p-y envelope response for N = 1 as discussed in Section 3.3. By definition this describes the monotonic loading response of a pile-soil element. Monotonic (first cycle loading) is a purely load controlled process and therefore for any given pressure, p1 , applied during the first cycle, the lateral displacement for this element will be y1 (Figure 32a). 3.4.2 Second cycle The p-y response assumed for the second cycle is defined from the envelope p-y response obtained for N = 2, but is not equal to the N = 2 envelope curve. For any given pressure, p1 , applied during the first cycle the lateral displacement obtained during the second

143 © 2011 by Taylor & Francis Group, LLC

Figure 32. p-y curve format for first and second cycles.

cycle, assuming that the same pressure p1 was still applied, would be equal to y2 (Figure 32b). However, in reloading to the pressure p1 during the second cycle, the pile-soil element is assumed to follow a different

pressure–displacement curve. This reloading curve is formed from 3 parts: Part 1 – Up to an applied pressure of 25% of p1 , the pile-soil element follows a linear path starting from the origin and giving a displacement of y0.25 at P0.25 (=25% of p1 ). (Evaluation of y0.25 is considered later.) Part 2 – For a pressure exceeding 25 % of p1 , the pile-soil element follows a second linear path that passes through the point (p0.25 , y0.25 ) and through the point (p1 , y2 ). Part 3 – Where the pressure defined from Part 2 of the p-y curve exceeds that which would be obtained using the N = 1 (monotonic) envelope curve, then the monotonic envelope curve is adopted. It can be seen that a key feature of this second cycle p-y response is an initial ‘soft’ zone followed by a hardening response. Such behaviour is encountered in all simple shear tests on uncemented carbonate soils where a reversal of the cyclic shear stress is imposed. The selection of an inflection point at 25% of p1 has been made based on inspection of the results from many simple shear tests. It may also be noted that the monotonic response may (eventually) be recovered, partly due to the dilatory soil response as large shear strains are applied following partial liquefaction. This is consistent with the philosophy that the generation of excess pore pressure during cyclic loading mainly affects the soil stiffness rather than its ultimate strength. However, it should be also appreciated that when a load cycle imposes a soil pressure that is close to the ultimate limit defined for cycle N, the degree of softening that occurs for cycle N + 1 may be such that there is a complete loss of stiffness under any further cyclic loads (Figure 32c). The pile will therefore be free to move through the softened soil, but would eventually encounter fresh soil, which would constrain the ultimate displacement somewhat. Large displacement hardening of this type has not been included in the model, but in practical applications it is unlikely to be an important limitation, due to the partially displacement controlled nature of lateral pile loading, since this also serves to limit the increase in displacement from one cycle to the next. 3.4.3 Subsequent cycles The form of the p-y curve used for all subsequent cycles (N = i) depends on whether or not the peak pressure applied to the pile-soil element in cycle N = i − 1 is the maximum pressure that has ever been applied to that element (pmax ), or is less than pmax . Where the peak pressure applied in cycle N = i − 1 (i.e. pi-1 ) is equal to pmax , the p-y curve for the next cycle is defined using the same procedure as used for the second cycle (Section 3.4.2), as illustrated on Figure 33a. This approach includes a degree of oversoftening in the p-y response, since inherently it is assumed that the pile has been subject to i − 1 cycles of pressure, pi-1 , even though it may have been subject to

Figure 33. p-y curve format for subsequent cycles.

a lower pressure during the earlier cycles, which would be expected to lead to a lower degree of softening for cycle N = i. Where the pressure applied in cycle N = i − 1 is less than pmax , two alternative criteria are used to assess the p-y response for cycle N = i, with the criterion that gives the greatest degree of softening adopted for the final p-y response. The first criterion is identical to that presented on Figure 33a: the displacement for cycle N = i is calculated assuming that the pile has been subject to i − 1 cycles of pressure, pi-1 , even though it may have been subject to a greater pressure during the earlier cycles. Inherently this means that the predicted degree of softening will underpredict that which would occur in practice, and indeed if pi-1 is significantly less than the maximum pressure ever applied, it can lead to a p-y curve for cycle N = i that is stiffer than that used for cycle N = i − 1, which is clearly a physically unreasonable outcome. Hence the second criterion is to first determine the p-y response that would be obtained if the pile-soil element had been subject to i − 1 cycles with an applied pressure equal to pmax , (i.e. as per Figure 33a), and then apply a scale back factor to the predicted degree of

144 © 2011 by Taylor & Francis Group, LLC

softening obtained from this calculation to account for the fact that the pile-soil element is now being cycled at a pressure that is less than pmax . The scale back factor used in this calculation has been derived by consideration of the available S-N data and will be discussed later. This process is illustrated on Figure 33b. 3.4.4 Modifications for surface effects The p-y curves discussed above are defined based on a horizontal slice through the pile and soil. This therefore represents the ‘deep’ failure mode where the soil can flow around the pile but with no vertical movement of the soil. However, near the mudline, vertical movements can occur in the soil giving rise to wedge type failures. To account for this effect a scaling factor (‘p-multiplier’) is applied to the pressures calculated using the ‘deep’ failure p-y curves. The scaling factor is determined as the ratio between the ultimate lateral resistance calculated accounting for the actual embedment depth, to the ultimate lateral resistance that would be obtained for the ‘deep’ failure mode. The ultimate peak lateral resistance has been defined at any depth based on the plasticity model developed by Murff and Hamilton (1993). The validity of this assumption was confirmed through a program of 3D finite element (FE) analyses, using the models discussed in the following section. These analyses showed that the simple ‘p-multiplier’ approach proposed here generally gave results within 10% of those obtained from the more rigorous FE analyses, which is considered acceptable for engineering design purposes. 3.5 Analysis of envelope p-y response The envelope p-y curves required for the new p-y algorithm have been developed from FE analyses performed with ABAQUS (ABAQUS Inc., 2006) and finite difference (FD) analyses performed using FLAC (Itasca, 2005). We used two programs for this work, since they have different strengths and weaknesses. In addition, a few cases were analysed using both programs to enable internal verification of the results obtained. Irrespective of the specific program used, the meshes generated for developing the envelope p-y curves are based on the same basic assumptions: – A plane strain ‘slice’ through the pile and soil was modelled. – Infinite elastic elements were included to model the far field behaviour; these were found to be essential during the equivalent work performed to determine the p-y curves for the CHIPPER model (Erbrich, 2004). – The pile was modelled as having a ‘smooth’ (i.e. frictionless) interface with the soil. – It was assumed that no separation would occur between the pile and the soil, due to the presumed undrained nature of the applied loading. – A small strain formulation was generally adopted (although we also performed some confirmatory analyses using a large strain formulation).

3.5.1 Constitutive model All of the numerical analyses were performed using an undrained total stress approach. For all cases we selected the Mohr-Coulomb soil model, which degenerates to a simple Tresca model when a friction angle of zero is specified, as required for the assumed undrained conditions. To account for the very high degree of non-linearity in the stress-strain response, the cohesive strength was defined with a hardening response as a function of the plastic shear strain. The elastic part of the response was modelled using a simple linear elasticity model, with a Poisson’s ratio of approximately 0.5 to ensure constant volume conditions. 3.5.2 Stress strain response Stress-strain curves characteristic of uncemented carbonate soils were established based on the available test data. From these tests, a variety of stress strain curves were developed for use in the FE/ FD analyses which reflected the full range of different normalised
monotonic strengths, su /σvo . For each case, we used the procedure presented in Section 3.3 in order to develop the shear stress versus shear strain curves required as input for the numerical analysis. These curves were defined for pure monotonic loading and for pure 2-way cyclic loading at various numbers of cycles. Typical stress-strain curves for a) soft/loose
= 0.25 and b) dense soil with soil with a low su /σvo
a high su /σvo = 10 are presented on Figure 34a and Figure 34b respectively. We performed verification analyses in both FLAC and ABAQUS under simple shear loading conditions in order to confirm that the intended stress-strain curves were indeed modelled accurately. Figure 35 presents one such comparison, which confirms that the model operated as expected. 3.5.3 Envelope p-y analysis – results Example analyses showing the resulting p-y curves obtained from the FE/ FD analyses using the input stress-strain curves shown on Figure 34 for a)
soft/loose soil with a low su /σvo = 0.25 and b) dense
soil with a high su /σvo = 10 are given on Figure 36a and Figure 36b respectively. These p-y curves are presented for a number of different applied load cycles. Similar curves were developed for a range of differ
ent su /σvo values, encompassing the relevant range. From the range of results obtained, it was found to be possible to interpolate linearly between the values obtained in order to determine the appropriate
p-y curves for any other desired combination of su /σvo and N.

145 © 2011 by Taylor & Francis Group, LLC

3.6

Parameters for cyclic p-y response

In addition to the envelope p-y response, the two other essential components of the new p-y model are the definition of the inflection point and the stiffness degradation function to account for reducing loads from one cycle to the next.

Figure 34. Typical envelope stress-strain responses. Figure 36. Computed envelope p-y curves.

(i.e. ‘stiffening’) load displacement path. The basic form for this p-y curve was obtained from an evaluation of the stress strain curves observed in cyclic simple shear tests performed on uncemented carbonate soil samples, where N > 1. Typical responses from these tests indicate that the convex nature of the stressstrain response in the first cycle contrasts with the concave response for all subsequent cycles (Figure 37). Furthermore, the test data implies a clear trend for γ0.25 (the shear strain at 25% of the maximum shear stress) to increase with increasing γcyc (the maximum cyclic shear strain in each cycle) and a conservative best fit to the data may be expressed as:

Figure 35. Comparison of stress-strain response.

3.6.1 Inflection point As discussed in Section 3.4.2 all load cycles after the first load cycle have a three part p-y format, with the first two parts describing a linear or concave upwards

Having obtained a relationship between γ0.25 and γcyc , it was necessary to define an equivalent relationship between y0.25 and yapp , where yapp is any given lateral movement of the pile. To assess this, some postprocessing was performed of the various horizontal slice FE/ FD analyses discussed earlier. This postprocessing took the form of evaluating the shear strain

146 © 2011 by Taylor & Francis Group, LLC

Figure 37. Detailed stress-strain responses.

(i.e. γcyc ) obtained at specific load levels throughout the FE/FD meshes used for these analyses, assessing γ0.25 at each element from Equation (6), and then determining appropriate weighted average values for γcyc and γ0.25 . It was then assumed that y0.25 can be calculated as a function of yapp as:

Figure 38. p-y inflection point data.

Various methods were tried in order to assess appropriate weighted values for (γ0.25 /γcyc )ave and it was found that the best approach, in terms of overall consistency and minimising the dependence of the final result on the size of the mesh used in the numerical analysis, was as follows:

where: Ne = the number of elements in the FE/FD mesh. dA = the area of element i in the FE/FD mesh. γ0.25 and γcyc are the relevant strains for element i in the FE mesh. This method gives greater weight to the values of γ0.25 /γrmcyc obtained in parts of the mesh that experience higher cyclic shear strains compared to those regions that are more distant from the pile and therefore experience lower cyclic shear strains. The weighting by γcyc dA was selected intuitively, but it would also be considered theoretically robust if a linear trend could be demonstrated between yapp and the integral of γcyc dA over the whole mesh. This was there
fore checked for a variety of cases (su /σvo = 0.35 to 1.5 and N = 1 to 30) and it was found that an approximately linear response was obtained in all cases, with no more than an 8% (and generally a lot less) deviation from a linear relationship, which is acceptable for practical engineering purposes. The minimal mesh dependency obtained with this approach is illustrated on Figure 38 which presents y0.25 /D versus yapp /D for four test cases, where D is the pile diameter. These results were obtained from four horizontal slice analyses, all of which have meshes that contain an equal number of elements but with the outer boundary steadily increased (i.e. increasing outer

Figure 39. Model tests; gap width vs. cyclic displacement.

diameter (OD) to inner diameter (ID) ratio). Very similar results were obtained irrespective of the magnitude
or N, which allowed for a single trendline to of su /σvo be fitted and incorporated into the p-y model, as also shown on Figure 38. It is interesting to compare the results presented on Figure 38 with the results of a series of laterally loaded model pile tests in carbonate soil undertaken at the University of Sydney (Randolph et al., 1988). Under 2-way loading these tests revealed a pronounced ‘S’ shaped hysteresis curve, with low lateral resistance obtained in the central region, which was assumed to be associated with a gap zone or softened soil. The size of this ‘gap zone’ (ug ) was measured and, as shown on Figure 39, was plotted in an identical format to that for y0.25 on Figure 38. Comparing these two figures it may be seen that the design line for y0.25 follows the same trend as the various lines presented for ug , but with a slightly steeper gradient.

147 © 2011 by Taylor & Francis Group, LLC

3.6.2 Stiffness degradation for reducing loads As discussed in Section 3.4.3, for N > 2 two criteria are used to determine the p-y response when the lateral pressure applied during the previous cycle (pi-1 ) is less than the maximum pressure that has ever been applied, pmax . The second of these criteria first determines the p-y response that would be obtained if the pile-soil element had been subject to i − 1 cycles with an applied pressure equal to pmax , and then imposes a scale back factor to the predicted degree of softening obtained from this calculation to account for the fact that the pile-soil element is now being cycled at a pressure, pi-1 , that is less than pmax . The approach that we have used to assess the magnitude of the scale back factor is to consider the relationship between normalised stress (or strength) and the corresponding equivalent amount of damage (i.e. incremental cyclic shear strain) at varying levels of cyclic strain. The initial step in this assessment was to examine the available data from 2-way strain controlled cyclic simple shear tests performed on uncemented carbonate soil samples. The results of these tests were assessed in terms of the normalised stress ratio τ/τref (where τref is the peak shear stress attained during the first cycle) versus the stiffness degradation ratio Kdeg at selected cycle numbers during each test. The value of Kdeg is in turn defined as γ/γref (where γref is the change in γ between the first and second cycles, while γ is change in γ between any other two cycles). For displacement controlled tests the strain does not change between one cycle and the next, but the stiffness does and hence it is still possible to ascertain γ and γref in an indirect manner as shown on Figure 40. Curves were fitted to the τ/τref versus Kdeg response for each test according to the following relationship:

where the exponent n was found (empirically) to vary as a function of the cyclic shear strain. The experimental results have demonstrated an appropriate form for Kdeg in terms of stresses and strains. However, for application to the lateral pile analysis, we need to transform these results into an equivalent format but with stress replaced by the lateral pressure acting on the pile, and with strain replaced by pile lateral movement. For this exercise we have assumed that pressure may be directly substituted for stress; i.e. P/Ppeak acting on the pile is assumed to be directly proportional to the values of τ/τref acting throughout the soil. We have then interrogated the various FE/FD analyses in order to determine a value for Kdeg-ave , which is the average stiffness degradation factor value applicable to a pile rather than a specific value that applies for a single soil element. For specified values of P/Ppeak , the appropriate values of γcyc and τ/τref are determined on an element by element basis from the FE/ FD analysis results, with γcyc first being used to determine the appropriate value for the exponent n,

Figure 40. Calculation of stiffness degradation from strain controlled tests.

which is then inserted along with τ/τref into Equation (9) in order to determine element specific values of Kdeg . The weighted average value of Kdeg-ave is then obtained following a similar approach to that used for assessing y0.25 , as discussed in the previous section:

Using this approach it was found that the best-fit relationships obtained for any specific value of yapp /D varied by only a modest degree for different soil types and cycle numbers, but there was a systematic variation with varying yapp /D. This is demonstrated on Figure 41, which presents the best fit curves of the form described by Equation (9) for varying yapp /D but with τ/τref replaced by P/Ppeak . As might be expected it can be seen that the stiffness degradation factor is higher (ie. more degradation) when the cyclic displacement amplitude is large. With Equation (9) recast in terms of P/Ppeak , unique values for the exponent n were then determined as a function of yapp /D. This simple equation finalises the determination of the p-y response when the lateral pressure applied during the previous cycle (pi-1 ) is less than the maximum pressure that has ever been applied, pmax . 3.7

A computer program named pCyCOS has been developed to incorporate the new p-y model that is presented here. This program is based on the platform that was developed for the CHIPPER program (Erbrich, 2004) since the new algorithm has many structural similarities to the CHIPPER algorithm. It was therefore

148 © 2011 by Taylor & Francis Group, LLC

Description of the pCyCOS program

– Pile geometry parameters (diameter, wall thickness, length, stickup, Young’s modulus). – Pile loads (various combinations of lateral load, head moment, lateral head displacement and pile head rotation are permitted). – Whether or not monotonic or cyclic loading is to be considered, and if the latter, how many cycles are to be applied. Post-cyclic monotonic loading can also be specified. Output obtained from pCyCOS includes full information along the pile length after the required number of cycles has been applied (displacement, bending moment, shear force and reaction pressure), and p-y data pairs along the pile. The latter data can be compiled from various analyses at different load levels (with a constant N) to give p-y curves that can be used as input into structural analysis packages such as SACS or SESAM. It should be appreciated that a single unique p-y curve does not exist for any particular depth in the soil, even for a constant N, since it is a function of the pile head fixity. Hence it is necessary to perform full pile analyses for each particular case, and an iterative approach will be required with the structural designer in order to ensure that the finally defined p-y curves are consistent with the pile head fixity arising out of the structural model.

Figure 41. Stiffness degradation for reducing cyclic loads.

relatively straightforward to incorporate the new p-y model into CHIPPER, while retaining much of the pre-existing code. The existing CHIPPER code has been extensively tested and verified and hence this approach also helped to minimise the risk of bugs being introduced. It should be understood that the new program does not perform a full time domain analysis, such as that used in the CYCLOPS program for axial pile design discussed in Part 1 of this paper. Cyclic loading in the pCyCOS program is assessed on a peak-to-peak basis, meaning that the lateral response is only calculated at the peak of each cycle. This approach is consistent with that generally used to determine the undrained cyclic bearing capacity of shallow foundations and suction caissons over the last 30 years (Hoeg, 1976). In principle, analyses of full storm histories could be performed but in practice there are several important reasons why this is not recommended. The most important of these is that the assumptions made about the p-y form are really only appropriate when considering relatively few cycles. The appropriate equivalent number of cycles should therefore be calculated from the full design storm history data using strain accumulation methods similar to those used for these other applications. The input parameters required for pCyCOS include: – Profile of monotonic undrained shear strength, su , as a function of depth and the effective unit weight of the soil, γ
.
– Envelope p-y curves for various values of su /σvo , sufficient to encompass the full range of monotonic strength ratios encountered in the design profile, and for various numbers of cycles between N = 1 to N = ne + 2 where ne is the number of equivalent cycles that need to be modelled. Linear interpolation is used to assess any intermediate values of
su /σvo and N from the input data set. – Parameters to define the stiffness degradation function. – Parameters to define the p-y curve inflection point.

3.8

Two examples are presented here to illustrate the performance of the pCyCOS model. It is unfortunate that there are no suitable prototype scale lateral load tests in uncemented carbonate soil which can be used to validate the model presented in this paper. The limited available historical data are generally for fully drained conditions, which are quite different to the undrained conditions that are applicable for the typical large diameter piles used in offshore construction, even for quite sandy conditions (albeit the equivalent number of cycles for these more sandy soils may be only 1 to 3). The first example presented demonstrates the key characteristics of the cyclic degradation model, and are compared qualitatively with the results of some centrifuge model tests. The second example considers a ‘typical’ offshore pile, and the results from pCyCOS are compared to those predicted with other commonly used methods. 3.8.1 Example 1 The first example illustrates the global stiffness response of a single rigid pile subject to 2-way cyclic loading with fixed displacement amplitude. A rigid 2.83 m diameter by 20 m long pile subject to pure translation is considered for this case. A monotonic undrained strength profile that varies linearly with
depth has been assumed, defined as su = 0.5 σvo . The response obtained for this pile in terms of pile head load versus number of applied load cycles is illustrated on Figure 42. Results are presented for a

149 © 2011 by Taylor & Francis Group, LLC

Example applications

Figure 42. Example 1 – Displacement controlled cycling.

range of pile head displacement amplitudes, varying between 50 mm and 3 m. At the smallest displacement amplitudes it can be seen that the pile head load in the first cycle is modest, but this load hardly degrades with continuing cycles. As the cyclic amplitude increases, the pile head load in the first cycle steadily increases but increasing levels of degradation occur during subsequent cycles. After 10 cycles, the highest head load that can still be sustained is found to be for a 200 mm head displacement, albeit the stiffness has degraded by around 40% from that obtained during the first cycle. As the pile head amplitude is increased even further, the 10 cycle response gradually degrades to a very low value; at 3 m head displacement over 5 MN can be applied during the first cycle but not much more than 100 kN during the 10th cycle. This sort of response is entirely consistent with expectation and implies that the soil is liquefied by the very large stresses imposed during the first cycle. The dramatic degradation of soil stiffness under cyclic loading, as simulated by the pCyCOS model, has also been demonstrated in the laboratory as shown on Figure 43. This figure presents p-y curves extracted from two tests (numbered 1111 and 1131) which are extracted from Dyson (1999). These were obtained from centrifuge model tests of laterally loaded piles in carbonate silt. The p-y curves are neither constant load nor constant displacement amplitude but still show the same mechanisms as revealed on Figure 42. For example, comparing the results at the shallowest depth of 0.5 pile diameters and the deepest depth of 8 diameters it may be noted that that a much higher p/su ratio was mobilised in the first cycle at the shallow depth than at the deep depth. However, after several load cycles, the resistance of the upper soil has degraded to almost nothing (disregarding a small negative offset, which is assumed to be an interpretation/ measurement error), whereas a reasonable resistance is still obtained (albeit reduced from the first cycle) at the deepest depth.

Figure 43. Typical p-y response in carbonate silts (centrifuge).

3.8.2 Example 2 The second example represents a typical large diameter offshore pile, 2.65 m diameter by 75 m long, with a wall thickness varying from 110 mm over the top 20 m, to 45 mm below a depth of 30 m, and 85 mm in between. We have performed analyses using pCyCOS, the API soft clay p-y criteria and the method proposed in Novello (1999). Lateral pile analyses were performed for the extreme cases of fully fixed or fully free pile heads and we have considered both monotonic and cyclic loading. The API soft clay procedure has often been considered for many of the silty and muddy carbonate soils found offshore Australia, particularly under monotonic loading where the complications of liquefaction under cyclic loading do not apply. The monotonic undrained shear strength profile used in both the pCyCOS andAPI analyses is presented on Figure 44, which is representative of a typical uncemented carbonate silt at most depths, interspersed with a few stronger and

150 © 2011 by Taylor & Francis Group, LLC

Figure 44. Undrained strength for example analysis.

more sandy layers. The API cyclic method is not tied to any specific number of cycles and simply represents an envelope of ‘residual’ cyclic response for clay. However, the pCyCOS response is a direct function of the number of applied load cycles. Hence for this example we have adopted 20 maximum load cycles, which is a typical value to represent a design storm in fine grained soil, where fully undrained conditions can be expected over the duration of a storm. The other parameters required for the API analysis are ε50 and J, for which values of 1% and 0.5 were adopted respectively. As discussed in Section 1.2 the Novello (1999) procedure is formulated in terms of cone resistance, qc , rather than undrained shear strength and hence it is necessary to derive an appropriate profile of qc for the analysis. For the current example we have adopted a cone factor, Nkt , of 12 which is a typical value for undrained carbonate silts. The required profile of qc is then simply determined as 12 times the undrained shear strength. The Novello procedure nominally incorporates cyclic degradation in accordance with Equation (3). It is suggested in Novello (1999) that by adopting a typical K0 of 0.4 for uncemented carbonate soil, Equation (3) can be simplified to pcyclic = (1-0.6U∗ )pstatic . A complication with this approach is that U∗ needs to be determined from a completely separate analysis, and hence retaining internal consistency is rather difficult. We have therefore not performed an explicit cyclic analysis with the Novello procedure. However, it is still important to note that, by definition, the Novello procedure implies a maximum softening of only 60%
in a fully liquefied soil (ie. u/σvo = 1), which is not consistent with available cyclic laboratory test data on carbonate soils, nor with the centrifuge tests discussed earlier (Dyson, 1999), and nor with the predictions of pCyCOS, as demonstrated by Example 1. The pile head load displacement responses obtained from the various cases considered are presented on Figure 45. The maximum bending moments at the pile

Figure 45. Comparison of pile head lateral deflection.

head for the fixed head cases are summarised on Figure 46, while the maximum bending moments within the body of the soil are summarised for both fixed and free pile heads on Figure 47. It may be noted that the pCyCOS and API results are remarkably similar for monotonic loading, with the pCyCOS response being marginally stiffer at low load levels and marginally softer at higher load levels. The pile bending moments also exhibit the same trend. This is surprising and probably coincidental given that the API approach is essentially a purely empirical method derived from load tests on soft clays, whereas pCyCOS is a theoretically based model, with the p-y curves determined from FE/ FD analyses using laboratory measured stress-strain curves from carbonate soils. We do not anticipate that such good agreement would always be obtained; for example in denser mate
rials with higher su /σvo ratios than considered in this example. The Novello (1999) procedure leads to a generally softer monotonic response than obtained from the other two methods. This is not entirely surprising since drained lateral pile load tests are the empirical basis for the method, and the high compressibility of the carbonate soils in those tests has led to a significantly softer

151 © 2011 by Taylor & Francis Group, LLC

Figure 46. Pile head bending moments.

response than would be expected under incompressible undrained conditions. Hence while the Novello procedure purports to be able to address undrained loading, with cyclic softening also included, we consider that the method, and all similar methods, are only applicable where fully drained loading conditions are expected to apply. Whereas close correlation was obtained between the pCyCOS and API methods for monotonic loading, a rather different result is obtained under cyclic loading. Except at the lowest load levels, a much softer response and significantly higher bending moments are obtained with pCyCOS compared to the API method. However, this is not unexpected since normal clays do not exhibit the significant cyclic degradation (‘liquefaction’) that occurs with carbonate soils. The softening behaviour can be clearly seen on Figure 48, which compares the pCyCOS reaction force distributions along the pile at 1 and 20 applied load cycles for a fixed head pile subject to a 12 MN head load. It can be seen that as the upper part of the soil softens, the load is shed down the pile. This inevitably increases the pile deflections and the pile bending moment as shown on Figure 49. The change in reaction force is also illustrated on a cycle by cycle basis on Figure 48, at three points on the pile.

Figure 47. Maximum bending moments below mudline.

Figure 48. Reaction forces.

The applied reaction pressure in each cycle is normalised by the monotonic ultimate resistance, pu , applicable at each specific depth on this figure. Once again it can be seen that shedding occurs over the

152 © 2011 by Taylor & Francis Group, LLC

Figure 49. Deflections and bending moments.

Figure 50. Monotonic and cyclic p-y curves.

upper section of pile, where high reaction pressures are mobilised in the first cycle. These are transferred down the pile into regions where lower, and hence nondegrading reaction pressures are mobilised in the first cycle, which allows the applied reaction pressures to increase with advancing cycles. Example p-y curves extracted from pCyCOS at selected depths for a fixed head pile are presented on Figure 50 for monotonic loading and for N = 20. These demonstrate nicely the substantial difference between monotonic and cyclic loading conditions. The resulting p-y curves can be input directly into structural analysis packages, although some manipulation of the cyclic p-y curves may be required to avoid numerical difficulties associated with their strain softening nature. 4

CONCLUSIONS

This paper has presented and demonstrated application of two new tools for the analysis and design of piles in carbonate soil: CYCLOPS for axial loading and pCyCOS for lateral loading. The procedures used in both

153 © 2011 by Taylor & Francis Group, LLC

are underpinned by a strong theoretical basis, but also include a number of factors determined from empirical calibrations in order to properly address observed soil behaviour. A key objective in the development of these methodologies was to simulate real behaviour as accurately as possible, but where the available data led to any doubt, conservative assumptions have been adopted. The procedures were also developed to be applicable to a wide variety of general situations, although as with all geotechnical design in carbonate soils, great care and good engineering judgement remain vital ingredients. With respect to axial pile design, we consider that the ‘CNS calibration approach’ embodied in CYCLOPS is an essential process for developing safe designs of cyclically loaded drilled and grouted piles in cemented carbonate soils.Actual sites exhibit highly variable properties; these may differ significantly from the specific characteristics encountered at Overland Corner, which remains the only site where a comprehensive program of cyclic pile tests has been undertaken in cemented carbonate soil. Direct scaling of the OC test results is therefore not recommended. The CNS device is the main tool available for differentiating the behavioural characteristics between different soil types and hence underpins the CYCLOPS method. With respect to lateral pile analysis, the pCyCOS procedure is also based on a rigorous link to laboratory monotonic and cyclic element testing, from which the basic p-y curves are developed. It is unfortunate that at this stage we do not have any equivalent of the OC tests for cyclic undrained lateral loading in uncemented carbonate soils. However, the centrifuge tests reported in Dyson (1999) show the same qualitative response as obtained with pCyCOS. It should also be appreciated that this method only represents a modest philosophical advance on the industry standard procedures that would be adopted for design of a shallow foundation or suction pile, albeit the practical implementation for lateral pile analysis proves to be necessarily complex. The reality of the destructive effects of cyclic loading on shallow foundations in carbonate soils has been amply demonstrated in full scale field situations (Erbrich, 2005). While the approach in pCyCOS may yield significantly more onerous results for cyclic loading than implied by current procedures (and this may have implications for existing structures), the paramount concern has been to ensure safe design, taking due account of the observed behaviour in laboratory testing. To date, the authors have utilised CYCLOPS and pCyCOS for the design of several major new facilities and all these designs have been accepted by third party certification bodies. Finally, it should be appreciated that while the discussions in this paper have centred on carbonate soils, the methodologies developed are of more general applicability. For example, pCyCOS could be used for any kind of uncemented soil, provided suitable laboratory test data is available to enable development of

the required p-y curves. In practice, this would probably only be a worthwhile activity where the soil in question might be subject to significant cyclic degradation under the design loading. Similarly, CYCLOPS has been developed specifically for weakly cemented carbonate rocks, but could also be applied to axial pile design in other kinds of soft rock, provided suitable CNS test data is available to allow determination of the necessary calibrated parameters. ACKNOWLEDGMENTS The authors would like to acknowledge their colleagues who made significant contributions to this work. In particular we would like to thank Professor John Carter who performed detailed independent analyses to verify the pCyCOS program, Edgard Barbosa Cruz, who performed the ABAQUS analyses used in the development of pCyCOS and to Mohamed Khorshid who provided support and encouragement throughout this work. We would also like to thank the many others in both AG and throughout our industry who provided constructive and valuable contributions to the development of these new design tools. REFERENCES ABAQUS Inc. (2006), ABAQUS Version 6.6, User Manual, Dsimuli Abbs, A.F. (1983). Lateral pile analysis in weak carbonate rocks. Proc. Conf. on Geotechnical Practice in Offshore Engineering, ASCE, Austin, 546–556. API (2000). API RP2A – Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms—Working Stress Design, 21st Edition, American Petroleum Institute, Washington. Coyle, H.M. and Reese, L.C. (1996). Load transfer for axially loaded piles in clay. J. Soil Mechanics and Foundations Division, ASCE, 92(SM2), 1–26. Dyson G.J. (1999), Laterally Loaded Piles in Calcareous Sediments, PhD Thesis, UWA. Dyson, G.J. and Randolph, M.F. (2001). Monotonic lateral loading of piles in calcareous sediments. J. Geotech Eng. Div, ASCE, 127(4), 346–352. Erbrich, C.T. (2004). A new method for the design of laterally loaded anchor piles in soft rock. Proc. Offshore Tech. Conf., Houston, Paper OTC 16441. Erbrich, C.T. (2005). Australian Frontiers – Spudcans on the Edge. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth.

Haberfield C.M. Paul D.R. Ervin M.C. and Chapman G.A. (2010) Cyclic Loading of Barrettes in Soft Calcareous Rock using Osterberg Cells. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth. Hoeg K. (1976), State of the Art – Foundation Engineering for Fixed Offshore Structures, Proc. BOSS76, Trondheim. ISO (2007). ISO 19902 – Petroleum and Natural Gas Industries — Fixed Steel Offshore Structures. International Standards Organisation, Geneva. Itasca (2005), FLAC Fast Lagrangian Analysis of Continua, User Manual, Itasca Consulting Group. Jeanjean, P., Watson, P.G., Kolk, H.J. and Lacasse, S. (2010). RP 2GEO: The new API recommended practice for geotechnical engineering. Proc. Offshore Technology Conf., Houston, Paper OTC20631. Kraft, L.M., Focht, J.A. and Amarasinghe, S.F. (1981). Friction capacity of piles driven into clay. J. Geotechnical Engineering Division, ASCE, 107(GT11), 1521–1541. Murff, J.D. & Hamilton, J.M. (1993), P-Ultimate for Undrained Analysis of Laterally Loaded Piles, Journal of Geotechnical Engineering Div. ASCE, 119(1), pp 91–107. Novello, E. (1999). From static to cyclic p-y data in calcareous sediments. Proc. 2nd Int. Conf. on Engineering for Calcareous Sediments, Bahrain. 1, 17–27. Randolph, M.F. (1986). RATZ: Load transfer analysis of axially loaded piles. Report No. Geo:86033, Department of Civil Engineering, The University of Western Australia. Randolph, M.F. (1988). Evaluation of grouted insert pile performance. Engineering for Calcareous Sediments, Perth, 2, 617–626. Randolph, M.F. (2003), RATZ Version 4.2 – Load Transfer Analysis of Axially Loaded Piles, User Manual. Randolph M.F., Jewell R.J. & Poulos H.G. (1988), Evaluation of Pile Lateral Load Performance, Proc. Eng. for Calcareous Sediments, ed. Jewell R.J. & Khorshid M.S. Randolph, M.F. and Jewell, R.J. (1989). Axial load transfer models for piles in calcareous soil. Proc. 12th Int. Conf. on Soil Mech. and Found. Eng., Rio de Janeiro, 1, 479–484. Randolph M.F., Joer H.A., Khorshid M.S. and Hyden A.M. (1996), Field and laboratory data from pile load tests in calcareous soil’, Proc. 28th Annual Offshore Tech. Conf., Houston, Paper OTC7992. Randolph, M.F. and Wroth, C.P. (1978). Analysis of deformation of vertically loaded piles. J. of Geotechnical Engineering Division, ASCE, 104(GT12), 1465–1488. Wesselink, B.D., Murff, J.D., Randolph, M.F., Nunez, I.L. and Hyden, A.M. (1988). Analysis of centrifuge model test data from laterally loaded piles in calcareous sand. Engineering for Calcareous Sediments, Perth, 1, 261–270. Wiltsie, E.A., Hulett, J.M., Murff, J.D., Brown, J.E., Hyden, A.M. and Abbs, A.F. (1988). Foundation design for external strut strengthening system for Bass Strait first generation platforms. Engineering for Calcareous Sediments, Perth, 1, 321–330.

154 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

New frontiers for centrifuge modelling in offshore geotechnics C. Gaudin Centre for Offshore Foundations Systems, UWA, Perth, Australia

E.C. Clukey BP America Inc., Houston, USA

J. Garnier Laboratoire Central des Ponts et Chaussées, Nantes, France

R. Phillips C-CORE, Newfoundland, Canada

ABSTRACT: Today, centrifuge modelling is a recognised tool, both within the academic and the industry community, to help understanding the behaviour of offshore structures and assist in their design. However, although expanding, its real impact on design practices is limited. The paper proposes, therefore, to revisit the perception of the advantages and disadvantages of centrifuge modelling and the technological developments achieved over the last few years, in the light of recent offshore projects undertaken. It aims to provide both academic and industry perspectives. Examples are presented that illustrate the difference between centrifuge testing and centrifuge modelling and which highlight the contribution that centrifuge methods can make in designing offshore structures and in providing new insights into soil-structure interaction. 1

INTRODUCTION

The initial use of centrifuge modelling in offshore geotechnics took place in Manchester University in 1973 where the behaviour and performance of gravity platforms to be used in the Gulf of Mexico was investigated (Rowe & Craig, 1981). The work encompassed a wide range of soil and loading conditions and provided pivotal insights into the failure mechanism taking place (Craig & Al-Saoudi, 1981). Even in the very early days of centrifuge modelling (when the technological capabilities were limited compared to present day), it was yet understood and acknowledged that centrifuge modelling could contribute significantly to design when novel or incompletely understood conditions (both in terms of soil and loading) prevailed (Craig, 1984). Since this pioneering work in 1973, a significant number of projects have been undertaken worldwide, on a wide variety of offshore geotechnical problems. A few centrifuge centres have emerged, in the wake of Manchester University, developing expertise on offshore geotechnics through collaboration with industry, such as Cambridge University in the UK (CUED), C-CORE in Canada, LCPC in France, Deltares (former Geodelft) in Netherlands, University of Colorado, Boulder in the USA and COFS at the University of Western Australia.

The work performed, focusing first on phenomenological and site-specific studies, developed progressively towards more general studies, including the observation of failure mechanisms and the understanding of soil-structure interaction, eventually aiding in the development of predictive design methods in some cases. This transition from what was initially centrifuge modelling to centrifuge testing is discussed further in the paper. As centrifuge modelling was technically and scientifically developing, along with an increasing need for performance data and understanding of offshore soil structure interaction, the acceptance and awareness by the offshore community of the benefits of the use of centrifuge grew significantly. A key milestone in that process was certainly the keynote address given by Don Murff at OTC (Offshore Technology Conference) in 1996 to the wider offshore community, advocating the benefits of centrifuge modelling (Murff, 1996). Since then, the number of offshore projects benefiting from centrifuge modelling input has increased significantly and industry users have developed a strong expertise in analysing centrifuge modelling outcomes and incorporating them into their design approach. Nevertheless, as new needs arise from the increasing complexity of offshore projects, along with new possibilities from the technological development of new modelling techniques, it appears necessary to

155 © 2011 by Taylor & Francis Group, LLC

revisit the benefit and the contribution centrifuge testing and modelling may provide to the offshore industry in designing offshore structures subsea facilities and pipelines. This paper aims then to answer the fundamental questions any potential user has once facing a challenging offshore project: Why use a geotechnical centrifuge? Is it cheaper or more expensive than alternative investigation technique?; is it more or less reliable?; is it quicker to obtain relevant data?; does it account for behaviour aspects other techniques do not?; should it be replace or complement other techniques such as 1g reduced scale modelling or field testing? How to use a geotechnical centrifuge? Is it better to use natural or artificial soils?; should it investigate a wide range of options or focus on a particular solution?; should it aim at validating a design or assisting in the design; should it be modelling or testing? When to use a geotechnical centrifuge? Should it be considered at the early stages of the project or latter once some key parameters are already established?; at what stage within the design process is centrifuge modelling the most valuable? To answer these three questions, this paper proposes: 1. to revisit the advantages (and disadvantages) of the use of centrifuge in light of the scientific and technological developments of the past decade; 2. to detail the contribution of the use of centrifuge on industry practices, highlighting the difference between centrifuge testing and centrifuge modelling, an; 3. to evaluate the impact of centrifuge modelling/testing in industry practice and eventually to present some perspective for the future.

to obtain qualitative or quantitative information about the problem. The difference between centrifuge testing and modelling follows similar concepts. By performing centrifuge testing, one aims at validating or confirming a general set of assumptions about a geotechnical problem, while by performing centrifuge modelling, one aims at predicting or anticipating the behaviour or response of an actual geotechnical structure. The difference is significant as the requirements, the modelling techniques and the objectives for the modeller are different, as will be the outcomes for the user. 2.1

Focusing on soil-interaction structure, examples are presented from the authors’ experience, providing the perspective of both academic modellers and an industry user. Indeed, this presents a small part of the entire centrifuge work performed world wide. However, the authors believe that the variety of examples is representative of the contribution and possibilities that centrifuge testing and modelling can offer for the offshore industry.

2

ROLES OF CENTRIFUGE TESTING AND CENTRIFUGE MODELLING

The process of designing, performing, analysing centrifuge experiments and eventually integrating the outcomes into the design of offshore structures cannot be undertaken correctly without a clear understanding of the difference between centrifuge testing and centrifuge modelling. In engineering, it is common to refer to in-situ testing and numerical modelling. The former is usually used to validate an assumption (soil properties, design loads, soil response, etc) while the latter resorts to an idealisation of the problem investigated

156 © 2011 by Taylor & Francis Group, LLC

Centrifuge testing

The aim of centrifuge testing is to create an idealised representation of a problem in order to obtain quantitative and/or qualitative prediction about the mode of behaviour of the structure investigated, (Lee, 2001). It is therefore necessary to interpret the results in light of a model for soil-structure behaviour, and then use that model to simulate the design situation. Being an idealisation, the model does not necessarily replicate all the features of a real structure (commonly called prototype) and consequently, some flexibility is permitted regarding the representativeness of the model as long as the key features of the behaviour which are investigated are modelled correctly. The difficulty is to know which features are simulated exactly, approximately or not at all and to account for these differences in the interpretation of the results. The complementary use of other modelling and testing methods (such as finite element analysis) may help in this interpretative process. The idealisation relates mostly to the soil and the structure as follows: – The soil used in the model may be different from the prototype as long as some pivotal features of the behaviour are reproduced (such as the dilatancy of dense sand or the shear strength of cohesive soil, for instance). Local heterogeneity and non-critical stratigraphy or topography are often not simulated, but may still be accounted for to interpret dissimilar behaviour. – Some particular features of the prototype (geometric, mechanics, etc), which are believed to not have a significant effect on the phenomenon investigated, are not simulated. – Phenomenon aside of the ones investigated, and for which the influence is not believed to be dominant, are not simulated (a specific method of installation, a particular loading sequence, etc). In designing the model and choosing which features to replicate or not, the engineer needs to make assumptions about the behaviour to be modelled, to understand the consequences of the simplifications made and to define clearly the purpose of the model. Centrifuge testing may be conducted to: – Develop an initial understanding of the engineering concern.

– Identify a particular failure mechanism on which an analytical solution can be developed. – Observe a particular mode of soil behaviour (is the response drained or undrained, does the soil flow or collapse? etc). – Gather performance data which can be used to calibrate a numerical model. – Perform a sensitivity study to understand the relative importance of various parameters in the behaviour of the structure. – Perform parametric studies to generate design charts. – Determine soil properties by performing reduced scale testing of in-situ soil properties.

2. The requirements to represent the prototype are much stricter, particularly with regards to the soil characteristics and the loading conditions. Hence, as opposed to centrifuge testing, one, when designing centrifuge modelling should ensure that: – The same soil, featuring the same properties as the prototype, is used in the model. – The model considers all the details of the prototype (e.g. particular attention must be paid to the fidelity of the model). – All key phenomena involved in the response of the structure must respect the similitude principles. – Loading conditions must be the same between the model and the prototype (rate of loading, amplitude and frequency of loadings, etc).

Indeed, the objectives sought may affect the idealisation made as much as the assumptions and some judgement will be required to decide whether or not the outcomes may be directly extrapolated to prototype conditions and whether or not they need to be checked or validated by other means. The objectives presented could sometimes be achieved by other means, such as by numerical modelling, but the centrifuge offers the advantage to replicate specific features of the soil behaviour or the structure loadings. These include soil softening and reconsolidation, large deformation, cyclic loadings, collapse and the installation process. All of these features are particularly relevant for offshore geotechnics. Often, the complementary use of both numerical modelling and centrifuge testing accelerates the understanding of the phenomena, as the two methods are based on different sets of assumptions.

In other words, in centrifuge modelling, all aspects of the model must be properly scaled to that of the prototype. The objectives of centrifuge modelling may be various, although they are limited to the case studied. Extrapolation to other conditions should be undertaken with caution. Centrifuge modelling should help to: – Validate a technical solution (type and geometry of structure). – Validate or evaluate the performance of a final design (design loads, displacements within the limits established, etc). – Validate a mode of behaviour upon which the design was based (i.e. mode of collapse). – Observe the response of a given structure under a particular type of loadings, possibly different from the one used for design (e.g. check the performance of the structure under cyclic loadings).

2.2 Centrifuge modelling The aim of centrifuge modelling is to establish the validity/relevance of certain assumptions, designs or models. It aims to directly replicate the field conditions. The results provide an immediate representation of the design situation. What characterises centrifuge modelling is the uniqueness of the model investigated. It must refer to a specific prototype (as opposed to centrifuge testing) and focus on a particular aspect of the validation. This allows definitive conclusions about the prototype behaviour to be made. Note that the assumptions, design or model to be validated might have arisen from centrifuge testing (Lee, 2001). The reference to a particular prototype results in two fundamental differences with centrifuge testing: 1. The use of similitude principles and associated scaling laws (Garnier et al., 2007) is required to extrapolate the information gathered on the model (typically the measurements of dimensions such as loads, displacements, etc) to the prototype. The use of similitude principles in centrifuge testing is indeed required (this is the justification for the use of the centrifuge), but the scaling laws (e.g. the use of prototypes dimensions) are not necessarily required.

The objective presented here could be also achieved by field testing. However, centrifuge modelling offers the advantage (while ensuring the correct stresses are applied within the soil), of being less costly, provides a better control of the modelling conditions (e.g. known soil characteristics, better quality of data), and allows more data to be collected in a shorter period. It also provides the opportunity to repeat the experiments if required.

2.3 The use of centrifuge outcomes into design The design of an offshore foundation structure follows a series of steps, which are simplified in Figure 1. The soil and loading conditions will dictate a particular type of foundation whose geometry and characteristics will be determined using design methods, either empirical or based on constitutive soil models linking strains and displacements to the stresses and loads. Along with these different steps, the geotechnical engineer will use various tools to help him assess the soil conditions, choose the relevant technical solution, develop or use an existing design method and eventually validate the design.

157 © 2011 by Taylor & Francis Group, LLC

Figure 1. Steps involved in designing geotechnical structures and potential input of centrifuge modelling and testing (white shapes symbolises design steps, shaded shape symbolises centrifuge modelling or testing).

These tools can be divided in two categories; field testing and numerical/analytical modelling. Field testing is used to characterise the soil and help validate the design, while the modelling methods would be used to develop the design. The objective of this paper is to describe the potential centrifuge methods may have to assist in geotechnical design. As illustrated in Figure 1, by performing both testing and modelling, centrifuge methods may be a valuable tool for geotechnical engineers in every phase of the engineering process. Centrifuge testing may be used to develop a new foundation concept, a new technical solution, or to characterise soil properties. It may also be used to investigate soil behaviour and failure mechanism to establish constitutive models. In contrast, centrifuge modelling may be used to validate a new foundation concept, to validate a design and to calibrate and validate a specific design method. In offshore geotechnical engineering, where guidelines and design rules play a lesser role than onshore, engineers’ judgement is paramount in establishing the most technically sound and economically efficient design. Basing judgment on reliable and quality data from a range of modelling and testing methods may make a significant difference between a satisfactory and an optimal design.

The following sections will present various examples of centrifuge testing and modelling, highlighting the contributions to design. It is not implied here that centrifuge methods should necessarily replace more classical methods. Indeed, for routine design, classical methods are often preferred as they provide a safe and efficient path towards satisfactory solutions. However, for more challenging geotechnical design issues, which frequently occur in offshore engineering, geotechnical engineers should seriously consider the potential centrifuge methods may offer. 2.4

The offshore industry has come to accept the advantages of centrifuge methods as more and more results from various studies became available. ExxonMobil (former Exxon) were initial leaders in this area with a number of programs performed in the early 1980s (Murff, 1996). With the publication of the ExxonMobil results and, as the industry encountered more complex problems where 1-g testing was either difficult and/or expensive to perform, centrifuge testing or modelling became a more viable option for other operators. Joint Industry Programs involving a significant centrifuge testing component have become increasingly more frequent in the past 2 decades, such as

158 © 2011 by Taylor & Francis Group, LLC

Industry perspective

the SEPLA: suction emplaced plate loading anchor JIP involving 5 operators and 2 consulting groups in early 2000 at C-CORE. The Pressure Ridge Ice Scour Experiment (PRISE) performed at C-CORE attracted over 7 operators and 2 regulatory agencies and developed guidelines for design of buried offshore pipelines subject to ice scouring (Phillips et al., 2005). The INSAFE JIP, gathering 19 offshore operators, jack-up manufacturers, safety regulators and consulting companies, has been using centrifuge testing and modelling outcomes initially from Cambridge University, and more recently from COFS and NUS (National University of Singapore) to provide guidelines for design and best practice of jack-up installation and performance. Other examples of offshore JIPs using centrifuge techniques, in complement of more traditional techniques, includes the SAFEBUCK JIP for the design of on—bottom pipeline subjected to buckling, the MERIWA JIP, involving six major offshore companies, to model submarine landslide and their impact on pipelines. In France also, several centrifuge projects were carried out at LCPC, founded by CLAROM (www.clarom.com), an institution which gathers marine and offshore companies to carry out R&D programs aiming at better understanding offshore structures behaviour and improving design methods. The success of centrifuge techniques outside the offshore community has also helped to convince management to support various centrifuge efforts. The ability, for instance, of centrifuge tests to accurately define the failure mechanism for the levy failures in New Orleans during Hurricane Katrina is a recent example of the usefulness of the technique for defining failure modes. (Sasanakul et al., 2008 and Ubilla et al., 2008). The use of centrifuge testing by numerous earthquake researchers involved in a very large United States program (Wilson et al., 2010) also helped improve the credibility of the methodology which led to testing of large gravity based structures under ice and seismic loads (DeWoolkar et al., 2006 and DeWoolker et al., 2008). More recently the industry has used centrifuge techniques for a large number of deepwater foundation as well as subsea and pipeline issues, based on an operator-testing contractor partnership, rather than based on a JIP (Gaudin et al., 2010a). The timeliness in which results could be obtained was often a major reason why centrifuge methods were selected vs. alternative techniques such as 1g testing. This will be developed further in section 5 and 6.

3

PERCEPTION OF ADVANTAGES AND DISADVANTAGES OF CENTRIFUGE MODELLING AND TESTING

The advantages and disadvantages of centrifuge modelling are well known, and have been listed and commented in many publications, including Murff (1996) for offshore related problems.

However, centrifuge testing and modelling is an advancing science and the advantages and disadvantages are continuously changing with the problems they are investigating. Hence, some previously perceived disadvantages are disappearing due to progress made in centrifuge modelling theory and techniques, while others are appearing as centrifuge methods are applied to new geotechnical problems. This section presents advantages and disadvantages of centrifuge methods, as they relate to offshore geotechnics today. 3.1 Advantages The main advantages are as follows: 1. Centrifuge modelling properly simulates the prototype body forces and, thereby, maintains proper relationships with stresses and strains. 2. Because actual model tests are small, tests can be performed in a relatively short amount of time and at reduced cost, allowing the investigation of a wide range of parameters. 3. The field in-situ conditions are more easily replicated than is usually the case for 1-g model tests. Advancements in in-situ devices used to measure soil properties in the centrifuge, such as the piezocone and T-bar (Stewart and Randolph, 1994) have significantly enhanced the knowledge of the material being tested. 4. Pore pressures dissipate much more rapidly than in the prototype condition. This is clearly an advantage when preparing a model for testing, since it reduces consolidation time, and for testing, as cyclic sequences over several years can be modelled within hours in the centrifuge. For monotonic loading in clays, undrained conditions in clays can usually be maintained by loading to failure at a higher rate and correcting for strain rate effects. 5. Centrifuge test data can be used to validate numerical models. This method is very efficient since the soil characteristics, the boundary conditions and the applied loads are known. This is not always the case in full scale field testing. 3.2

Disadvantages are more difficult to list, as it is necessary to separate the perceived disadvantages resulting from a somewhat limited knowledge of the physical modelling principles and the real disadvantages, which the centrifuge modelling community clearly identifies (Garnier et al., 2007), and for which mitigation or correction measures may be applied. The following presents a list of the disadvantages and the measures to undertake to avoid or mitigate them: 1. Scale effects (different from size effects) were for a long time listed as the main disadvantage of centrifuge methods, supposedly preventing extrapolation of centrifuge modelling and testing outcomes to field problems. As the size of the particles may

159 © 2011 by Taylor & Francis Group, LLC

Disadvantages

not be negligible when compared to the dimensions of the model, some particular soil behaviour may be generated, that is different to that expected in-situ. This is a scale effect, first identified by Ovesen (1975) who observed different responses for footing of various widths loaded on sand. Ovesen concluded that, for flat footing resting on sand, the width of the footing should be higher than 30 times the mean diameter of the particles. This prevents any scale effects to occur (in this case, an increase of the prototype bearing capacity with a reduction of the footing width). Indeed, the identification of a scale effect as the ratio of a relevant model dimensions to the particle size indicate that this type of problem is more likely to be critical for sand than for clay. Over the years, various research has been undertaken to quantify the minimum model dimension to particle size ratio for a variety of problems. This includes a significant number of bearing related geotechnical problems and, maybe more importantly, interface shearing problems where the scaling of the interface roughness may lead to significant difference in behaviour between the model and the prototype (Garnier & Köenig, 1998). The outcomes of all this work and some significant progress made in establishing similitude principles (notably through dimensional analysis) have been gathered in a scaling law catalogue published by Garnier et al. (2007). This catalogue provides modellers with a set of rules to design valid modelling and testing, free of potential scale effects, or to account for them when extrapolating the results to prototype conditions if they cannot be avoided (e.g. increased shearing forces due to scale effect can be calculated following procedure developed by Lehane et al. (2005). In practice, as already pointed out by Murff (1996), scale effects are rarely an issue in offshore geotechnics, especially as marine soils are mostly constituted of small particles (clay and silt in majority). If scale effects are really a major concern (for instance, if prefailure displacements in sand need to be accurately scaled, Palmer et al., 2003), modelling of models (testing different size models at different accelerations to model the same prototype) is a relevant technique to validate the modelling outcomes. 2. Creep is a complex phenomenon which may not be modelled correctly in the centrifuge. If creep relates to volumetric compaction (e.g., the displacement of a flat footing subjected to sustained vertical loading), the rate of creep is governed by the rate water escapes or moves into a volume of soil. It is therefore scaled by n2 , similarly to consolidation. Consequently, any creep phenomenon would occur n2 time faster in the centrifuge and would be captured by the model within the time frame of the experiment. If creep relates to shear deformations (e.g. the displacement of a pile subjected to sustained tensile loading), the rate of creep is governed by the viscous resistance of the soil structure and displacements in the centrifuge would occur at

the same rate as that in the field (as the strain rate would be equivalent). This result has, for example, been observed in centrifuge loading tests of shallow footing resting on dry sand (Canepa et al., 1988). Consequently, creep phenomena related to shear strains are difficult to capture in a centrifuge model within the usual time frame of an experiment. 3. Inertia effects scale differently than body forces. Therefore, for earthquake problems the loading frequencies need to be significantly increased. Even with these enhanced frequencies the pore pressure response is much faster than the prototype case in cohesionless sandy soils and a more viscous pore fluid is required to properly model this response. This is now a commonly used technique as demonstrated in section 4.4. 4. For problems where shear strain localization issues are important, such as ice gouging of the seafloor, centrifuge test interpretation of discontinuities versus shear zones is important. The use of physical modelling of models and complementary numerical modelling has been found to be beneficial in correctly interpreting the observed behaviour (Phillips et al., 2010). 5. The acceleration field across the models are variable and dependent on the radius from the centre of rotation to any point on the model. Corrections are available to account for this effect. Although some of the potential problems can be overcome or mitigated, as described above, the different scaling relationships for body forces, pore pressures viscosity and inertia can make testing more complex and difficult to perform. It relies then on the experience of the modeller to clearly establish the limits of the validity of the model. 3.3

An implicit advantage of centrifuge testing is that similitude relationships are almost always directly addressed. Conditions where similitude is or is not achieved are more carefully considered. Therefore, the advantages and limitations of the technique are better defined. In contrast, 1-g model tests have significant similitude issues, requiring the modification of the analytical models used, to match the experimental model test data. However, if the experimental model tests do not properly account for all the key factors that influence the prototype behaviour, then the potential for errors when the analytical models are extended to the prototype condition is significant. A potential useful non-geotechnical analogy to illustrate the importance of considering similitude is the model testing performed for deepwater production facilities such as TLPs, SPARs and FPSOs. For almost any large scale offshore project that utilizes one of these facilities, extensive tests are performed in model test basins (Newman, 1977). The modellers and engineers who perform these tests are always aware of the key model parameters, such as the Froude number and

160 © 2011 by Taylor & Francis Group, LLC

Consideration of similitude

Reynolds number when interpreting results. Because all the key similitude parameters for their problem are usually not satisfied (Lyle Finn, personal communication), the model tests are augmented with extensive field measurements to further enhance measurements used to calibrate analytical and design models. Perhaps regrettably, comparable field measurements are not often performed on offshore foundations. Therefore, there is a greater reliance on the model test data for the foundation design procedures and a greater demand on centrifuge modellers to provide modelling validation. The laws of similitude need to be carefully considered when performing model tests. To assist modellers in validating their model tests, the centrifuge community has been working extensively to produce a scaling law catalogue, listing scale factors and dimensional analysis procedures for a wide range of geotechnical problems (Garnier et al., 2007). A particularly relevant example is the dimensional analysis carried out by Boylan et al. (2010a) to validate the centrifuge modelling of the behaviour of the runout of submarine slides. While 1g laboratory are commonly performed to investigate runout behaviour, the modelling of a phenomena which relies on both soil mechanics and fluid mechanics necessitated detailed similitude analyses. 4

CENTRIFUGE TECHNOLOGY DEVELOPMENT

Over the last 30 years, the progress made in miniaturised electronics, micro-computing, software engineering and digital imagery have been incorporated almost immediately in centrifuge technology. It results in more realistic simulations, more accurate measurements and more detailed observations, subsequently improving the benefit centrifuge modelling can yield to the geotechnical community. This is particularly relevant for offshore geotechnics where complex environmental conditions, emphasis in capacity and saturated soil conditions prevail. Hence, sophisticated motion control to apply environmental cyclic loadings, image acquisition techniques to observe failure mechanisms and enhanced data acquisition systems to measure accurately pore pressure developments and fast responses have been key developments in centrifuge modelling techniques. Some of the most noticeable developments recently undertaken in centrifuge technology are presented below, with examples highlighting the benefits obtained from these improvements presented in sections 5 and 6. 4.1

Data acquisition system

In the early days of centrifuge testing data acquisition was limited to post-testing measurements and observations. With the development of computing and electronic, on-board data acquisition systems appeared. A typical data acquisition system would

Figure 2. New high-speed data logging unit developed at COFS. The unit replaces standard PC-data acquisition card combination (after Gaudin et al., 2009a).

include transducers, a signal amplifier and an analog/digital converter system in a slave computer onboard of the centrifuge (Liu et al., 1988; Garnier & Cottineau, 1988). The signal would be transmitted through electrical slip rings to a master computer inside the control command room. In such a system, electrical slip rings constitute the weak point (Taylor, 1995). They are prone to failure and generate significant electrical and mechanical noise. Over the years, improvements in sampling, signal resolutions and signal conditioning were undertaken, with 128 channels sampled at 100 Hz with a resolution of 16-bits being the norm. The use of either optical slip rings or wireless transmission also contributed to improve the quality to the data acquired. Nevertheless, the overall architecture remained unchanged as one can see in the development of the newest centrifuge facilities (Ng et al., 2006). Recently, some significant breakthroughs in data acquisition technology have been made at COFS. An independent self-powered unit 150x60x40 mm in size, embarking processing and storage units and capable to monitor up to 8 transducers, has been developed for centrifuge application (Gaudin et al., 2009a). Each unit (see Figure 2) is capable of powering and monitoring eight instrument channels at a sampling rate of up to 1 MHz at 16-bit resolution. The data are stored within the logging unit in solid-state memory, but may also be streamed in real-time at low frequency (up to 10 Hz) to the centrifuge control room, via wireless transmission. Unlike PC-based data acquisition solutions, this system performs the full sequence of amplification, conditioning, digitization and storage on a single circuit board via an independent micro-controller allocated to each pair of instrumented channels. It results in a much cleaner signal and the faster sampling rate, coupled with automatic triggering features, proves to be extremely useful in acquiring data during cyclic events or fast events such as submarine landslide (Boylan et al., 2009a). 4.2

Actuation devices have followed an improving trend similar to data acquisition systems. Early devices

161 © 2011 by Taylor & Francis Group, LLC

Robotic control

such as hydraulic mono-axis actuators developed into bi-axis electric or hydraulic actuators, to eventually multi-axis robots, first in LCPC (Derkx et al., 1998) and then in HKUST (Ng et al., 2002). In parallel, the development of sophisticated motion control systems occurred (De Catania et al., 2010). Complex loading sequences, including varying control modes on various axes are now possible, mimicking more realistically typical offshore loadings and permitting the study of particular aspects of soil structure interaction. One example is the investigation of pipeline embedment during laying process, when the pipe is subjected to a combination of horizontal and vertical cyclic motion with varying amplitude (Gaudin & White, 2009). This example is developed in more detail in section 6. The last ten years have also witnessed the development of actuation devices to generate wave loadings and wave-induced phenomenon. The device commonly includes a wave paddle (Sassa & Sekigushi, 1999) or a plunger moved by a horizontal (Baba et al., 2002). In both cases, frequency and amplitude of the motion of the wave generator device may be controlled in order to generate the appropriate wave spectrum. Particular attention is paid to limit or control the wave refraction against the edges of the centrifuge container and the use of fluids with viscosity higher than water to satisfy the similarity rule with regard to laminar flow (e.g. conservation of the Reynolds number). The drum centrifuge, which features a high developed length, was initially used for such tests (Phillips & Sekiguchi, 1992) and is particularly appropriate to model waves (Gao & Randolph, 2005). The steel catenary riser tests at C-CORE described in section 5.7 utilise 2 synchronised servo-hydraulic actuators to provide the necessary broad band frequency surge and heave motions (Figure 3). The actuators are counter balanced to minimise driving forces to maximise the system frequency response. 4.3

Digital imagery

In the early ages of centrifuge modelling, computer image processing techniques were used in combination with digital cameras (Garnier et al., 1991). The subsequent development of the particle image velocimetry (PIV) and photogrammetry techniques, for soil mechanics, by White et al. (2003) have constituted a major breakthrough for centrifuge technology, expanding considerably the range of investigations. More detailed observations can be gathered and insight into failure mechanisms taking place around foundations can be obtained. This is particularly relevant for offshore foundations where failure often governs the design, as opposed to onshore foundations where designs are governed more by displacements or serviceability. The technique consists in processing digital images of the soil and structure investigated, placed against a transparent observation window (see Figure 4), in order to obtain information about the displacements of the soil and the structure (White et al., 2005).

Figure 3. Servo hydraulic actuators developed at C-CORE.

The use of the PIV technique has been pivotal in a significant number of breakthroughs on the behaviour and performance of offshore structures. One may highlight: 1. The identification of the back flow mechanism, as opposed to the wall collapse mechanism previously assumed, governing the limiting cavity depth during spudcan penetration (Hossain et al., 2005). 2. The identification of the bearing capacity mechanism for spudcans penetrating sand overlaying clay, which showed the changes in overall failure mechanism due to varying geometric and strength conditions of the layered soil (Teh et al., 2008). 3. The identification of the spudcan extraction mechanism and the resulting insight provided about the development of suction at the spudcan invert (Purwana et al., 2006) or the identification of the ground movement generated by spudcan penetration leading to indirect loading of nearby piles (Leung et al., 2008). 4. The identification of the berm creation mechanism and its influence on the soil resistance during lateral pipeline motion (Dingle et al., 2008). 5. The identification of a Hill-type reverse end bearing mechanism during uplift of skirted foundation, due to the development of suction at the foundation invert (Mana et al., 2010).

162 © 2011 by Taylor & Francis Group, LLC

Other pivotal examples include the identification of the strain path followed by a soil particles nearby the tip of a pile which contributed to the development of theoretical solutions for the assessment of pile

Figure 4. Typical PIV setup in the COFS drum centrifuge (after White et al., 2005).

Figure 5. Soil displacement vectors for skirted foundation with embedment ratio = 0.2 subjected to uplift (Mana et al., 2010).

base resistance (White & Bolton, 2004) and the understanding of the phenomenon of friction fatigue which governs the shaft resistance along piles driven in sand (White & Bolton, 2001).

The PIV technique may provide insights, not only on the soil displacement pattern and the failure mechanism taking place, but may also reveal the particular behaviour of the structure investigated. A

163 © 2011 by Taylor & Francis Group, LLC

typical example of such insight of offshore structures behaviour is the identification and the understanding of the mechanism governing the behaviour of keying flap of suction embedded plate anchors. Recent experiments clearly demonstrated the inefficiency of the keying flap in reducing the loss of embedment during keying (Gaudin et al., 2010b). All these examples highlight the dramatic improvements in the utility of physical modelling in providing important insights in the mechanisms and behaviours which can subsequently be used to develop, improve or validate design and calculation procedures. 4.4

Soil sample reconstitution and characterisation

Improvements in soils sample reconstitution results more from the experience gathered by modellers and from the better understanding of soil behaviour, than from technological developments. Automatic air pluviation techniques to reconstitute sand samples and in-flight consolidation of normally consolidated clay samples or in-press consolidation of over consolidated clay samples are now well established techniques in every centrifuge centre, resulting in homogeneous and well controlled soil samples (Garnier, 2002). Nevertheless, recent projects have seen a significant increase in sophistication of the soil samples reconstituted, with natural soils sourced in-situ (as opposed to commercially available clay and sand) becoming more common, both in full-size samples, or in small boxes for PIV analysis. This has become common practice at, for example, LCPC (natural clay from North Sea and from West Africa) and at COFS (carbonate sediments from North West Shelf, soft clay from West Africa). The reconstitution of multi layered soils (typically a sand layer within a clay seabed) and the use of alternative pore fluid in granular material (such as silicon oil or aqueous methyl cellulose) to increase the pore fluid viscosity in order to generate and observe liquefaction phenomenon is also becoming common practice. A typical example is the reconstitution of a sample featuring a kaolin clay layer saturated with water overlaying a silica sand layer saturated with silicon oil. This technique has been used to replicate in-situ conditions offshore Hong Kong, where suction caissons to found wind turbine tripod will be installed (Gaudin & Randolph, 2008). 100 cst silicon oil (e.g. 100 times more viscous than water) was used as the pore fluid to model correctly the pore pressure regime resulting from environmental cyclic behaviour whose frequency could not be scaled correctly by a factor of 100. The silicon oil, therefore allowed proper scaling of the pore fluid. Unsaturated conditions may also be simulated in centrifuge models as demonstrated during the European project NECER (Rezzoug et al., 2000). In parallel to the increasing sophistication of reconstitution techniques using natural soils, alternative materials have been developed, replicating the geotechnical properties of natural soil, with the aim to either simplify and increase the reliability of the

reconstitution techniques or provide particular features to maximise testing outcomes. Artificial soil made from mineral ingredients, high boiling liquid and solvent and exhibiting shearing behaviour identical to soil, has been developed in order to overcome the existing limitations of reconstitution techniques (Sarma, 2006). However, the very good control of the reconstitution process of natural and kaolin clays and the improvement in soil characterisation techniques have dramatically limited the need of such artificial soil. Another artificial material which is seeing much use is the “transparent soil” (Iskander et al., 2001). By mixing flumed silica with paraffin oil and white spirit, one can obtain a transparent slurry, which upon consolidation, exhibits a strength profile similar to clay materials. The process permits the visual observation, through a window, of the offshore structure investigated. This is similar to the PIV technique, but it also permits accurate measurements of the loads applied, which is not possible with the PIV technique due to the contact of the structure with the observation window. An example offshore application using observation, through transparent soil, has been the investigation of the behaviour and trajectory of an embedded plate anchor (Figure 6) during keying and pullout (Song et al., 2009). Improved in-flight soil characterisation techniques may also be partly responsible for the improvement in sample reconstitution techniques. Reduced scale versions of cone penetrometer, piezocone and vane apparatus, identical to the in-situ tools, were used successfully since the early day of centrifuge testing to characterise soils. The last two decades have, however, seen the development of new characterisation tools in the centrifuge. This has proved to provide more accurate and more comprehensive information about the soil properties. This is notably the case of the T-bar penetrometer (Stewart & Randolph, 1994), initially developed for the centrifuge, which is now widely used offshore. The T-bar provides an accurate measurement of the soil undrained shear strength and offers a valuable alternative to cone penetrometer devices. The latter requires various corrections and have a disadvantage of a low ratio of the change in resistance to the ambient hydrostatic pressure acting on the cone tip (Randolph et al., 2005). In addition, more information can be extracted from T-bar results as illustrated by recent research. By performing successive cyclic sequences, one may access the strength degradation rate, the soil sensitivity and the post reconsolidation strength (Hodder, et al., 2010, Zhou & Randolph, 2009). By performing tests at various penetration rates, one may access information about the consolidation characteristics of the soil and the strength enhancement due to viscous effects (Chung et al., 2006, Lehane et al., 2009). More recently, a new analysis technique, based on the T-bar, has been developed which allows the accurate determination of the shear strength in the first meter of the

164 © 2011 by Taylor & Francis Group, LLC

Figure 7. O-bar developed at C-CORE.

Figure 6. Anchor model in transparent soil (after Song et al., 2009).

seabed, accounting for buoyancy of the apparatus and surface effects (White et al., 2010). This analysis technique is particularly relevant to assess the soils shear strength for soil-pipeline interaction. This is developed further in section 6.3. C-CORE has recently developed an O-bar for the assessment of near surface strength by reducing the bar diameter to 2.4 mm while increasing its length to 110 mm using the form of an annular ring (Figure 7). Other full flow penetrometer such as the ball penetrometer (Kelleher & Randolph, 2005) and more recently the piezoball penetrometer (Boylan et al., 2010b) have also been developed and have started to be introduced offshore (Figure 8). The centrifuge environment, with perfect control of the testing conditions, is the appropriate environment to develop and validate new soil characterisation tools. Near future will surely see new devices arising, such as the “doughnut” shearing device developed to measure peak and post peak pipe-soil friction factors and the associated pore pressures (Yan et al., 2010). Soil containers have also benefited from some technological developments. Thermally controlled strongboxes (Phillips et al., 2002) and ice generators (Lau et al., 2002) have been introduced to investigate the behaviour of offshore structures in artic regions. These

Figure 8. Piezoball penetrometer developed at COFS to investigate strength and consolidation properties of soft soils (after Boylan et al., 2010b).

are based on the similitude principles for soil freezing established by Yang & Goodings (1998). 4.5

165 © 2011 by Taylor & Francis Group, LLC

Models and instrumentation

The general improvement in instrumentation and sensing technology has indeed benefited centrifuge technology. Laser sensors offering high resolution are progressively replacing linear displacement transducers. The miniaturisation of commercially available load cells, pore pressure and total pressure transducers results in more accurate and more numerous measurements. This is giving access to information pivotal to understand detailed phenomena. One relevant example is the use of contact stress transducers to measure the radial stress changes on the outside of the 0.4 mm thick skirt of a suction caisson during installation, consolidation and subsequent extraction (Chen & Randolph, 2004). Measurements provided direct insight onto the skin friction of the caisson skirt, which was previously back calculated from penetration and extraction resistance. Jeanjean et al. (2006) used a double walled suction caisson to separate internal and external skin friction by nesting two stainless steel tubes and measuring the

Figure 10. Example of model suction embedded plate anchor (30 mm high, 53 mm wide) replicating all key features of prototype anchors, including keying flap and flap hinge eccentricity (from COFS internal report).

Figure 9. “Stroud” type load cell with protective sleeve removed (after Daiyan et al., 2010).

loads applied both to the inner wall and on the total system. There was a 0.4 mm gap between the outside wall and the inside wall of the anchor. Specific instrumentation for centrifuge applications has also been developed. For instance, Take & Bolton (2002) and Oung & Bezuijen (2002) have developed pore pressure tensiometers which overcome some of the limitations of the commercially available pore pressure transducer Druck PDCR-81. Daiyan et al., (2010) have used a miniature sealed ‘Stroud’ type load cell to measure the axial, transverse and moment loads transmitted through a pipeline subject to oblique loading in sands and clays (Figure 9). The use of bender element to measure in-flight shear wave velocity of a soil sample is also becoming regular practice (Brandenberg et al., 2006). The manufacturing of models has also improved significantly, with more precise foundation and anchor models now commonly available. An example of this improvement relates to drag and plate anchor models which feature particular details such as shank and fluke geometry and keying flap. The use of devices to replicate suction installation of caissons (Clukey & Phillips, 2002, Chen & Randolph, 2007b, Raines et al., 2005, Jeanjean et al., 2006, Colliat et al., 2010) or suction embedded plate anchors (Gaudin et al., 2006c), has also become common practice (Figure 10).

5 THE CONTRIBUTION OF CENTRIFUGE TESTING/MODELLING The contribution of centrifuge testing and modelling to the design of offshore structures and the understanding of offshore structure soil interaction has been addressed successively by Murff (1996), Garnier (2004) and Gaudin et al. (2006a). With the continuous development of new centrifuge technology and the great state of flux exhibited by

the offshore industry, continuously moving towards deeper waters and unusual soil conditions, it is anticipated that new areas will emerged in the near future. While the offshore industry will most likely use more and more centrifuge modelling techniques and benefits to its outcomes, it is as important to appreciate that the offshore industry will contribute significantly to the development of new modelling and testing techniques. The following provides a review and some additional examples of where centrifuge testing has enhanced offshore foundation engineering. 5.1

As opposed to field tests, centrifuge modelling provides homogeneous and well characterised soil conditions, known boundary conditions, accurate measurements of parameters and repeatable testing conditions. Therefore centrifuge testing offers reliable performance data for a given idealised problem which can be used to calibrate analytical and numerical models. These can subsequently be applied to specific field problems, which by nature are more complex. Some relevant examples, outside those already presented by Murff (1996), Martin (2001) and Gaudin et al. (2006a), include: 1. The calibration of numerical model for the design of the foundations of the Rion-Antirion bridge in Greece (Garnier & Pecker, 1999), a novel type of foundation including a sand raft resting on pilereinforced marine clay and subjected to seismic loading in addition to the more common environmental loadings. 2. The calibration of the geotechnical model used to design the pier foundation of the Confederation bridge in Canada (Phillips et al., 1998). The Confederation Bridge main pier foundation design

166 © 2011 by Taylor & Francis Group, LLC

Provide performance data to calibrate analytical or numerical models

was carried out using the LRFD approach and a probabilistic assessment (Becker et al., 1998). A reliable assessment of bearing resistance requires that the failure mechanism that is likely to develop in a foundation material be known and understood. There is limited field data regarding the performance of structures founded on ring footings, particularly in an offshore environment where high horizontal loads and moments are generated. Therefore, centrifuge model tests of the main pier foundations were carried out to investigate potential mechanisms of failure, to examine how these mechanisms vary with changes in foundation and loading conditions, and to evaluate the foundation design analysis carried out as described above. This testing also helped to reduce some of the inherent uncertainties associated with the analyses, such as limitations in constitutive models, and to improve the fundamental understanding of the geotechnical aspects of foundation design under high horizontal loads and associated eccentricities. 3. The calibration of the numerical model developed to design the sloping sea wall of the pier foundations of the Gwang Yang bridge in Korea against the ship impacts (Gaudin & Colwill, 2007). The Gwang Yang bridge is a 1.28 km long suspended bridge in South Korea. Its piers are funded on a 60.5 m long by 36 m wide raft resting on rubble, soft marine clay and sand compaction piles. In turn, the raft is supported by a network of 38 piles, 32.6 m long, embedded in the rock bed. The piers are protected against potential ship impact by a sloping sea wall, whose geometry was initially designed from empirical analytical models. In order to calibrate the numerical model used to finalise and validate the design, centrifuge tests were performed on an idealised model, where the rubble, soft marine clay and sand compaction piles were replaced by loose sand. Other features of the prototype such as the embedment of the piles into the rock bed through a layer of soft marine clay and the geometry and velocity of the bow and energy realised by the ship during impact were accurately replicated. The model was heavily instrumented and provided performance data such as the total pressure and the pore pressure in the front and at the invert of the raft, displacements and acceleration of the raft, and 3D profiling of the geometry of the sloping sea wall after impact. These data were used to calibrate a finite element model, which was subsequently improved to account for the properties of the rubble and sand compaction piles. The FE model was then used for design (Figure 11). 5.2 Providing qualitative insights into soil-structure interaction and mechanisms This aspect is particularly important when novel concepts or unusual conditions are encountered. Understanding the structure behaviour, observing the failure mechanism taking place, is a pivotal step into the

Figure 11. Model of the GwangYang bridge pier foundation and protecting seawall. (a) Footprint left by the impact of a ship at 30 degrees, and (b) comparison with the numerical model (b)(after Gaudin & Colwill, 2007).

development of sound design methodology. Some of the most noticeable breakthrough includes: 1. The identification of the shallow failure mechanism for gravity platform under horizontal and vertical cycling loading in comparison to the much deeper mechanism resulting from monotonic loading (Craig & Al-Sauodi, 1981). 2. The identification of the flow failure (Figure 12) taking place during spudcan penetration, as opposed to a cavity wall failure (Hossain et al., 2005). This has resulted in the amendment of the SNAME (2008) guidelines. 3. The observation during pipeline lateral breakout of the growth of a soil berm in front of the pipe and of the suction developed at the rear of the pipe, governing both the trajectory of the pipe and the peak breakout resistance (Dingle et al., 2008). This aspect is developed in more details in section 6.5. 4. The observation of the behaviour of the keying flap equipping suction embedded plate anchors. The flap was designed to rotate during keying and hence limit the loss of embedment. Centrifuge results demonstrated that the flap does not rotate (Figure 13) due to the rotational mechanism of the soil, resulting in the soil applying a bearing pressure at the back of the flap (Gaudin et al., 2010b).

167 © 2011 by Taylor & Francis Group, LLC

Figure 13. Plate anchor pulled out vertically. Demonstration of the non-activation of the keying flap during anchor keying (after Gaudin et al., 2010b).

with full-scale measurements to provide confidence to the end-users that this is a viable approach.

5.3 Validate the structure design for a particular site or a specific design approach By allowing the use of natural soils, sampled in-situ, by applying complex loading sequences directly relevant to design and by using a models replicating precisely prototypes, centrifuge modelling offers a valuable tool to validate or justify specific offshore structures. Among the relevant examples are:

Figure 12. Observation of the back flow mechanism during spudcan penetration (after Hossain et al., 2005).

Centrifuge modelling has also provided an efficient and cost-effective complement to conventional ice tank modelling to simulate the deformation of level ice and rubble. Lau et al. (2002) presented an overview of the ice-structure interaction research conducted at C-CORE. They discuss the advantages of centrifuge modelling of ice problems in terms of the scalability of ice properties, the controllability of test environments, and the quality of test data. There was an excellent agreement between data obtained from the centrifuge, an analytical algorithm and other conventional refrigerated model basins for a range of ice conditions. These included level ice, ridges and rubbles interacting with structures including conical piers with level ice sheets; unconsolidated ice rubble and ridges with cylindrical structures; and consolidated rubble ice with conical structures. Due to the small size of the model required, centrifuge modelling also has the advantage of simulating and maintaining a controlled test environment at a fraction of the cost of conventional tanks. Centrifuge modelling is relatively new to the ice engineering community and additional comparisons are still needed

1. The investigation of the response under cyclic horizontal loading of the Maari platform in New Zealand. Operated by OMV, the Maari platform is founded on a rectangular skirted mat (Figure 14a), installed in sand overlaid by impermeable silt (Gaudin et al., 2006b). Tests were used to investigate both the feasibility of suction installation through layered soil, and the performance of the platform under cyclic horizontal loading. Results demonstrated that suction installation was achievable, with limited silt plug uplift, provided that sufficient direct load is applied and the pumping rate is high enough to result in a fast installation, limiting the amount of plug uplift. Similar observations were made on another offshore project in Gulf of Mexico where suction installation on sand overlaid by clay was required (Watson et al., 2006). Results for the Maari platform, under horizontal cyclic loading highlighted the very stiff response of the top silt layer and the very limited vertical displacements experienced by the platform, even after several packages of cyclic loading (Figure 14b). 2. The investigation of the performance of a dolphin made of sheet pile walls for the protection of the Korena Incheon bridge foundation against ships

168 © 2011 by Taylor & Francis Group, LLC

Figure 14. (a) The model Maari platform and (b) Accumulation of vertical displacements under cyclic horizontal loading (after Gaudin et al., 2006b).

impact (Bezuijen et al., 2007). A total of 31 tests were performed at 200g investigating the response of the dolphin under static and dynamic tests loadings. Other parameters investigated included the density of the soil, the diameter of the dolphin and the location of impact. The results show that for dynamic tests (as opposed to static tests) suction was generated underneath the dolphin increasing its stability. 3. The validation of design of the Genesis Spar platform installed in the Gulf of Mexico, and operated by Chevron. The 70 m long piles anchoring the platform were modelled in a series of centrifuge tests performed at LCPC to investigate notably the performance of the piles under combined vertical and horizontal loading, and cyclic storm loading. The bending moment profiles (see Figure 15a) along the piles were monitored to optimise the pile thickness and to develop p-y reaction curves (see Figure 15b) these were subsequently used to validate the pile design.

Figure 15. (a) Bending moment (MN.m) vs. depth (m) recorded during the Genesis pile loading tests ranging from 10 to 23 MN, and (b) experimental P-y reaction curves derived from the bending moment profiles (depth ranging from 21 m to 58.8 m). (From internal LCPC report).

5.4

The very good control of the testing conditions, the possibility of measuring a significant number of parameters, such as displacement, loads, pressures and even strains from digital imagery and the relatively low cost of centrifuge testing compared to field testing (even in reduced scale) make the centrifuge an ideal tool to develop new concepts and investigate the feasibility of particular foundations, which have not been tested in-situ yet. The modelling does not intend to replicate precisely a prototype structure, but rather puts emphasis on a particular aspect which requires validation or deeper understanding. Clukey & Morrison (1993) investigated the efficiency of multi-cell suction caissons as anchoring system for tension leg platforms (Figure

169 © 2011 by Taylor & Francis Group, LLC

Investigating the feasibility or developing new foundation concept

Figure 16. Model multi-cell suction caisson (source LCPC).

16). Tests were performed on kaolin clay and particular attention was paid on the potential for suction to be developed at the caisson lid invert to resist monotonic and cyclic tensile loads. Gaudin & Cassidy (2007) investigated the feasibility of suction-induced preloading of a skirted mat as an alternative to spudcans as a foundation for jack-up platform in intermediate water depth, where ballast tank may not provide enough preloading. Tests did not model any particular type of foundations, but focused on the preload level which could be achieved by suction and the drainage conditions within the foundation. Results demonstrated the feasibility of the concept but also the mechanism governing suction-induced preloading. It was notably observed, the absence of drained plug failure during the application of suction-induced preload, and the dominant effect of the consolidation time versus the preload level on the post preloading bearing capacity of the foundation (Figure 17). Another typical example of a novel design that has greatly benefited from centrifuge methods is the development of dynamic anchors, commonly called torpedo anchors. Torpedo anchors are usually 1 to 1.2 m in diameter and 10 to 15 m high. They are released from a height of 50 to 100 m above the seabed, achieving a free fall a velocity of 10 to 30 m/s, before impacting soft seabed sediments and embedding by 2 to 3 times their length. Key uncertainties relate to (i) the embedment depth resulting from the dynamic penetration and (ii) the soil setup after installation and the resulting holding capacity. Centrifuge techniques were particularly appropriate to conduct model tests, using the high g environment to achieve a fast free fall velocity from limited drop height. They played a significant role in both providing key insight about the soil-anchor interaction and establishing techniques to predict anchor embedment and anchor capacity. Early centrifuge work, reported notably by O’Loughlin et al. (2004) and Richardson et al. (2006) focused on anchor installation and highlighted the effect of the very high strain rate generated by the

Figure 17. (a) Hybrid foundation model with the suction caisson in the centre and the skirted mat around the caisson (b) Increase of bearing capacity due to suction induced preloading (after Gaudin & Cassidy, 2007).

penetration high velocity. This results in an increase in shear strength due to viscous effect, potential entrainment of water at the soil anchor interface and a decrease of skin friction. The necessity to account for a drag factor, in addition to the soil bearing factor, in calculating the anchor embedment depth was also demonstrated. More recent work, reported by Richardson et al. (2009) focused on the soil strength recovery after installation and the subsequent holding capacity (Figure 18). Results demonstrated the significant increase of capacity with time (by a factor of 5 from immediate extraction to full consolidation) and the significant time required to achieve full consolidation, compare to piles or suction caisson (up to 7 years in the case presented here). 5.5

As highlighted by previous examples, the range of benefits from centrifuge application to offshore

170 © 2011 by Taylor & Francis Group, LLC

Developing new design methods

the pile head lateral displacement may be estimated by:

Figure 18. (a) Typical model torpedo anchors, without fluke and with four flukes (b) Illustration of the gain in capacity with consolidation time after penetration. QS refers to static installation, DY to dynamic installation (after Richardson et al., 2009).

geotechnics offers the opportunity of providing valuable information at all stages of foundation design. Centrifuge modelling can also be an appropriate means to develop empirical design methods for some particular cases, where the phenomenon taking place are too complex to be captured by numerical or analytical models. One example is the development of a new design method for piles under lateral cyclic loading, which was poorly accounted for in common design guidelines. Centrifuge tests have been carried out by Rosquöet et al. (2007) and Rakotonindriana (2009), where up to 75000 lateral cycles were applied to model piles installed in dry sand. By normalising these cyclic loads by the maximum applied load Fmax and by introducing the amplitude of the load variation DF, Rosquoet et al., 2007 demonstrated that, for service conditions, the effect of the number of cycles n on

where y1 and yn are the pile head displacement under the maximum load Fmax , at respectively the first loading and at the nth cycle. Cyclic p-y curves were determined from bending moment measurements during loading (Figure 19). The degradation of the soil lateral resistance at different depths was then evaluated and P-multiplier coefficients r were established. These permits the determination of the cyclic p-y curves from the static ones (Table 1). Another typical example is the empirical method developed by Bienen et al. (2009) to determine the flow rate required to successfully extract spudcans in soft clay using on-bottom jetting (Figure 20a). The method was based on insights provided by a series of centrifuge tests, investigating various parameters such as the extraction rate, the jetting flow rate and the jetting pressure (Bienen et al., 2009; Gaudin et al., 2010c) and the demonstration that the efficiency of the jetting related to its capability to fill the gap at the spudcan invert created by its uplift. Hence, jetting flow was shown to be a dominant factor over jetting pressure, in contrast to generally accepted belief in the jack-up industry. A jetting methodology was then established based on (i) the pullout force resulting from the buoyancy of the jack-up hull, and a filling ratio f, defined as the ratio of the jetting flow rate to the product of the extraction rate to the spudcan area (i.e., the volume at the spudcan invert created by the extraction) (Figure 20b). The Pressure Ridge Ice Scour Experiment (PRISE) developed the capability to design pipelines and other seabed installations in regions gouged by ice, taking into account the soil deformations and stress changes, which may be caused during a gouge event (Phillips et al., 2005). This capability has been used to design offshore pipelines in regions such as the US Beaufort Sea and off Sakhalin Island, Russia. One of the main activities for the PRISE program was to assess the significance of subgouge deformations imposed on buried pipelines during ice gouge events. The PRISE joint industry research program conducted analytical, numerical, experimental and field studies. Comparison of field investigations on relict gouge events with centrifuge experiments demonstrated that subgouge deformations should be considered. Analytical and numerical models were developed to assess the gouge forces in sands and clays, based on the experimental failure mechanism observations. The implications of the models on ice keel ablation, spoil heap development and the development of subgouge deformations are discussed by Phillips et al. (2005). Eulerian based finite element models for clay have also been validated against the centrifuge model test results (Phillips et al., 2010).

171 © 2011 by Taylor & Francis Group, LLC

Figure 19. Cyclic p-y curves at different depths obtained from one-way lateral loading tests (after Rosquoet et al., 2007).

Table 1. P-multipliers r at different depths for determining the cyclic p-y reaction curves (after Rosquoet et al., 2007). Depth z

P-multiplier r

0 < z < 1.5B 1.5B < z < 3B 3B < z < 5B

r = 0.7 − 0.12 DF/Fmax r = 0.94 − 0.058 DF/Fmax r = 0.97 − 0.029 DF/Fmax

5.6

5.7

Characterising in-situ soils

The improvement of soils reconstitution methods allows the replication of complex soil stratigraphy using natural soils exhibiting the same properties (such as permeability, compressibility, void ratio, water content, unit weight and shear strength) as found in the field (see development in section 1.3). The centrifuge offers the possibility of performing a large number of tests at a reduced cost, using characterisation tools identical to the ones used in-situ. Also, in addition, the techniques provides extra information by using newly developed characterisation tools not yet commonly used in-situ. One such example is the recently developed piezo-ball penetrometer presented in section 4.4, for characterisation of soft soils. The test, in addition to the measurement of the penetration resistance (and hence undrained shear strength), provides information on the drainage characteristics of the soil and potentially the consolidation behaviour (Boylan et al., 2010b). With the acquisition of large soil samples taken with box core becoming more standard practice, it

172 © 2011 by Taylor & Francis Group, LLC

is also possible to perform typical soil characterisation tests in a relative undisturbed state (White & Gaudin, 2007). Centrifuge testing helps augment these tests. Providing data for assessment of fatigue for conductors and SCRs

Designing for fatigue of conductors (uppermost outer pipe in a well) and Steel Catenary Risers (SCRs) is an important design challenge in deepwater. The interaction of the conductor and riser with the soil is an important part of both these fatigue assessments. Jeanjean (2009) describes a case where the API p-y curves were modified based on finite element analyses and centrifuge testing to satisfy fatigue requirements for a number of offshore conductors. For this particular case the small displacement lateral soil response was critical for predicting the expected fatigue life. Figure 21 shows a comparison between theAPI recommended lateral soil response and the response derived from FE analyses and centrifuge testing. In this problem a stiffer soil response reduced fatigue life. Cyclic centrifuge tests did show some reduction in the soil stiffness. However, based on the design methods used to assess fatigue, during the unloadreload soil response appropriate to cyclic loading, the soil stiffness was never less than the tangent stiffness from the monotonic test results used in the analyses (Jeanjean, 2009). The fatigue of an SCR in the soil touchdown point region (Figure 22) depends on both small and large amplitude displacements. Hodder et al. (2008) describes a series of centrifuge tests to investigating

Figure 20. (a) Model spudcan with jetting in action and (b) conceptual chart of jetting extraction efficiency. The chart indicates, for a given extraction load, the required flow rate for a successful jetted extraction (after Gaudin et al., 2010c).

the response of a small portion of pipe in the soil touchdown point region. The tests show that the soil stiffness can be dramatically reduced by the large displacements incurred from the riser motions, especially when

173 © 2011 by Taylor & Francis Group, LLC

the riser pipe separates from the seafloor. However, a significant portion of the stiffness loss is regained when the motions stop and the soil has an opportunity to recover from thixotropic and consolidation effects

Figure 21. Comparison of API p-y curves versus p-y curves obtained with centrifuge testing and FE analyses (after Jeanjean, 2009).

Figure 22. Schematic of offshore facility and SCR interacting with soil in the touch down area.

(Hodder et al., 2009). The large motions of the SCR will also enhance trench formation which will also affect fatigue life. Centrifuge tests to investigate the response of a complete riser through the touchdown point region are now being performed at C-CORE. A special actuator has been developed for these tests to simultaneously provide both heave and surge motions to the riser. The motions are applied to both these motions at two frequencies to capture the overall riser motions expected in the field. Comparisons of trenches observed in the field and in the centrifuge are shown on Figure 23 and Figure 24. 5.8

Figure 23. SCR trench developed in centrifuge test.

distribution of loads. One such example is the work performed by Gaudin & Landon (2008) and Gaudin et al., 2009b who investigated the efficiency of a rock cover to protect a buried offshore pipeline against the dragging of anchors. The assessment of the clearance between the anchor and the pipeline over a series of centrifuge tests resulted in the determination of frequency of failure of the pipeline which was subsequently introduced into a probabilistic analysis on the frequency of hazard affecting the pipeline. 6

Providing data for probabilistic analyses

One recognised advantage of centrifuge modelling is the ability to conduct parametric studies in a well controlled environment. It is therefore possible to focus the study on a particular parameter, such as the natural variation of soil characteristics or the statistical

6.1

Suction caisson technology development

The history of the development of suction caisson as anchoring system for deep water solutions illustrates particularly well the contribution of centrifuge

174 © 2011 by Taylor & Francis Group, LLC

EXAMPLE OF CONTRIBUTIONS

Figure 26. Evolution of penetration resistance and external radial total stress of a suction caisson in centrifuge during jacking and suction installation (After Chen & Randolph, 2007a).

Figure 24. SCR trench in field.

Figure 25. Results of static pullout tests. Comparisons between in-situ 1g tests and centrifuge tests performed on natural (Lysaker) and artificial (kaolin) clays. (after Morrison et al., 1994).

testing and modelling in investigating behaviour and validating design procedures throughout their development. Centrifuge results have impacted installation procedure and assessment of the capacity, first under monotonic vertical loading and then under combined cyclic loading. In contrast to the development of driven piles technology for offshore structures, which was primarily based on a series of 1-g model tests, suction caisson technology has achieved significant advancements through centrifuge testing. The initial tests, however, for suction caissons used in deepwater were in-situ 1-g tests for the Snorre TLP in the North Sea. (Dyvik et al., 1993). These tests were subsequently replicated with centrifuge tests at LCPC (Morrison et al., 1994). When the undrained shear strength profile is well simulated, the load-displacement behaviour between the in-situ and centrifuge tests are very close, both in peak resistance and stiffness (Figure 25), and provided added confidence in the use of centrifuge testing for suction caisson applications.

Additional testing for TLP applications (Clukey & Morrison, 1993, Clukey et al., 1995) provided additional information on suction caisson behaviour under both monotonic and cyclic loads for TLPs, notably about the cyclic ratios resulting in failure and about the increase of post-cyclic static capacity. With ever increasing water depths, the design trends moved from multi-cell configurations systems, to single cell arrangements with the mooring line attached about 2/3 down the side of the caisson to enhance the holding capacity. The utilization of these single cell suction caissons expanded considerably as the industry moved into water depths too deep for conventional offshore pile driving. Centrifuge testing at a number of facilities then began to provide important information on the behaviour of these foundations both during installation and under a variety of loading conditions. Andersen et al. (2003) showed that suction caissons could potentially be installed to over ten times their diameter without failing the internal soil plug. Clukey & Phillips (2002) described monotonic tests which showed a potential for the reduction in external skin friction for suction installed versus jacked in caissons. Subsequent tests by Raines et al. (2005), however, showed no difference in external skin friction for suction versus jacked installed suction caissons. Similar results were obtained by Chen & Randolph (2007a) (see Figure 26) and Jeanjean et al. (2006). The latter used a double walled suction caisson to separate internal and external skin friction. Their results did show, however, a reduction in external skin friction versus API recommended values for driven piles. The reverse end bearing observed by Jeanjean et al. (2006) had a peak bearing capacity factor (Nc ) of 12 at large displacements and 9 when the external skin friction and total holding capacity reached peak values. Similar magnitudes of design parameters were obtained from research on suction caisson response carried out by Chen & Randolph (2007a). More recently, the effect of stiffeners in both the installation resistance and the pullout capacity has been investigated by Westgate et al. (2009) and Colliat

175 © 2011 by Taylor & Francis Group, LLC

verified the effectiveness of design tools in predicting suction caisson holding capacity in stiff overconsolidated soils in the Gulf of Mexico. It is most noteworthy that the results were obtained in a time frame, at a cost and with a level of details that could not have been matched by field testing alone. 6.2

Figure 27. Effects of stiffeners and consolidation on normalized pull-out capacity (after Colliat et al., 2010).

et al. (2010) at LCPC. The former investigated various stiffener configurations in kaolin clay, while the latter, using in-situ deepwater Nigeria clay, focused on the effect of consolidation time following installation, for caissons with and without stiffeners. Results, from both studies, demonstrated the significant contribution of the stiffeners to the pullout capacity. For undrained pullout, the increase ranges from 20% shortly after installation, up to 33% when full consolidation is permitted (Figure 27). Combined loading was investigated by Randolph et al. (1998), who described tests that demonstrated the potential for gapping behind the caisson, in soft carbonate sediments, due to interactions between horizontal and vertical loadings. Clukey et al. (2003), further investigated combined loading and demonstrated that limit analysis techniques could be used to predict these interactions. Clukey et al. (2004) also described a combination of independent centrifuge test programs at UWA and C-CORE that showed that reverse end bearing could be maintained for several months without reduction in capacity from pore water pressure dissipation. These results were also further corroborated with finite element analyses (Clukey et al., 2004). These results were important in assessing the performance of suction caisson for long term loop current load. As demonstrated above, the use of centrifuge techniques from various centres in the world, in collaboration with offshore operators have resulted in significant breakthrough in understanding the behaviour of suction caissons, both during installation and under various conditions of loadings. These various research programmes had significant impact in operators’ practice, both validating and optimising innovative designs. Hence, centrifuge model tests formed part of the design validation process for applications in Australia, off the west coast of Africa (Randolph et al., 1998), as well as in the Gulf of Mexico. Jeanjean et al. (2006) also described results obtained at the University of Colorado, Boulder, which

1. Accelerated time frame. Centrifuge modelling and testing require a limited volume of soil, accelerating sample preparation. For soft soil, the process may be further accelerated by in-flight self-weight consolidation. Similarly, testing sequences are considerably shortened while ensuring the correct drainage conditions. This allows collection of a significant number of data in a short time frame and at a reasonable cost. 2. Accurate loading sequences. New motion control techniques permit accurate replication of complex loading sequences, including cyclic vertical and/or horizontal motion under either load or displacement control. This is particularly relevant when mimicking the complex motion of the pipe at the touchdown zone during laying, or the large deformations resulting from lateral buckling.

176 © 2011 by Taylor & Francis Group, LLC

Pipeline-soil interaction

Offshore pipelines are laid on the seabed and used as flowlines or trunklines tied back to shore. They are becoming more and more common as oil and gas extraction is taking place in deeper and more remote areas. Geotechnical design procedures for offshore pipeline and risers have not yet reached the maturity exhibited by design procedures for piles, shallow foundations and anchoring systems. Reasons are multiple and relate mainly to the unknown geometry of the problem (the final embedment of the pipe is uncertain), the uncertainties of the soil conditions (both difficult to assess at very shallow depth and significantly affected by the installation process) and the very large deformations (and associated post-failure behaviour) the pipeline may experience (notably during controlled or unexpected lateral buckling). Whilst the on-bottom stability of pipelines (i.e. its stability against hydrodynamic forces) has been extensively investigated since the early 60s, notably in the USA, lateral buckling and axial walking (due to changes in internal pressure and temperature in the pipeline) have only emerged recently as a major research topic. Significant advances have been made over the last 5 years, using data from 1g model tests (performed notably in Cambridge, Cheuk et al., 2007, and NGI, Dendani & Jaeck, 2007), in-situ tests (notably using the SMARTPIPE device developed by Fugro, Jacob & Looijen, 2008) and also centrifuge tests. While each method has its own advantages and disadvantages (see Hill & Jacob, 2008 for a comparison of each method as applied to pipelines), centrifuge tests have certainly boosted knowledge of pipeline-soil interaction and generated significant breakthroughs. The main reasons are:

3. Use of natural soils. By requiring a limited volume of soil, it is possible to use in-situ soil with reasonable supply costs. The use of natural soils for model tests is maybe more important for pipelines than for other geotechnical structures, because of the heavy remoulding and the large deformations experienced by the soil. This may trigger specific behaviour that would not be captured by common artificial laboratory soils. 4. Accurate seabed characterisation. The good control of the soil reconstitution process results in a homogeneous sample. By using standard soil characterisation tools (such as the T-bar) or dedicated ones (such as the O-bar) in controlled conditions, it is possible to determine the soil characteristics accurately. 5. Enhanced instrumentation. By using image acquisition systems and pore pressure measurements, in addition to the measurements of load and displacements, one gains access to particular features of the pipe-soil interaction, such as the development of lateral berms during buckling, the creation of a trench during dynamic laying and the drainage conditions around the pipe during the pipe motion. All this information is pivotal in understanding and describing pipe-soil interaction. 6. Case-specific study. Centrifuge testing can be used to provide insight in particular issues related to pipeline design. This includes site-specific storm loading conditions or geohazard conditions, such as scarp crossings. The rest of this section aims to illustrate by a few relevant examples, some of the insights and breakthroughs, obtained from both centrifuge testing performed to gain insight into specific aspects of pipeline interaction, and from centrifuge modelling performed to assist in the design of specific pipeline projects. It is not a state-of-the-art review of pipeline behaviour and design. Such reviews have been presented by Cathie et al. (2005) and White & Cathie (2010).

6.3 Seabed characterisation Improving the reliability of design predictions of pipeline embedment during laying and lateral buckling relies on accurate measurements of the seabed characteristics in the upper 0.5 m of the seabed, which is the zone relevant to pipeline-soil interaction. The T-bar penetrometer, originally developed for the geotechnical centrifuge (Stewart & Randolph, 1994) is increasingly favoured in the field for characterising soft soils. The interpretation of T-bar penetration tests to assess the strength of soft seabed soils in the upper 0.5 m can be refined beyond the use of a constant NT-bar = 10.5 factor (Stewart & Randolph, 1994). These refinements have been described by White et al. (2010) and account for the soil buoyancy and the reduced bearing factor arising from the shallow failure mechanism mobilised prior to full flow of soil

© 2011 by Taylor & Francis Group, LLC

Figure 28. Example the influence of near surface and buoyancy correction on the determination of the strength at shallow embedment of lightly overconsolidated clay. (after White et al., 2010).

around the bar. These corrections, which are marginal for assessing the strength of clay at moderate depths (>1 m) can be significant at shallow depth in soft materials. White et al. (2010) demonstrated that the omission of these effects may result in an underestimation of the shear strength in the depth range relevant to pipeline analysis (Figure 28). More recently, the use of T-bar penetrometers has been extended to investigate in the centrifuge phenomena relevant to pipeline behaviour, including the successive cycles of consolidation and remoulding experienced by the soil (Hodder et al., 2010) and the potential entrainment of water (and consequent swelling) during cyclic events. Figure 29 illustrates the loss of strength (expressed as the ratio of the remoulded shear strength su,r to the intact shear strength su,int ) experienced by different soils during cycles of remoulding for artificial clay and naturals clayey silts typical of the North West Shelf region, off the coast of Australia. The most striking feature of this cyclic test is the ten-fold reduction in strength during cycling for the Western Australian carbonate silt A in contrast to the two-fold reduction in strength for normally consolidated Kaolin. Further strength loss may be experienced by the soil as the pipeline remoulds and softens the soil, and water is entrained, increasing the moisture content of the soil. This is illustrated in Figure 30, which compares the strength loss for normally consolidated Kaolin clay resulting from deep cycles, from cycles 22 mm deep, breaking at the surface and from cycles 50 mm deep, breaking at the surface (using model scale units). The deep cycles do not break the surface, so water entrainment does not occur. For the kaolin test to 50 mm depth, the strength loss has been quantified at four different depths, from 10 to 40 mm, while for the kaolin

177

test to 22 mm, the loss of strength has been quantified at a depth of 20 mm. This test clearly indicates that the entrainment of water contributes to a continuous strength loss down to a value 5 times lower than the fully remoulded shear strength determined from deep cycles (without water entrainment). For typical normally consolidated kaolin, this strength reduction corresponds to an increase of water content by about 30%. The cycles to 50 mm depth indicate a similar trend with the magnitude of strength loss decreasing with depth. At 10 mm depth, the strength loss reaches a value similar to the one exhibited by the 22 mm cycles, but at a much faster rate. At deeper depths the degree of strength degradation reduces to eventually reach a value similar to the one obtained from deep cycles. As the T-bar penetrates deeper, the backflow of soil gradually limits the water entrainment. Even at very shallow depth, the effect of the water entrainment seems to reach a maximum, resulting in the soil still exhibiting some shear strength, even after a significant number of cycles. The simple examples presented above illustrate (i) the wide range of sensitivity exhibited by different types of offshore soils and (ii) the significant loss of strength resulting from water entrainment – a phenomenon likely to occur during pipe-soil interaction but which is not accounted in current design

approaches. In both cases, simple characterisation tests performed on in-situ reconstituted samples in the centrifuge may provide valuable and reliable information for design. 6.4

The knowledge of the pipe embedment is pivotal for subsequent on-bottom stability and lateral buckling design. The effect of dynamic laying on pipe embedment has been first observed by Lund (2000) who concluded on the necessity to account for pipeline laying history in subsequent on-bottom stability design, as it results in excessive embedment compared to a static laying process. During the lay process, an element of pipe moves through the touchdown zone, from an initial contact with the seabed to a stationary position, where the pipe weight is supported by an equal upwards seabed reaction force. The dynamic behaviour of the pipe through this process is complex and difficult to replicate. It is a function of the following parameters:

Figure 29. Degradation of strength during cyclic T-bar penetrometer tests (after Gaudin & White, 2009).

Figure 30. Strength loss at different depths in normally consolidated clay during deep and surface-breaking cyclic T-bar penetrometer tests (after Gaudin & White, 2009).

178 © 2011 by Taylor & Francis Group, LLC

Dynamic lay embedment

– The profile of stress concentration in the touchdown zone due to the catenary shape, resulting in a vertical load on the soil varying along the pipe and higher, at some locations, than the as-laid pipeline weight. These pipe stresses can be estimated, based on a structural analysis of the hanging pipeline using the planned lay tension and hang-off angle. – The horizontal and vertical oscillation due to the motion of the laying vessel, resulting in a sweeping and damping of the soil at the touchdown zone. It can be estimated from the analysis of the motion of the vessel under for the wave motion considered. – The number of oscillations during the entire lay process. It can be assessed, based on the estimated pipeline laying rate, the touchdown zone length and the wave-induced oscillation frequency. Figure 31 presents a typical pipeline setup used to model dynamic laying and lateral buckling. The pipe is modelled as a short section (with an aspect ratio of at least 6 to prevent any end effect) and may be sandblasted to achieve a particular roughness. It is connected to a VHM loading arm, used to measure the horizontal loads, via a shear load cell, insensitive to bending moment. The purpose of this load cell is to provide a very accurate reading of the vertical load applied (permitting a resolution of 1 N at model scale) that is essential for pipelines, which impose stresses on the seabed that are orders of magnitude lower than typical geotechnical structures. Additional instrumentation may include pore pressure transducers at the pipe invert, as shown in Figure 31. A purposely developed motion control system (De Catania et al., 2010) is used at COFS to (i) apply a targeted vertical load depending on the stress concentration profile in the touchdown zone, via a feedback loop on the axial load cell and (ii) apply lateral oscillation of varying amplitude, linked via a second feedback loop to the pipe embedment.

The following example presents centrifuge modelling performed recently to determine the final embedment of a flowline to be installed offshore Australia (Gaudin & White, 2009). In-situ soil was used and the dynamic motion applied accurately replicated the motions determined from a numerical analysis of the wave, vessel and pipeline motion. The pipeline was divided into six sections, from the initial contact point to a far away position when the pipe was considered to be unaffected by the laying motion. The six sections featured six different loading sequences. Each sequence included a cyclic vertical loading of specified amplitude (from 0 to 1.55 times

Figure 31. Typical pipeline model setup.

the as-laid pipeline weight Vlay ) concurrent with horizontal cyclic motion, also of specified amplitude, as presented in Figure 32. The first sequence, modelling the first segment of the pipe at the touchdown zone, features a specific cyclic motion in which the pipe was pushed into the soil to a specified load of 0.7Vlay before being lifted up until separation between the soil and the pipe occurred. The resulting accumulation of pipe embedment is presented in Figure 33 for each sequence as a function of the cumulative number of imposed cycles. The key observation from Figure 32 and Figure 33 is the much larger pipe embedment (about 0.24 m) compared to a static embedment (about 44 mm using the formulation presented by Randolph & White, 2008a), even accounting for the stress concentration. Other important observations are the suction developed at the pipe invert during the first sequence of loading, where the pipe was lifted up from the soil (which potentially affects the stresses in the pipeline), and the dominant effect of the lateral motion amplitude compared to the cyclic vertical load. The first sequence, which featured the largest amplitude of lateral motion, resulted in a deeper embedment that the subsequent sequences, which featured a higher maximum vertical load but smaller lateral motion amplitudes. In other words, for these lay conditions and this soil type, the action of pushing the soil to either side of the pipeline during lateral motion (and concurrently remoulding it) has a dominant effect on

Figure 32. Dynamic lay simulation. Cyclic vertical loading and the associated pipe invert trajectory (after Gaudin & White, 2009).

179 © 2011 by Taylor & Francis Group, LLC

Figure 33. Accumulation of pipe embedment resulting from the dynamic lay simulation of Figure 32 (after Gaudin & White, 2009).

the pipe embedment in comparison to any remoulding or penetration of the soil resulting from the vertical cyclic loading. These results stress the necessity to model accurately dynamic laying motion, in order to determine the pipe embedment, but also the remoulding of the soil resulting from the laying process, which are going to govern both on-bottom stability and lateral buckling. Further, the results demonstrate the capability of centrifuge modelling for simulating accurate laying motions, capturing particular soil behaviour features and delivering useful performance results. 6.5

Lateral buckling

To reduce the structural load resulting from thermal expansion, pipelines are permitted to buckle at targeted location, resulting in lateral displacements of the order of 10–20 times the pipeline diameter (Bruton et al., 2006). Whilst lateral bucking reduces the axial load on the pipeline, it also generates bending moment in the buckling zone, which can ultimately lead to local bending failure. The structural analyses used to assess lateral buckling in design require as input the soil resistance when the pipeline moves laterally. Unlike on-bottom stability design, overestimating the lateral resistance is not necessarily conservative as a high ‘friction’ factor can result in a more onerous structural load. The determination of an accurate friction factor is therefore pivotal for a safe and sound design. This requires the understanding of the soil behaviour at large displacements and through many cycles of loading, well beyond the point of failure, all features that can be well investigated by centrifuge methods. Two different examples are presented of centrifuge studies of lateral buckling. One of centrifuge

testing performed to understand mechanisms taking place during lateral breakout, and one of centrifuge modelling to provide friction factors for a particular soil and loading conditions, to assist in the design of a pipeline offshore Australia. In the first example a specific device was developed to observe and analyse, using PIV techniques, the deformation of the soil around the pipe by placing it behind a Perspex window (Dingle et al., 2008). In order to accommodate the large displacements occurring during lateral breakout, the camera was placed on a second actuator, which followed the displacement of the pipe, so the pipe remained constantly at the centre of the image. The corresponding setup is presented in Figure 34. Lightly over consolidated kaolin was used, and the model pipe was lowered onto the seabed to a targeted depth, hence ignoring any dynamic process and consequent soil remoulding. The test setup aimed at replicating the lateral breakout from known initial conditions. It deliberately ignored side aspects, which although important, are not believed to affect the qualitative response of the soil. Hence it qualified as testing rather than modelling. Dingle et al. (2008) presented a thorough analysis of the testing results, covering notably, the soil heave resulting from penetration and the assessment of the mobilised soil shear strength during lateral breakout. Two pivotal observations for heavy pipes (i.e. penetrating significantly into the seabed) were made, unknown at the time of project, regarding lateral breakout. First, the soil behaviour exhibited a brittle response, characterised by a sudden drop in resistance once suction at the rear of the pipe was lost. (Figure 35a) and resulting in the pipe rising up to shallower embedment. Hence, the peak lateral resistance is governed by the available tensile resistance at the rear of the pipe. Note that theoretical solutions for the two-sided mechanism

180 © 2011 by Taylor & Francis Group, LLC

Figure 34. Centrifuge testing of pipe lateral breakout (after Dingle et al., 2008).

and for the one-sided mechanism were derived from upper bound plasticity limit analysis by Cheuk et al. (2008) and Randolph & White (2008b), respectively, following the observations from the centrifuge testing. Second, the post peak lateral response is governed by the growth of a soil berm at the front of the pipe (see Figure 35b; this can also be seen in Figure 31 and Figure 34). Further analysis presented by White & Cheuk (2008) indicated that the lateral resistance could be expressed as a conventional frictional term plus a berm resistance term proportional to the volume of the berm. This finding has critical consequences for design as it can lead to an overall friction factor, combining both phenomena, higher than unity while conventional friction factors used for design lied in the range 0.2–0.8. The second example presents centrifuge modelling performed to provide design data (dynamic embedment and friction factors) to assist in the design of a pipeline offshore Australia (White & Gaudin, 2008). Natural soils sampled from the site were used and reconstituted so they exhibited the same strength characteristics and index parameters as the in-situ soil. The dynamic laying process was accounted for, replicating the change of vertical contact force along the touchdown zone during laying. It was followed by lateral breakout and a series of lateral sweeps. By replicating soil conditions and key phenomena of the soil-pipe interaction, and by aiming to provide data directly applicable for design, the project qualifies as modelling rather than testing. Detailed results are presented by White & Gaudin (2008). Some key observations, related to initial breakout in fine grained soils are presented in Figure 36. Load developments, pore pressure measurements at the pipe invert and vertical displacement with pipe lateral displacements from Figure 36, and the knowledge obtained from centrifuge testing from Dingle et al.

181 © 2011 by Taylor & Francis Group, LLC

Figure 35. Instantaneous velocity fields during pipe lateral breakout. Illustration of the tensile crack at the rear of the pipe (left) and formation of the soil berm at the front of the pipe (right) (after Dingle et al., 2008).

(2008), indicate the mechanism operating during initial breakout. Prior to separation the failure mechanism is two-sided, with soil being pulled behind the pipe in addition to being pushed ahead. The opening of a crack

Figure 36. Centrifuge modelling of lateral breakout (replotted from White & Gaudin, 2008).

at the rear of the pipe was demonstrated by the pore pressure measurements. After separation, a one-sided mechanism (e.g. with a berm forming at the front of the pipe) is operative. After breakout, the pipe rises to a shallow embedment. During this phase the lateral resistance also drops, reflecting the shallower failure mechanism that is mobilised. Once the pipe reaches a steady elevation the lateral resistance remains approximately constant and is created by shear at the base of the pipe and passive resistance against the berm of soil being pushed ahead. It is noteworthy that the pipe-soil interaction features observed during centrifuge testing, from which a design method has been refined (Bruton et al., 2008) were also observed during centrifuge modelling. Hence, the friction factor extracted from modelling and applicable to the specific site conditions could be confidently used for design using enhanced design methods developed from centrifuge testing. The two examples illustrate particularly well the combined contribution of centrifuge modelling and testing. The knowledge about a particular feature of soil-pipe interaction obtained from well controlled centrifuge testing helped interpret the data obtained from centrifuge modelling on specific soil and loading conditions. 6.6

Integration into design

The integration of physical model methods within the design process for geotechnical aspects of pipeline design such as lateral buckling has been discussed by White & Gaudin (2008) and is summarised in Figure 37. Several recent projects worldwide have included

182 © 2011 by Taylor & Francis Group, LLC

centrifuge testing of pipe-soil interaction as part of the design process, making pipeline design the offshore area where centrifuge methods are the most commonly used. As, for any other offshore site specific project, centrifuge modelling needs to be planned in the early stages of the project so sufficient volume of soil can be recovered from the seabed during the geotechnical survey. The modelling programme may include specific centrifuge soil characterisation test to complement results from the geotechnical survey and/or validate the soil sample reconstitution process. A preliminary design of the pipelines will have been undertaken, from which the likely pipe weights, laying motion and in-service behaviour can be extracted in order to provide the required parameters to develop the model and instrumentation and to specify the number of tests needed to provide the required information for design. The centrifuge modelling programme should then be tailored to replicate the anticipated lay process and the in-service pipe movements (with an appropriate range to capture uncertainties) aiming for the results to provide a suitable representation of the design situation. Best practice is for the reporting to follow the same pattern as the conventional geotechnical investigation, with a sequence of factual and interpretive reports. The factual report provides a full record of the modelling activity, typically presenting pipe load-displacement responses during laying, lateral breakout and lateral sweeping, as well as the development of excess pore pressure if measured. The interpretive report analyses the results in light of the calculation model that the design team is anticipating to use for the pipe-soil behaviour. Key values from the testing, such as the as-laid embedment and

Figure 37. Integration of pipe-soil centrifuge modelling within the pipeline design process (after White & Gaudin, 2008).

the friction factors at key stages, should be extracted from the load displacement responses, allowing the uncertain aspects of the calculation model to be calibrated for the particular site. The calibrated calculation model, which may have been originally developed from centrifuge testing, is then used to simulate the particular design situation. This generalisation is usually required because the scope of the centrifuge modelling cannot encompass all patterns of pipe movement during operation. These vary longitudinally along the pipeline. It is therefore necessary to interpolate between the patterns of pipe movement simulated in the model tests. Similarly, the project may involve various weights and diameter of pipeline, which cannot all be simulated in the centrifuge study. Depending on the complexity of the project, outcomes from centrifuge modelling may be completed using finite element analysis calibrated from the centrifuge results.

7

DIRECTIONS FOR THE FUTURE

7.1 Hybrid modelling Real time substructure testing is an emerging technology within civil engineering modelling (Blakeborough et al., 2001). In this approach, the physical model represents only part of a larger dynamic system, with the remainder being simulated numerically in real time, during the experiment.

This permits the modelling of geotechnical structures under realistic and complex loading sequences which will be updated in real time from numerical analysis looped to the motion control system. This is particularly relevant for offshore structures, such as pipelines or jack-up, where the structural response and the geotechnical response are often dependant. This will surely constitute a major breakthrough in physical modelling techniques in the near future. One such example is the hybrid actuator, currently in development at COFS for centrifuge application, to investigate the behaviour of jack-up legs and spudcan when penetrated close to an existing footprint. The actuator models a single leg of a jack-up unit and can apply or measure independent or combined vertical V, horizontal H and moment M loading. In order to include the structural response of the jack-up unit, a stiffness matrix is incorporated at the top of the actuator leg, governing the leg displacement response as a function of the loads developed within the leg. The stiffness matrix is updated in real time, depending on the structural response of the jack-up unit, using a finite element programme. This provides a more realistic modelling of the footprint interaction problem. An even more advanced development is the UKNEES project (Madabhushi et al., 2010). The UKNEES project is a distributed testing network, where testing facilities in one site can be controlled by numerical or physical models located in a different site. In addition, the network features tele-participation with video feed, allowing participants to communicate

183 © 2011 by Taylor & Francis Group, LLC

Figure 38. A pile featuring a wireless transmission module at its head: Mounted on the centrifuge actuator (left), details of the wireless transmission module (right) (after Deeks et al., 2010).

and share information simultaneously. In the example provide by Madabhushi et al. (2010), different parts of a bridge structure with shallow foundations were tested at different locations. The shallow foundation was tested in the Cambridge centrifuge facility. Loading on the foundation was achieved by an actuator, whose input came from physical and analytical models for the pier and sub-structure of the bridge deck tested at Bristol and Oxford. This type of network opens new possibilities for soil-structure interaction, with potential applications in the field of offshore geotechnics. A distributed testing network could for instance integrate the real-time numerical calculation of environmental loadings in one facility, which would feed the demand of a motion control system of a physical model in another facility. With the rapid development of networking facilities and the need for more realistic model testing, such networks will likely become more common in the future. 7.2

Enhanced instrumentation

The continuous development in computing and sensing technology will likely result in improved modelling techniques and new types of investigations. Two trends are identified, which will augment the benefit provided by centrifuge methods, namely image acquisition and intelligent sensors. As described in section 4.3, the development of PIV techniques has already constituted a major

breakthrough in physical modelling. Further improvements are expected with the increasing resolution and miniaturisation of cameras, leading to enhanced soilstructure investigations. High definition high speed cameras have started to be used more commonly for model testing at 1-g (Chow et al., 2010) and in the centrifuge (Boylan et al., 2009b), providing valuable information on high speed events. A good example is provided by Boylan et al. (2010a), where the usual geotechnical analysis is augmented by a geomorphologic analysis of submarine landslide, providing key information about the runout behaviour, and notably about the fluidisation of material and the transition from debris flow to turbidity current. This information could not be captured by usual data monitoring devices. The second trend relates to the development of intelligent sensors. One such example is the system presented by Deeks et al. (2010) to model pile installation by rotary jacking. The challenge associated with the continuous rotation of the pile during installation required the development of a wireless miniaturised data acquisition system to be located within the pile (Figure 38). The system is approximately 90 mm × 40 mm × 40 mm in size, and provides logging, conditioning and wireless transmission of up to 8 channels of data, eliminating the need of a data acquisition card on-board a computer. Such development opens new possibilities in term of data monitoring opening the field for new type of investigations.

184 © 2011 by Taylor & Francis Group, LLC

8

CONCLUSIONS

The paper presented a review of typical offshore centrifuge applications, with a particular emphasis on the contribution of centrifuge methods to offshore design and the benefit of collaboration with industry. The paper lists the advantages and disadvantages of the use of centrifuge methods. It details the difference between centrifuge testing and modelling, with an aim to assist potential centrifuge users and existing modellers in optimising the outcomes of their model tests by defining why, how and when centrifuge methods should be used. The authors advocate the use of centrifuge methods as a valuable tool to assist the offshore industry in developing and designing solutions for a wide range of geotechnical problems. Indeed, centrifuge modelling and testing should not been considered as the unique tool to be used, but as a particular one whose outcomes may be maximised if integrated in a global approach. It is believed that while technological and scientific developments will, without doubt, increase the utility of centrifuge methods, the most spectacular improvement will be in understanding centrifuge methods contribution and their integration at key stages into a global design procedure, incoporating in-situ, numerical and analytical methods. ACKNOWLEDGEMENTS Centrifuge modelling and testing are experimental techniques which rely upon the expertise and dedication of technical staff. The authors would like to gratefully acknowledge the contribution of the technical teams at COFS, C-CORE and LCPC which were invaluable in carrying out successfully the projects presented in this paper. The authors would also like to thank BP for allowing the 2nd author to participate on this paper and funding several studies discussed herein. Progress of centrifuge techniques have greatly benefited from industry incentive and support. The contribution and financial support of the various industry partners involved in the projects presented in the paper are gratefully acknowledged. Fruitful and thought provoking discussions with various colleagues have helped shape this paper. The authors would like to thank particularly Prof. Mark Cassidy, Prof. Dave White and Dr Philippe Jeanjean for their contribution. REFERENCES Andersen, K., Jeanjean, P., Luger, D., Jostad, H.P. 2003. Centrifuge tests on installation of suction anchors in soft clays”. Proc., Intern. Symposium, Deepwater Mooring Systems, Houston, TX. Baba, S., Miyake, M., Tsurugasaki, K., Kim, H. 2002. Development of wave generation system in a drum centrifuge. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 265–270.

Becker, D.E., Burwash, W.J, Montgomery, R.A., Liu,Y. 1998. Foundation design aspects of the Confederation Bridge. Canadian Geotechnical Journal, 35(5), 750–768. Bezuijen, A., Kim, J.H., Schaminée, P.E.L., Lottum H. Van 2007. The protection efficiency of dolphins tested in a geotechnical centrifuge. Proc. 17th Intern. Conf. Offshore and Polar Eng., Lisbon, Portugal, 2, 1609–1625. Bienen, B., Gaudin, C., Cassidy, M.J. 2009.The influence of pull-out load on the efficiency of jetting during spudcan extraction. Applied Ocean Research, 31, 201–211. Blakeborough, A., Williams, BM.S., Darby A.P., Williams, D.M. 2001. The development of real-time structure testing. Earthquake Eng. Struct. Dyn. 30, 431–448. Boylan, N., Gaudin, C., White, D.J., Randolph, M.F. 2010a. Centrifuge modelling of submarine slides. Ocean Engineering. (submitted in May 2010) Boylan, N., Randolph, M.F., Low, H.E. 2010b. Enhancement of the ball penetrometer test with pore pressure measurements. Proc. 2nd Intern. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia. Boylan, N., Gaudin, C., White, D.J., Randolph, M.F., Schneider, J.A. 2009a. Characterisation of submarine slide runout in geotechnical centrifuge. Proc. of the 28th Intern. Conf. Ocean, Offshore and Artic Eng, Honolulu, Hawaii, USA, OMAE2009-79059. Brandenberg, S.J., Choi, S., Kutter, B.L., Wilson, D.W., Santamarina, J.C. 2006. A bender element system for measuring shear wave velocities in centrifuge models. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 165–171. Bruton, D.A.S., White, D.J., Carr, M.C., Cheuk C.Y. 2008. Pipe-soil interaction during lateral buckling and pipeline walking: the SAFEBUCK JIP. Offshore Technology Conf., OTC19589. Bruton, D.A.S., White, D.J., Cheuk, C.Y., Bolton, M.D., Carr, M.C. 2006. Pipe-soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Offshore Technology Conf., OTC17944. Canepa,Y., Garnier, J., Amar, S. & Corté, J.F., 1988. Comparison between field and centrifugal model tests on shallow foundations, Centrifuge 88, Paris, France, 313–323. Cathie, D.N., Jaeck, C., Ballard, J.-C., Wintgens, J.-F. 2005. Pipeline geotechnics – State-of-the-art. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 95–114. Chen, W. & Randolph, M.F. 2004. Radial stress changes around caissons installed in clay by jacking and by suction. Proc. 14th Int. Offshore and Polar Eng., 2, 493–499. Chen, W. & Randolph, M.F. 2007a. Radial stress changes and axial capacity for suction caissons in soft clay. Géotechnique, 57(6), 499–511. Chen, W. & Randolph, M.F. 2007b. Uplift capacity of suction caissons under sustained and cyclic loading in soft clay. ASCE J. of Geotech. and Geoenv. Eng., 133(11), 1352– 1363. Cheuk, C.Y., White, D.J. Dingle, H.R.C. 2008. Upper bound plasticity analysis of a partially-embedded pipe under combined vertical and horizontal loading. Soils & Foundations. 48(1), 137–148. Cheuk, C.Y., White, D.J. and Bolton, M.D., 2007. Large-scale modelling of soil-pipe interaction during large amplitude cyclic movements of partially embedded pipelines. Canadian Geotechnical Journal, 44(8), 977–996. Chung, S.F., Randolph, M.F., Schneider, J.A. 2006. Effect of penetration rate on penetrometer resistance in clay. ASCE J. of Geo-tech. and Geoen. Eng, 132(9), 1188–1196.

185 © 2011 by Taylor & Francis Group, LLC

Chow, S.H., Nazhat, Y., Airey, D.W. 2010. Applications of high speed photography in dynamic tests. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Clukey, E.C. & Morrison, M. J. 1993.A centrifuge and analytical study to evaluate suction caissons forTLP applications in the Gulf of Mexico. ASCE Proc. Conf. on Foundations, Geotechnical Special Publications, 38, 141–156. Clukey, E.C. & Phillips, R. 2002. Centrifuge model tests to verify suction caisson capacities for taut and semi-taut legged mooring systems. Proc. Deep Offshore Technology Conf., New Orleans. Clukey, E.C., Aubeny, C.P., Murff, J.D. 2003. Comparison of analytical and centrifuge model tests for suction caissons subjected to combined loads. Proc. 22nd Int. Conf. on Offshore Mechanics and Artic Engineering. OMAE-37503. Clukey, E.C., Morrison, M.J., Garnier, J., Corté, J.F. 1995. The response of suction caissons in normally consolidated clays to cyclic TLP loading conditions. Proc. Offshore Technology Conf., OTC 7796. Clukey, E.C., Templeton, J.S., Randolph, M.F., Phillips, R.A. 2004. Suction caisson response under sustained loopcurrent loads. Proc. Offshore Technology Conf. OTC 16843. Colliat, J.L., Dendani, H., Jostad, H.P., Andersen, K.H., Thorel, L., Garnier, J., Rault, G. 2010. Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners end consolidation time, Proc. 2nd Intern. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia. Craig, W.H. 1984. Preface. Proc. Intern. Symp. Application of Centrifuge Modellingto geotechnical Design, Manchester, UK, 1. Craig, W.H., & Al-Sauodi, N.K.S. 1981. The behaviour of some model offshore structures. Proc. 10th Intern. Conf. on Soil Mech. And Found. Eng., Stockholm, Sweden, (2), 541–556. Daiyan, N., Kenny, S., Phillips, R. and Popescu, R. 2010. Investigation of axial/lateral interaction of pipes in dense sand. Proc. 2nd Intern. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia. De Catania, S., Breen, J., Gaudin, C., White D.J. 2010. Development of a multiple axis actuator control system. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Deeks, A.D., White, D.J., Breen, J., Hortin, P., De Catania, S., Brown, T.A. 2010. The use of a wireless data acquisition system for monitoring rotary pile jacking (‘Gyropiling’) in the centrifuge. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Dendani, H., & Jaeck, C. 2007. Pipe-soil interaction in highly plastic clays. Proc. 6th Intern. Conf. Offshore Site Investigation and Geotech. pp. 115–124. Derkx, F., Merliot, E., Garnier, J., Cottineau, L.M. 1998. On-board remote-controlled centrifuge robot. Proc. Centrifuge 98, Tokyo, Japan, 97–104. DeWoolkar, M.M. , Hwang, J. and Ko, H. Y. 2006. Lateral stability of offshore skirted gravity structures. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong Kong, (1), 631–637. DeWoolkar, M.M. , Hwang, J. and Ko, H. Y. 2008. Physical and finite element modelling of lateral stability of offshore skirted gravity structures subjected to iceberg impact load. Int. Ocean Engineering, 35(16), 1615–1626. Dingle, H.R.C, White, D.J., Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout in soft clay. Canadian Geotechnical Journal, 45, 636–652. Dyvik, R., Andersen, K.H., Hansen, S.B., Christophersen, H.P. 1993. Field tests of anchors in clay.

1: Description. ASCE J. Geot. Eng. Div., 119(10), 1515–1531. Finn, Lyle, 2010. Personal communication to Ed Clukey. Gao, F.P. & Randolph, M.F. 2005. Progressive ocean wave modelling in drum centrifuge. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 583–588. Garnier, J. & Cottineau, L.M. 1988. The LCPC centrifuge : means of model preparation and instrumentation, Centrifuge 88, Paris, France, 83–92. Garnier, J., Chambon, P., Ranaivoson, D., Charrier, J. & Mathurin, R. 1991, Computer image processing for displacement measurement, Centrifuge 91, Boulder, USA, 543–550. Garnier, J., König, D. 1998. Scale effects in piles and nails loading tests in sand. Proc. Centrifuge 98, Tokyo, Japan, 205–210. Garnier, J. 2002. Properties of soil samples used in centrifuge models, Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada (1), 5–19. Garnier, J. 2004. Centrifuge modelling in offshore engineering. Proc. 14th Int. Offshore and Polar Eng, Toulon, France. Garnier, J., Gaudin, C., Springman, S.M., Culligan, P.J., Goodings, D., Konig, D., Kutter, B., Phillips, R., Randolph, M.F., Thorel, L. 2007. Catalogue of scaling laws and similitude questions in centrifuge modelling. Intern. J. of Phy. Mod. in Geotech., 7(3), 1–24. Garnier J. & Pecker A. 1999. Use of centrifuge tests for the validation of innovative concepts in foundation engineering. Proc. 2nd Intern. Conf. on Earthquake Geotech. Eng. Lisbonne. Gaudin, C., White, D.J. Bezuijen, A., Schaminée, P.E.L., Garnier J. 2010a. Physical modelling with industry – Overview of practices and benefits. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Gaudin, C., Simkin, M., White, D.J., O’Loughlin, C.D. 2010b. Experimental investigation into the influence of a keying flap on the keying behaviour of plate anchors. Proc. 20th Intern. Conf. Offshore and Polar Eng., Beijing, China. Gaudin, C., Bienen, B., Cassidy, M.J. 2010c. Investigation of the potential of bottom water jetting to ease spudcan extraction in soft clay. Géotechnique. (accepted). Gaudin, C., White, D.J., Boylan, N. Breen, J., Brown, T.A., De Catania, S., Hortin, P. 2009a. A wireless data acquisition system for centrifuge model testing. Measurement Science and Technology, 20, No 095709. Gaudin, C., Landon, M.M., Pederson, V., Kiat, C.H., Lequin, W. 2009b. Design of rock berm cover for offshore pipeline LNG project in Hong Kong. Proc. of the 19th Intern. Conf. Offshore and Polar Eng., Osaka, Japan, 546–553. Gaudin, C. & White, D.J. 2009. New centrifuge modelling techniques for investigating seabed pipeline behaviour. Proc. of 17th Intern. Conf. of Soil Mech. and Geotech. Eng., Alexandria, Egypt, 1, 448–451. Gaudin, C. & Landon, M.M. 2008. CLP Power Ltd Hong Kong LNG project – Rock berm centrifuge testing. The University of Western Australia, GEO 08437, 67 p. Gaudin, C. & Randolph, M.F. 2008. CLP Power Hong Kong Ltd metrological mast. Laboratory centrifuge compression loading tests for Buildings Department in Hong Kong. The University of Western Australia, GEO 08460, 85 p. Gaudin, C. & Cassidy, M.J. 2007. Feasibility and performance of a hybrid foundation system. Report to Keppel Offshore and Marine Technology Centre PTE LTD. The University of Western Australia, GEO 07401, 85 p (confidential). Gaudin, C. & Colwill, R. 2007. Centrifuge modelling of a ship impact against a sloping seawall. Proc. 17th Intern.

186 © 2011 by Taylor & Francis Group, LLC

Conf. Offshore and Polar Eng., Lisbon, Portugal, 4, 3616–3622. Gaudin, C., Randolph, M.F., O’Loughlin, C.D. 2006a. New insights from model tests of foundation and anchoring systems in offshore geomechanics. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 47–62. Gaudin, C., Randolph, M.F., Acosta, H.E., Tran, M. 2006b. Maari field development – Centrifuge modelling of suction foundations. The University of Western Australia, Centre for Offshore Foundation Systems, GEO 06374, 135 p. Gaudin, C., O’Loughlin, C.D., Randolph, M.F., Lowmass, A.C. 2006c. Influence of the installation process on the behaviour of suction embedded plate anchors. Géotechnique, 56(6), 381–391. Hill, A.J. & Jacob, H. 2008. In-situ measurements of pipe-soil interaction in deep water. Offshore Technology Conference, OTC 19528. Hodder, M.S., White, D.J., Cassidy, M.J. 2008. Centrifuge modelling of riser-soil stiffness degradation in the touchdown zone of a steel catenary riser. Proc. Intern. Conf. on Offshore Mechanics and Arctic Engineering, Estoril, Portugal. Hodder, M.S., White, D.J., Cassidy, M.J. 2009. Effect of remolding and reconsolidation on the touchdown stiffness of a steel catenary riser: Observations from centrifuge modeling. Offshore Technology Conference, OTC-19871. Hodder, M.S., White, D.J., Cassidy, M.J. 2010. Analysis of soil strength degradation during episodes of cyclic loading, Illustrated by the T-Bar penetration test. ASCE Inter. J. Geomechanics, (10)3, 117–123. Hossain, M.S., Hu, Y., Randolph, M.F., White, D.J. 2005. Limiting cavity depth for spudcan foundations penetrating clay. Géotechnique, 55(9), 679–690. Iskander, M., Liu, J., Sadek, S. 2001. Transparent amorphous silica to model clay. ASCE J. Geotech. And Geoenv. Eng., 128(3), 262–273. Jacob, H. & Looijen, O. 2008. Development of a deep water tool for in-situ pipe-soil interaction measurements and its benefits in pipeline analysis. Offshore Pipeline Technology Conf., Amsterdam, Netherlands. Jeanjean, P.J., 2009. Development of p-y springs for fatigue analyses. Proc. Offshore Technology Conf., OTC 20158. Jeanjean, P., Znidarcic, D., Phillips, R., Ko, H.Y., Pfister, S., Schroeder, K. 2006. Centrifuge testing on suction anchors: Doublewall, stiff clays, and layered soil profile. Offshore Technology Conf., OTC 18007. Kelleher, P.J. & Randolph, M.F. 2005. Seabed geotechnical characterisation with a ball penetrometer deployed from the Portable Remotely Operated Drill. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 365–371. Lau, M., Phillips, R., McKenna, R.F., Jones, S.J. 2002. Modelling ice-structure interactions in a geotechnical centrifuge. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 325–330. Lee, F.H. 2001. The philosophy of modelling versus testing. Proc. Intern. Symp. Constitutive and Centrifuge Modelling; Two extremes, MonteVerita, Switzerland, 113–131. Lehane, B.M., Gaudin, C., Schneider, J. 2005. Scale effects on tension capacity for rough piles buried in dense sand. Géotechnique. 55(10), 709–719. Lehane, B.M., O’Loughlin, C.D., Gaudin, C., Randolph, M.F. 2009. Rate effects on penetrometer resistance in kaolin. Géotechnique, 59(1), 41–52. Leung, C.F., Xie, Y., Chow, Y.K. 2008. Use of PIV to Investigate Spudcan-Pile Interaction. Proc. 18th Intern.

Conf. Offshore and Polar Eng., Vancouver, Canada, 721–728. Liu, L., Jiang, W., Ma, J. 1988. Automatic data acquisition and processing of centrifugal model tests. Proc. Centrifuge 88, Paris, France, 93–96. Lund, K.M. 2000. Effect of increase in pipeline soil penetration from installation. Proc. Intern. Conf. Ocean Offshore Artic Eng. New Orleans, USA, OMAE2000/PIPE-5047. Madabhushi, S.P.G., Haigh, S.K., Ali, A., Williams, M., Ojaghi, M., Lamata, I., Blakeborough,T.,Taylor, C., Dietz, M. 2010. Distributed testing of soil-structure systems using web-based applications. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Mana, D., Gourvenec, S.M., Hossain, M.S., Randolph, M.F. 2010. Kinematic failure mechanisms of skirted circular foundations under tension and compression. Submitted to journal. Martin, C.M. 2001. Impact of centrifuge modelling on offshore foundation design. Proc. Intern. Symp. Constitutive and Centrifuge Modelling; Two extremes, Monte Verita, Switzerland, 135–154. Morrison, M.J., Clukey, E.C., Garnier, J. 1994. Behaviour of suction caissons under static uplift loading. Proc. Centrifuge94, Singapore, 823–828. Murff, J.D. 1996. The geotechnical centrifuge in offshore engineering. Proc. Offshore Technology Conf., OTC 8265. Newman, J.N. 1977. Marine Hydrodynamics, MIT Press, 354 p. Ng, C.W.W., Zhang, L.M., Wang, Y.H. 2006. Physical Modelling in Geotechnics – 6th ICMPG06, Taylor&Francis. Ng, C.W.W., Van Laak, P.A., Zhang, L.M., Tang, W.H., Zong, G.H., Wang, Z.L., Xu, G.M., Liu, S.H. 2002. Development of a four-axis robotic manipulator for centrifuge modelling at KHUST. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 71–76. O’Loughlin, C.D., Randolph, M.F., Richardson, M.D. 2004. Experimental and theoretical studies of deep penetrating anchors. Offshore Technology Conference, OTC 16841. Oung, O. & Bezuijen, A. 2002. Development of selective pore pressure transducers and their use to check the validity of classical capillary pressure-saturation curves. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 95–100. Ovesen, N.K. 1975. Centrifuge testing applied to bearing capacity problems of footings on sand. Géotechnique, 25(2), 394–401. Palmer, A.C., White, D.J., Baumgard, A.J., Bolton, M.D., Barefoot, A.J., Finch, M., Powell, T.A., Faranski, A.S., Baldry, J.A.S. 2003. Uplift resistance of buried submarine pipelines: comparison between centrifuge modelling and full-scale tests. Géotechnique, 53(10), 877–883. Phillips, R., Barrett, J.A., Al-Showaiter, A.S. 2010. Ice keel-seabed interaction: numerical modelling validation. Offshore Technology Conference, OTC 20696. Phillips, R., Clark, J.I. and Kenny, S. 2005 PRISE studies on gouge forces and subgouge deformations. 18th Int Conf on Port & Ocean Engineering under Arctic Conditions, 1, p75–84. Phillips, R., Clark, J.I., Hanke, R. 2002. Pipeline frost heave modelling. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 313–318. Phillips, R. & Sekiguchi, H. 1992. Generation of water wave trains in drum centrifuge. Proceedings of Techno-Ocean ’92, Yokohama, Japan, volume 1, pp 29–34. Phillips, R., Walter, D.J., Kosar, K.M. 1998. Physical modelling of foundations for Confederation Bridge, Canada. Proc. Centrifuge 98, Tokyo, 447–452. Purwana, O.A., Leung, C.F., Chow, Y.K. 2006. Breakout failure mechanism of jackup spudcan extraction. Proc.

187 © 2011 by Taylor & Francis Group, LLC

6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 667–672. Raines, R.D., Ugaz, O., Garnier, J. 2005. Centrifuge modelling of suction piles in clay. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, 303–308. Rakotonindriana, J. 2009. Comportement des pieux et groupes de pieux sous chargement latéral cyclique, PhD Thesis, Ecole Nationale des Ponts et Chaussées, Paris, 381 p. Randolph, M.F. & White, M.J. 2008a. Pipeline embedment in deep water: processes and quantitative assessment. Offshore Technology Conference, OTC 19128. Randolph, M.F. & White, D.J. 2008b. Upper bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique, 58(4), 297–301. Randolph, M.F., Cassidy, M.J., Gourvenec, S.M., Erbrich, C. 2005. Challenges of offshore geotechnical engineering. Proc. of 16th Intern. Conf. of Soil Mech. and Geotech. Eng., Osaka, Japan, 1, 123–176. Randolph, M.F., O’Neill, M.P., Stewart, D.P. & Erbrich, C. 1998. Performance of suction anchors in fine-grained calcareous soils. Proc. Offshore Technology Conference, OTC 8831. Rezzoug, A., König, D., Triantafyllidis, T.. 2000. Scaling laws in centrifuge modelling for capillary rise in soils, Proc. Int. Symp. On Physical Modelling in Environmental Geotechnics, La Baule, France, 217–224. Richardson, M.D., O’Loughlin, C.D., Randolph, M.F. and Gaudin, C. 2009. Setup following installation of dynamic anchors in normally consolidated clay. ASCE J. of Geotech. and Geoenvironmental Eng., 135(4), 487–496. Richardson, M.D., O’Loughlin, C.D., Randolph, M.F., Cunnigham, T.J. 2006. Drum centrifuge modelling of dynamically penetrating anchor. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, 673–678. Rosquöet, F., Thorel, L., Garnier, J., Canepa, Y. 2007. Lateral cyclic loading of sand-installed piles. Soils and Foundations, (47)5, 821–832. Rowe, P.W. & Craig, W.H. 1981.Applications of models to the prediction of offshore gravity platform foundation performance. Proc. Intern. Conf. on Offshore Site Investigation, London, 269–281. Sasanakul, I., Vanadit-Ellis, W., Sharp, M., Abdoun, T., Ubilla, J., Steedman, S., and Stone, K. 2008. New Orleans levee system-performance during Hurricane Katrina. Proc. 17th Street Canal and Orleans Canal North. ASCE J. Geotech. Eng. Div., 134(5), 657–667. Sassa, S. & Sekiguchi, H. 1999. Wave-induced liquefaction of beds of sand in centrifuge. Géotechnique, 49(5), 621–638. Sarma, D. 2006. Geotechnical modelling by artificial soil. A new approach in model study. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 235–240. SNAME. 1997. Guidelines for site specific assessment of mobile jack-up units. Society of Naval Architects and Marine Engineers, Technical and Research Bulletin 5-5A Rev 1, New Jersey. Song, Z., Hu, Y., O’Loughlin, C.D., Randolph, M.F. 2009. Loss in anchor embedment during plate anchor keying in clay. ASCE J. of Geot. And Geoenv. Eng., 135(10), 1475–1485. Stewart, D.P. & Randolph, M.F. 1994. T-bar penetration testing in soft clay. ASCE J. Geotech. Eng. Div., 120(12), 2230–2235. Take, AW.A. & Bolton, M.D. 2002. A new device for the measurements of negative pore water pressure in centrifuge models. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 89–94. Taylor, R.N. 1995. Geotechnical centrifuge technology. Blackie Academic and Professional.

Teh, K.L., Cassidy, M.J., Leung, C.F., Chow, Y.K., Randolph, M.F. and Quah, C.K. 2008. Revealing the bearing capacity mechanisms of a penetrating spudcan through sand overlying clay. Géotechnique, 58(10), 793–804. Ubilla, J., Abdoun, T., Sasanakul, I., Sharp, M., Steedman, S., Vanadit-Ellis, W., and Zimmie, T. 2008. New Orleans levee system performance during Hurricane Katrina: London Avenue and Or-leans Canal South Canal and Orleans Canal North. ASCE J. Geotech. Eng. Div., 134(5), 668–680. Watson, P.G., Gaudin, C., Senders, M., Randolph, M.F. 2006. Installation of suction caisson in layered soil. Proc. 6th Intern. Conf. Physical Modelling in Geotechnics, Hong-Kong, (1), 685–692. Westgate, Z., Tapper, L., Lehane, B.M., Gaudin, C. 2009. Modelling the installation of stiffened caissons into overconsolidated clay. Proc. 28th Intern. Conf. Ocean, Off. and Artic Engineering, Honolulu, Hawaii, USA. OMAE200979125. White, D.J. & Hodder, M. 2010. A simple model for the effect on soil strength of episodes of remoulding and reconsolidation. Canadian Geotechnical Journal, In Press. White, D.J. & Cathie, D.N. 2010. Geotechnics for subsea pipelines. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth, Australia. White, D.J., Gaudin, C., Boylan, N., Zhou, H. 2010. Interpretation of T-bar penetrometer tests at shallow embedment and in very soft soils. Canadian Geotechnical Journal, 47(2), 218–229. White, D.J. & Cheuk, C.Y. 2008. Modelling the soil resistance on seabed pipelines during large cycles of lateral movement. Marine Structures, 21, 59–79. White, D.J. & Gaudin, C., 2008. Simulation of seabed pipeline behaviour using centrifuge modelling. Proc. Intern. Deep Offshore Tech. Conf., Perth, Australia. White, D.J. & Gaudin, C. 2007. Browse LNG Development Phase 2C: Centrifuge modelling of cone penetration tests. Report to WorleyParsons and Woodside Energy Ltd. The University Of Western Australia, GEO 07422, 21 p. (confidential). White, D.J., Take, W.A., Bolton, M.D. 2003. Soil deformation measurement using particle image velocimetry PIV and photo-grammetry. Géotechnique, 53(7), 619–631. White, D.J, Randolph, M.F., Thompson, B. 2005. An imagebased deformation measurement system for the geotechnical centrifuge. Int. J. of Physical Modelling in Geotechnics, 5(3), 1–12. White, D.J. & Bolton, M.D. 2004. Displacement and strain paths during plane-strain model pile installation in sand. Géotechnique. 54(6), 375–397. White, D.J. & Bolton, M. D. 2001. Observing friction fatigue on a jacked pile. Proc. Intern. Symp. Constitutive and Centrifuge Modelling; Two extremes, Monte Verita, Switzerland, 347–355. Wilson, D.W., Kutter, B.L., Boulanger, R.W. 2010. A summary of the Davis centrifuge facility, recent history and current research, NEES@UC Davis. Proc. 7th Intern. Conf. on Phy. Model. in Geotechnics, Zurich, Switzerland. Yan, Y., White, D.J., Randolph M.F. 2010. Investigations into novel shallow penetrometers for fine-grained soils. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. Perth, Australia. Yang, D. & Goodings, D.J. 1998. Climatic Soil Freezing Modeled in Centrifuge. ASCE J. of Geot. And Geoenv. Eng., 124(12), 1186–1194. Zhou, H. & Randolph, M.F. 2009. Numerical investigations into cycling of full-flow penetrometers in soft clay. Géotechnique, 59(10), 801–912.

188 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Risk and reliability on the frontier of offshore geotechnics R.B. Gilbert The University of Texas at Austin

J.D. Murff Consultant

E.C. Clukey BP America

ABSTRACT: Operating in the frontiers of offshore energy production poses significant risk. The objective of this paper is to present lessons learned in managing risk on the frontier of offshore geotechnics. Case histories are described to underscore the following themes: (1) Achieving an appropriate risk requires balancing risk and conservatism; (2) Managing risk requires understanding the loads on our designs as well as the capacities; (3) Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions; and (4) Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. The paper concludes with guidance to improve the state of practice. Better communication between all parties involved in design, construction and operation and earlier application of risk and reliability principles in the life cycle of a project will enhance the practical value of these principles.

1

INTRODUCTION

Operating in the frontiers of offshore energy production poses significant risk. The consequences of a failure are severe and the costs to mitigate risks are enormous. The offshore industry has been a leader in explicitly considering risk in design and decision making. One of the first reliability-based design guidance documents was developed for offshore facilities. In addition, this industry has been at the forefront in articulating and managing risk levels. Most importantly, due to the extreme conditions under which we operate, we have had the opportunity to learn from experience to innovate and improve practice to better manage risk. The objective of this paper is to present lessons learned in managing risk. Case histories are described to underscore the following themes: 1. Achieving an appropriate risk requires balancing risk and conservatism. 2. Managing risk requires understanding the loads on our designs as well as the capacities. 3. Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions. 4. Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. The case histories presented represent a wide range of facilities, operators and locations. While what

is described is derived from real projects, it only represents a small and geotechnically-biased window into the numerous decisions and factors being considered in making those decisions. The paper concludes with guidance for improving the state of offshore geotechnical practice to better manage risk in the future. 2 ACHIEVING APPROPRIATE RISK Risk is the possibility of a loss. An appropriate risk is one that balances the cost of reducing the risk against that of accepting the risk.Appropriate risk is never zero risk, and excessive conservatism can be as troublesome as excessive risk. Risk is quantified as the expected consequence of loss. In the simplest case, if the possibility of loss is a binary event that either does or does not occur, then the risk is equal to the probability of the loss multiplied by its consequence. Therefore, risk is commonly represented by the probability and consequence of a loss. Consequences can include human safety, environmental and economic losses. General guidance for striking the balance between reducing versus accepting risk has been developed by different governments and industries. Examples include USNRC (1975) for nuclear power plants, AIChE (1989) for chemical process facilities, ANCOLD (1998) and USBR (2003) for dams, and Bea (1991), Stahl et al. (1998) Goodwin et al. (2002) and Gilbert et al. (2001) for offshore production facilities.

189 © 2011 by Taylor & Francis Group, LLC

ballast would provide an acceptable reliability, even if the riser was installed immediately after pile driving. 2.2

Figure 1. Required ballast versus set-up time to achieve target probabilities of overload during installation for a riser tower foundation.

or on the capacity side of the design check (increasing the capacity),

The following case histories illustrate how the cost of reducing risk against that of accepting risk can be balanced in practice. 2.1

where rn is the nominal design capacity of the suction caisson in uplift due to side shear and reverse end bearing, sn is the nominal design load in uplift, and ϕR and γS are resistance and load factors, respectively. The ratio γS /ϕR is greater than 1.0, meaning that a larger nominal design capacity would be required if the weight of the caisson were considered to be a component of the capacity (Equation 1b). This case underscores the arbitrariness implied typically in different design formulations. The design-build contractor proposed using Equation (1a) versus Equation (1b). The basis for their proposal was that there was relatively less uncertainty in the weight of the caisson than the components of side shear and reverse end bearing. In addition, this proposal provided for a less expensive foundation. In order to evaluate the appropriateness of this proposal, a reliability analysis was performed. This analysis has the advantage that the results do not depend on a particular formulation for the design check. The event of interest was failure of the suction caisson in uplift under the long-term sustained load from the riser tower (short-term dynamic loads were very small in comparison to the total uplift load). The calculated probability of uplift failure in the 20-year design life was between 0.01 and 0.1. This probability of failure was considered to be intolerably high, and the design standard was subsequently clarified and revised in order to provide for a higher level of reliability.

Driven pile foundations for riser towers

The riser towers for an oil production system in deep water were to be held in place with driven pile foundations. Since the preliminary design was based on suction caissons, the soil borings were relatively shallow. The driven pile lengths were subsequently constrained by the depth of the deepest soil boring due to concerns about driveability. In addition, the largest load would be applied during riser installation, which potentially could occur before significant set-up after pile driving. Based on a design check using nominal loads and capacities, the axial uplift capacity of the piles alone was not sufficient. In order to increase the capacity, a ballasting system was developed, with boxes at the head of the piles that could be filled with steel ballast to increase the uplift capacity. A reliability analysis was performed to assess the need for the ballast boxes and the optimal amount of ballast (Fig. 1). The event of concern was overload during riser installation when the uplift force would be greatest because the riser is empty and the pile capacity would not necessarily have reached its maximum value due to full set-up. Based on a consideration of the consequences of a failure (primarily economic), the target threshold for the probability of overload was set between 0.001 and 0.0001. The results in Figure 1 provide a range of possible means to achieve an appropriate level of risk, including using larger ballast with a shorter set-up time or smaller ballast with a longer set-up time. Based on the proposed check, in which nominal or conservative values were used both for the uplift load and pile capacity, a ballast weight that was nearly 70 percent of the design load was selected (Fig. 1). The reliability analysis indicated that less than half that amount of

3

UNDERSTANDING LOADS

The focus in offshore geotechnics is generally on the capacity of geotechnical systems. However, the reliability of these systems can be influenced as much by the loads they experience as by their capacity. The

190 © 2011 by Taylor & Francis Group, LLC

Suction caisson foundations for riser towers

A suction caisson was designed to anchor a bundle of risers for an oil production system in deep water. This design was unconventional in that the maximum load was a sustained tension load and not a storm load and the capacity was primarily from the weight of the foundation and not the shear strength of the soil. This particular situation meant that conventional factors of safety in the design standard did not necessarily apply. Specifically, the question was whether to include the weight of the foundation on the load side of the design check (reducing the total uplift load),

Figure 2. Schematic of wave-induced mudslides.

loading

mechanism

for

following case histories highlight the importance in understanding loads.

Figure 3. Results of slope stability analyses for offshore site in 130 m of water.

3.1 Wave-induced mudslides Wave-induced mudslides have caused considerable damage to offshore facilities in the Gulf of Mexico, particularly to pipelines that are an essential link in the supply of oil and gas. In Hurricanes Ivan (2004) and Katrina (2005), more than 50 pipelines and one platform were damaged or destroyed by mudslides in the Mississippi Delta (OTRC 2008). A schematic of the loading mechanism for a wave-induced mudslide is shown in Figure 2. The differential pressure on the sea floor underneath the waves (Fig. 3) imparts shear stresses in the soil. If these shear stresses exceed the shear strength of the soil, then a slope failure will occur. Results of stability analyses for a site subjected to large waves in Hurricanes Ivan and Katrina are shown in Figure 3 (OTRC 2008). The analysis assumes that the soft clays are sheared under undrained conditions by the waves and uses a profile of undrained shear strength versus depth from a nearby soil boring. Mudslides occurred here in both hurricanes, which is consistent with the stability analyses (Fig. 3). While the maximum wave heights in Hurricane Ivan at this location were nearly 20 percent smaller than those in Hurricane Katrina, the bottom pressure loading was actually higher in Hurricane Ivan and the factor of safety was smaller due to the longer wave period (Fig. 3). This result is significant because the eye for Hurricane Ivan was nearly 150 km to the east of the Delta (in contrast to Katrina’s eye, which passed over the Delta); the relatively small wave heights for Ivan in the Delta would not have been expected to induce mudslides, particularly in water depths of 130 m. However, the magnitude of bottom pressures, and therefore the risk of mudslides, depends both on the wave height and period. A comparison of wave heights and periods in the Delta for these two storms is shown in Figure 4. The conventional approach to characterize sea states in a hurricane is to relate wave period to wave height

Figure 4. Comparisons of wave periods and wave heights for waves in Mississippi Delta during recent hurricanes (from OTRC 2008).

based on empirical data collected from the largest waves in a storm (i.e., those waves near the eye). The curve labeled “Average” in Figure 4 is typically used to associate wave periods with wave heights for design purposes, such as in designing a production platform. However, the average curve under-represents the potential for long period waves away from the eye; the periods for the waves in the Delta during Ivan were similar to those near the eye, 150 km to the east, even though the wave heights were smaller (Fig. 4). The risk for mudslides in the Delta would be underestimated if this potential for longer period waves were not considered. Since platform loads are not very sensitive to the wave period, this issue was not recognized by and is not captured by the standard design guidance documents developed for platforms. However, it is an important consideration for mudslides. This case history underscores the significance of having a careful understanding of loads and not using information from design guidance documents without knowing its source and limitations.

191 © 2011 by Taylor & Francis Group, LLC

Figure 6. Conceptual relationship between design conservatism, information and acceptable reliability.

of designers to apply conservatism at every decision point, it can clearly result in costly designs.

Figure 5. Annual probability that the actual load on the foundation will exceed the design load.

4

3.2 Tendon loads There can be significant conservatism embedded in the design loads that geotechnical engineers use to size foundations. To illustrate this point, the following conservative assumptions were made in establishing the design tendon load for a driven pile foundation of a Tension Leg Platform. First, an unlikely combination of wave height and period was assumed to establish the maximum tendon load. The 100-year design load corresponded to a significant wave height having a 0.01 annual probability of exceedance (the “100-year” wave height) combined with an associated peak spectral period with only a 5-percent probability of exceedance. Since the maximum tendon load increases with decreasing wave period due to the natural period of the structure, the resulting design load can be significantly greater than the load that will be exceeded with an annual probability of 0.01 (the “100-year load”). In addition, a conservative estimate of the tide and the worst possible wave loading direction were used for design, giving the maximum possible tendon load for a sea state. However, this combination of tide and wave loading direction was relatively unlikely in the conditions expected to produce the largest waves. Second, conservative assumptions were made concerning pre-tension measurement error in the tendon, mis-positioning of the pile head, and subsidence of the sea floor. The result of these compounded conservatisms was a design load with an extremely small probability of being exceeded. Figure 5 shows a probability distribution for the maximum annual tendon load on the pile. The annual probability that it will be exceeded is less than 1 in 100,000 (Fig. 5). Therefore, the “100-year” design load was really a “100,000-year” design load, which was subsequently factored up to establish the required design capacity. Therefore, the probability of failure for this design was many orders of magnitude smaller than what would generally be considered tolerable. This approach, selecting a series of conservative parameters in a design, is often referred as “double dipping.” Although it may be the natural tendency

A common decision point in offshore design is whether or not additional data are needed or would be beneficial enough to justify the cost of acquiring the data. A conceptual schematic illustrating how reliabilitybased design can be used to guide this decision is shown in Figure 6. This figure shows the cost of a foundation versus the cost of the information that is used to design the foundation. An acceptable level of reliability can be achieved by a variety of combinations of design information to reduce uncertainty in the foundation performance and design conservatism to limit the effect of uncertainty on the performance of the design. Design conservatism is reflected in the load and resistance factors or factors of safety as well as the nominal values used for loads and capacities. Increasing the cost of design entails increasing the difference between the expected capacity and the expected load. The level of information used in developing a design is rarely expressed explicitly in a code; however, it is typically implied through conventional practice. For example, design codes for foundations typically assume that a site-specific geotechnical investigation has been conducted to develop the design. Once a reliability-based design code is calibrated, the combination of the design information and the design conservatism presumably gives foundations with an acceptable level of reliability. The value of additional information depends on how much that information is expected to reduce design conservatism. This potential cost savings can then be compared against the cost of obtaining additional information, which is often governed by the time it will take to acquire it and the subsequent delay in development. Maximizing the value of information entails finding the optimal combination of design information and conservatism, such as the combination that will minimize the total expected cost. The following case histories illustrate projects where risk and reliability principles were used to improve the value of information.

192 © 2011 by Taylor & Francis Group, LLC

MAXIMIZING THE VALUE OF INFORMATION

4.1 Design without site-specific soil boring A mature field had a rather substantial amount of geotechnical data, with more than 50 soil borings and geophysical survey lines. In addition, the cost and time required to obtain site-specific borings had been steadily increasing. Therefore, the question was whether or not a site-specific soil boring was worth its cost for every new facility in this field. The conceptual approach shown in Figure 6 can be captured in the design checking equation for a reliability-based code as follows:

where rn is the nominal design capacity and sn is the nominal design load; ϕR and γS are the conventional load and resistance factors, respectively, to account for the uncertainty in these quantities even if a new soil boring were available at the site location; and ϕspatial is a partial resistance factor to account for the added uncertainty if the amount of design information is less than the convention. As the magnitude of added uncertainty increases due to having less design data, the quantity (1/ϕspatial ) increases, meaning that a more conservative design is required to achieve the conventional reliability. In order to relate the value of (1/ϕspatial ) to the amount of design information, a geostatistical model was developed for this offshore field and calibrated with the available geotechnical data. The details of this model are described elsewhere (Gambino and Gilbert 1999 and Gilbert et al. 2008), and its important features are summarized here: • The model accounts for horizontal and vertical

correlation in the pile capacity. • The model incorporates data from modern borings (pushed sampling methods) as well as data from older borings where samples were obtained using wire-line percussion methods and were more disturbed. The effect of the method of sampling was included both in the expected pile capacity as well as the standard deviation in the capacity. • The model predictions reflect uncertainty both due to spatial variations as well as systematic uncertainty in the model due to the limited amount of data available to calibrate it. The output from this model is shown in Figure 7 for a small section in this field. The mean capacity at the location of an available, modern boring is equal to the design capacity obtained directly from that boring (Fig. 7a). The mean capacity at the location of an available, older boring is adjusted from the boring data to reflect what would be expected if a modern boring were drilled at that location. The unconditional mean or average for the field is 30 MN (Fig. 7a); the data at both the modern and the older boring suggest that the pile capacity within this section is higher than that for the field on average. As the location of the structure moves away from the location of the borings, the

Figure 7. (a) Expected value (b) and coefficient of variation (c.o.v.) for estimated pile capacity versus location for a 100-m long, 1-m diameter steel pipe pile (from Gilbert et al. 2008).

expected value for the design capacity approaches the unconditional mean (Fig. 7a). The added uncertainty in the capacity, expressed as the coefficient of variation or standard deviation divided by the mean, is zero at the location of the modern boring because this data point represents the conventional practice (Fig. 7b). The added uncertainty in the capacity is greater at the location of the older boring, reflecting the greater uncertainty associated with these data compared to modern practice. The ceiling of Figure 7b corresponds to the unconditional variance for the field. As the location of the structure moves away from the boring locations, the added variance approaches the unconditional variance (Fig. 7b). The required value for (1/ϕspatial ) is related to the added uncertainty due to not having a modern boring at the site location. Figure 8 shows this relationship for this field, where the added uncertainty is expressed as a coefficient of variation. In this application, the coefficient of variation for the added uncertainty was typically between 0.05 and 0.08 (Figure 7b); therefore, the required safety margin is about 10 percent greater than for the conventional case (i.e., 1/ϕspatial is about 1.1 in Fig. 8). These results provided the owner of the structure with guidance in making a decision between drilling a

193 © 2011 by Taylor & Francis Group, LLC

Figure 8. Partial resistance factor versus coefficient of variation for spatial variability.

Figure 9. Probability distributions for tendon load and tensile pile capacity.

new boring or relying on the existing data and increasing the safety margin by about 10 percent. If the cost of increasing the pile length was less than the cost of a site-specific boring, then the alternative of moving ahead without another soil boring (i.e., “Less Data” on Fig. 6) would be preferred. The value of information from a site-specific boring depends on the geology; the greater the spatial variability, the greater the penalty to be paid by not having a site-specific boring (Fig. 8). The value of information also depends on the use of the information. For example, decisions about pile driveability may be more or less sensitive than those about pile capacity to the information from a site-specific boring. 4.2

Updated design using pile driving data

The pile foundation for a tension leg platform in a frontier area was designed based on a preliminary analysis of the site investigation data. The geotechnical properties of the site were treated in design as if the soil conditions were similar to other offshore areas where the experience base was large. The steel was then ordered. Subsequently, a more detailed analysis of the geotechnical properties showed that the soil conditions were rather unusual, calling into question how the properties should be used in design and leading to relatively large uncertainty in the estimated pile capacity. With three different commonly-used methods for estimating the tensile capacity of the piles, the factor of safety ranged from less than 2.5 to more than 4 (Fig. 9). For reference, the target factor of safety was 3. The owner was faced with a series of decisions. First, should they stay with the original pile design or change it given the uncertainty in the pile capacity, considering that changing the pile design after the steel had been ordered would substantially impact the cost and schedule of the project? Second, if they decided to stay with the original pile design, should they monitor the installation to confirm the capacity was acceptable, considering that this approach required a flexible contract where the pile design may need to be updated after the pile is installed?

194 © 2011 by Taylor & Francis Group, LLC

Figure 10. Expanded view of probability distributions for tendon load and tensile pile capacity.

A reliability analysis was performed to provide guidance in answering these questions. One of the first findings from this analysis was that the uncertainty in the maximum tendon load was small compared to that in the pile capacity (Fig. 9). The second finding from this analysis was that the reliability of the pile foundation was governed by the possibility of the capacity being far below what was predicted by any of the design methods being considered. Figure 10 shows how the most probable combination of load and capacity leading to failure (the “Most Probable Failure Point”) is well below the estimated axial capacity assuming that the side shear is equal to the remolded undrained shear strength, which is a reasonable lower-bound on the available capacity. To support this concept of a lower-bound capacity, the pile load database that was used to develop and calibrate the API Design Method is shown in Figure 11 together with the calculated lower-bound for each data point. In all cases, the measured capacity exceeds the calculated lower-bound.

Figure 12. Probability of foundation failure in design life versus a lower-bound estimate of the pile capacity.

Figure 11. Comparison of measured capacity for driven piles in clay soils with estimated lower-bound capacity calculated assuming the side shear is equal to the remolded undrained shear strength of the clay; measurements are shown as a range to account for uncertainty in interpreting the load tests results (adapted from Najjar 2005).

A reliability analysis that explicitly accounted for the existence of a lower-bound or minimum value of the pile capacity was then performed. The lower bound was modeled as uncertain, with a normal distribution having a coefficient of variation equal to 0.2. The results of this analysis are shown in Figure 12, where the estimated value for the lower bound represents the most likely lower-bound value. If the estimated lower bound was greater than 0.25 times the estimated (or median) pile capacity after set-up, then the probability of foundation failure in the design life was smaller than the target of 0.001 (Fig. 12). The estimated value for the lower bound, obtained using the side shear equal to the remolded undrained shear strength, was about 0.4 times the estimated capacity after set-up, meaning that the existing design would have an acceptable reliability. Pile driving data during and after driving were used to update the reliability of this foundation. In general, pile monitoring information for piles driven into normally to slightly overconsolidated marine clays cannot easily be related to the ultimate pile capacity due to the effects of set-up following installation and uncertainties in the pile driving model. The capacity measured at the time of installation may only be 20 to 30 percent of the capacity after set-up. However, the pile capacity during installation does arguably provide a lower bound on the ultimate pile capacity, which governed the reliability of the foundation in this case. The estimated capacity based on the soil resistance to driving indicated that the reliability of the foundation was acceptable (Fig. 12). A re-tap after several days of set-up further supported this conclusion (Fig. 12). 4.3 Communication of uncertainty due to lack of information An important aspect in evaluating information is to clearly communicate uncertainty. As geotechnical

Figure 13. Stratigraphic cross-section with locations of soil borings and proposed platform (adapted from Gilbert et al. 2006).

engineers, we are rarely in the position of making decisions directly; rather we are conveying what we do and do not know as guidance to the decision makers. If a decision maker does not understand the magnitude of uncertainty, then they may not appreciate the potential value of additional information. A common offshore example is shown in Figure 13 where the final location of the platform does not correspond to the locations of the available soil borings. The geologic setting is a normally consolidated marine clay that is interbedded about every 30 m with overconsolidated clay crusts. In addition, there are buried alluvial channels that are filled with a variety of clay, silt and/or sand. A channel filled with dense sand was encountered at the location of Boring B. The risk is whether the driven piles at the platform location will encounter a thick enough layer of dense sand to cause driving problems. The conventional means to convey the available information is shown in Figure 13. A cross-section is drawn between the borings that includes geophysical data together with the data from the borings. The difficulty is that this cross-section conveys what is

195 © 2011 by Taylor & Francis Group, LLC

Figure 14. Uncertainty multiples showing uncertainty in straigraphy at platform location (adapted from Gilbert et al. 2006).

expected, not what might occur. Even though there are question marks on the channel location away from the boring, a decision maker will tend to assume that there is no possibility of sand at the platform location because it is not shown below the platform location in the cross-section (Fig. 13). However, there is considerable uncertainty in whether or not and how much dense sand lies below the platform location. The available geophysical survey lines do not match up directly with the location of the platform. In addition, there is considerable uncertainty in the interpretation of the geophysical data, time-slice (seiscrop) plots from high resolution, surface-towed boomers. Comparisons of geophysical interpretations with subsequent boring data in this field indicated that the geophysical interpretations were not very accurate at identifying and locating channels buried with sand. An alternative depiction of the uncertainty in this information is shown in Figure 14. This depiction makes use of multiples, which are small-scale images positioned within the eye span on a single page or screen (Tufte 1990). In this case, “uncertainty multiples” (Gilbert et al. 2006) are used to convey the range of possible interpretations of the information based on the collective judgment of the geotechnical engineers and geologists, with each image representing an equally probable possibility. Figure 14 shows that there is a reasonable possibility (4/9 probability) of encountering sand at the platform location. It shows that if sand is encountered, its thickness could range from several meters to tens of meters. It shows that if a channel does lie below the platform location, it may or may not be the same channel or filled with the same material as the channel encountered at Boring B. It even shows that the thickness of the sand layer at the location of Boring B is uncertain; there is a 1/9 probability that it is thinner due to sand running into the borehole when the layer was encountered. The intent of Figure 14 is that a decision maker, who is most likely not a geotechnical engineer or a geologist, will be readily able to understand the uncertainty in the subsurface conditions. Whether or not a new site-specific soil boring is warranted will depend

Figure 15. Schematic plan view of mooring system for study spar.

on the cost of the boring, the cost of being prepared to drive through a dense sand layer if it is encountered, and the expected cost (or risk) of pile driving difficulty. 5

Most design considerations for foundations are based on individual components. However, these components generally perform within a system of foundation and structural elements. Ultimately, it is the performance of the system that is of most interest and concern. The following case histories underscore the importance of considering the system as well as the components. 5.1

Mooring system

The mooring system for a spar is shown in Figure 15. This spar design was developed by industry to be representative of typical practice for the purposes of studying and assessing that practice (OTRC 2006). There were fourteen lines, each composed of a section of chain connecting the line to the hull, a section of rope from the upper chain to just above the mudline, a section of lower chain connecting the rope to the suction caisson anchor at a padeye below the mudline. Three different water depths were considered for this system: 1,000 m with a semi-taut mooring spread and wire rope lines; 2,000 m with a taut mooring spread and polyester lines; and 3,000 m with a taut mooring spread and polyester lines. The probabilities of failure for individual components in the most-heavily-loaded line are shown in Figure 16. For each design (water depth), the probability of failure for the anchor is more than two orders of magnitude smaller than that for the rope and chain segments of the line; hence, failure of a mooring line during a storm is expected to be a break in the line itself versus a pull-out of the anchor. There is a potential to make the anchor designs more efficient (e.g., using a lower factor of safety) without jeopardizing the performance of the mooring system. Second, the consequence of a failure may depend on how the

196 © 2011 by Taylor & Francis Group, LLC

CONSIDERING SYSTEMS AS WELL AS COMPONENTS

Figure 16. Probability of failure for components in most-heavily-loaded line.

failure occurs. Failure in the lines means that the hull could move off station by hundreds of kilometers during a storm and collide with other offshore facilities or coastal facilities. Failure by pull-out of the anchors means that the hull may not move off station as far due to the restoring force provided by the weight and dragging resistance of the anchors. However, the dragging anchors could damage seafloor facilities such as well heads and pipelines. The results on Figure 16 also show that the reliabilities of the components (chains, rope and anchor) and of the system of components that make up a single line (total) depend on the type of mooring system. The probabilities of failure are notably larger for the semi-taut system in 1,000 m of water compared to the taut systems in 2,000 and 3,000 m of water. These differences in reliability are due to using the same design recipe for all mooring systems, which is the standard of practice. Uncertainty in the maximum line load for a taut system is smaller than that for a semi-taut system because a greater proportion of the total load is pre-tension and controlled. Hence, the probability of failure for a taut system is smaller since it is less likely that the actual load will exceed the design capacity when the same factors of safety or load and resistance factors are used. Redundancy in the mooring system is shown in Figure 17 as the probability that the mooring system will fail (i.e., loss of station keeping for the spar) in the event that a single line has failed during a hurricane. A measure of redundancy is the inverse of the probability of system failure given component failure; for example, this redundancy factor for the semi-taut system is about 1.6 (or 1/0.6). Again, there is a significant difference between the different designs (water depths). For the semi-taut system, there is relatively little redundancy between failure of a single line and failure of the system. Conversely, the taut system distributes the environmental load more evenly between lines, and the corresponding redundancy factor is between 15 and 20. For comparison, the redundancy for an eight-leg fixed jacket in shallow water is between 30 and 125, depending on the loading direction (Tang and Gilbert 1993).

Figure 17. Probability of failure of the mooring system in the event that one line fails during a hurricane.

5.2

197 © 2011 by Taylor & Francis Group, LLC

Jacket pile system

System effects can be even more pronounced for fixed jackets compared to floating production systems. To illustrate interactions between the jacket and the foundation system, the capacity of the foundation system for an 8-pile jacket that experienced large loads in Hurricane Ike is shown in Figure 18. The curves in Figure 18 represent capacity envelopes for various combinations of base shear and overturning moment applied to the foundation system; a combination of shear and moment that is outside of the envelope is expected to cause system failure. These system capacity curves were obtained using an upperbound, plasticity model of the pile system (Murff and Wesselink 1986 and Gilbert et al. 2010). The case study platform in Figure 18 is of particular interest because while the maximum load in Hurricane Ike was greater than the estimated capacity for the foundation system, the foundation survived the hurricane. The structural engineers assessing this structure after the hurricane increased the undrained shear strength of clay layers along the pile sides by a factor of three times in order to “explain” its survival. This practice is common place, and has led to a perception that foundation designs are excessively conservative (e.g., Gilbert et al. 2010). This case also points out the fallacy of trying to adjust an overall result by dramatically varying a parameter that has a relatively weak effect. However, the survival of this foundation system can more plausibly be explained by considering the entire structural system. The ultimate capacity of the system will be reached at large lateral deformations of the individual piles; therefore, any degradation of lateral soil resistance due to cyclic loading at lower load levels are not expected to affect the ultimate lateral capacity. In addition, the steel members in the structure, including the piles and jacket legs, are expected to have a higher yield strength than the nominal value used in design. Using expected lateral soil resistance and steel yield strength alone are enough to explain the

Figure 18. Increase in foundation system capacity of jacket platform in the hurricane loading direction due to higher lateral soil resistance (static versus cyclic p-y curves), increased steel yield strength (fY ) and modeling jacket leg stubs (adapted from Gilbert et al. 2010).

survival of this foundation in Hurricane Ike (Fig. 18). In addition, the jacket legs actually extend 3 to 4 m below the mudline, which increases the shear capacity of the system by forcing the plastic hinges in the piles deeper below the mudline. Accounting for this system effect further increases the capacity of the foundation system beyond the loading experienced in Hurricane Ike. Therefore, the performance this platform does not necessarily suggest that foundation designs are excessively conservative when the entire structural system is considered. A comparison of system effects for two jacket pile systems is shown in Figure 19. These figures were developed by assessing the sensitivity of system capacity to variations in the capacity of individual piles. The axial and lateral capacities of each pile were varied by ±30 percent from the design capacities to reflect possible variations in soil properties, installation conditions and structural capacities between piles. The maximum and minimum capacities obtained from this variation in any single pile are shown in Figure 19 to assess robustness of the pile system (i.e. the ability of the system to accommodate variations in component strengths). Both jacket pile systems are robust to variations in individual pile capacities for a shear-dominated failure (i.e., smaller overturning moments). These results complement those on Figure 18, where the pile system is more sensitive to structural factors than geotechnical factors for a shear-dominated failure. Conversely, both jackets are less robust to variations in individual pile capacities for an overturning-dominated failure (Fig. 19).Therefore, these jacket pile systems are much more sensitive to uncertainties in axial pile capacity than to uncertainties in lateral pile capacity. The three-pile system (Fig. 19a) is less robust in overturning than the six-pile system (Fig. 19b). For the same ±30-percent variation in axial capacity for individual piles, the three-pile system exhibits

© 2011 by Taylor & Francis Group, LLC

Figure 19. Robustness in pile system capacity for (a) a three-pile jacket and (b) a 6-pile jacket (b) (adapted from Chen et al. 2010).

about a ±25-percent variation in system overturning capacity while the six-pile system exhibits less than ±10-percent variation in system overturning capacity. This difference in system behavior is real. Both pile systems experienced hurricane loads near their system capacities. The three-pile system failed in overturning while the six-pile system survived in shear without detectable damage. This difference in system behavior is not explicitly reflected in design practice. The same design recipe is used to calculate the loads and required design capacities for individual piles, regardless of whether the piles are in a single-pile system (like a caisson or a riser tower), a three-pile system, or a six, eight or more pile system. Therefore, the resulting reliability for these various systems will not necessarily be consistent. 6 THE FUTURE The keys to better managing risk in the future on the frontier of offshore geotechnics are embodied in the preceding lessons and case histories. Achieving an appropriate level of risk requires that we pay as much attention to the consequences of failures as to the probabilities of failures and costs associated with reducing the probabilities of failure.

198

A thoughtful examination of the consequences of failures can also highlight useful means to reduce risks. It may be more cost-effective to reduce risk by mitigating consequences compared to increasing reliability. A fundamental understanding of what affects the performance of a geotechnical system, including both the loads and the capacities, is far more effective in engineering appropriate solutions than simply being conservative to manage risk. Realistic estimates for loads and capacities are needed as well as the “nominal” values used in design. Realistic estimates include what is most likely, how much uncertainty or variation there is, what the sources of uncertainty are (e.g., temporal variations, spatial variations, inference errors, model uncertainties, installation tolerances, human errors, etc.), and any lower or upper bounds that physically limit what is possible. Furthermore, being “conservative” can have unintended consequences that result in taking on greater risk. Difficulty installing a “conservatively” designed foundation that is larger or longer than usual may leave a facility with less capacity in the foundation than if a “less conservative” design had been used. There is ample potential to improve the benefitto-cost ratio for the information used in offshore geotechnics. Realizing this potential will mean focusing on acquiring data that will be most informative or have the greatest impact on project-specific decisions. This approach is in contrast to always getting similar information for each project. It requires a frontend investment to develop preliminary decisions and designs based on available information so that the value of additional information can be assessed. It requires developing decisions and designs that are adaptable so that they can be updated throughout their lifetime based on new information. It requires considering how information from an existing project may provide value in future projects. Adopting a system-wide perspective can pay large dividends in better managing risks. Designs need to implicitly consider how the performance of geotechnical components will affect the overall performance of the system. These system considerations are not necessarily captured in design codes and guidelines, and they are project specific. These system considerations extend beyond the physical facilities to the factors driving decisions on the project. An elegant and economical foundation design using new technology may not be useful if the contractor selected to install it is unwilling to use different technology. These system considerations require that geotechnical engineers interact closely and communicate clearly with other disciplines, from conceptual design through construction and operation. Finally, humans are an integral part of managing risk. Humans collect and interpret the available information, make decisions and develop designs, construct and operate facilities, and are impacted directly and indirectly when failures occur. Since geotechnical engineers are not formally trained in psychology, sociology or communication, we need to reach out to

professionals in these non-technical areas in order to more effectively manage risk. 7

The intent of this paper was to provide case histories to guide practitioners in managing risk on the frontier of offshore geotechnics. The major themes are: 1. Achieving an appropriate risk requires balancing risk and conservatism. We need to become as involved in assessing and mitigating the consequences as we are assessing and minimizing the probabilities of failures, particularly in the context of the larger systems. 2. Managing risk requires understanding the loads on our designs as well as the capacities. We need to be careful to convey realistic ranges of possibilities and to avoid inadvertently compounding conservatisms. 3. Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions. We need to identify where information will be and will not be most valuable on a project-specific basis, develop designs and decisions that can readily be adapted based on new information, and convey uncertainty as clearly as possible to the decision makers. 4. Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. We need to interact closely with other disciplines, from conceptual design through construction and operation. The use of risk and reliability principles has become an important part of offshore geotechnics, especially on the frontier. Better communication between all parties involved in design, construction and operation and earlier application of these principles in the life cycle of a project will enhance the practical value of the principles. ACKNOWLEDGMENTS We would like to acknowledge the following organizations that have directly or indirectly supported the material in this paper: US Minerals Management Service, American Petroleum Institute, National Science Foundation, Offshore Technology Research Center, BP, ExxonMobil, Sonaghal, Chevron, Shell, and numerous other operators and contractors that constitute the offshore oil industry.The views and opinions presented herein are ours alone and do not necessarily reflect those of any sponsor or employer. REFERENCES AIChE. 1989. Guidelines for Chemical Process Quantitative Risk Analysis. Center for Chemical Process Safety of the American Institute of Chemical Engineers. New York, New York.

199 © 2011 by Taylor & Francis Group, LLC

CONCLUSIONS

ANCOLD. 1998. Guidelines on Risk Assessment, Working Group on Risk Assessment. Australian National Committee on Large Dams. Sydney, New South Wales, Australia. Bea, R.G. 1991. Offshore platform reliability acceptance criteria. Drilling Engineering, Society of Petroleum Engineers. June Issue, 131–136. Chen, J-Y., Gilbert, R.B., Murff, J.D.,Young, A. and Puskar, F. 2010. Structural factors affecting the system capacity of jacket pile foundations. Proceedings of International Symposium on Frontiers in Offshore Geotechnics. in press. Gambino, S.J. and Gilbert, R.B. 1999. Modeling spatial variability in pile capacity for reliability-based design. Analysis, Design, Construction and Testing of Deep Foundations, Roesset Ed., ASCE Geotechnical Special Publication No. 88, 135–149. Gilbert, R.B., Najjar, S.S., Choi, Y.J. and Gambino, S.J. 2008. Practical application of reliability-based design in decision making, Book chapter in Reliability-Based Design in Geotechnical Engineering: Computations and Applications, Phoon Ed., Taylor & Francis Books Ltd., London. Gilbert, R.B., Chen, J.Y., Materek, B., Puskar, F., Verret, S., Carpenter, J., Young, A. and Murff, J.D. 2010. Comparison of observed and predicted performance for jacket pile foundations in hurricanes. Proceedings of Offshore Technology Conference. OTC 20861. Gilbert, R.B., Tonon, F., Freire, J., Silva, C.T. and Maidment, D.R., 2006. Visualizing uncertainty with uncertainty multiples. Proceedings, GeoCongress 2006, ASCE, Reston, Virginia. Gilbert, R.B., Ward, E.G. and Wolford, A.J. 2001. A comparative risk analysis of FPSO’s with other deepwater production systems in the Gulf of Mexico. Proceedings of Offshore Technology Conference. OTC 13173.

Goodwin, P., Ahilan, R.V., Kavanagh, K. and Connaire, A. 2000. Integrated mooring and riser design: target reliabilities and safety factors. Proceedings, Conference on Offshore Mechanics and Arctic Engineering. 185–792. Murff, J.D. & Wesselink, B.D. 1986. Collapse analysis of pile foundations. 3rd International Conference on Numerical Methods in Offshore Piling, Nantes, France: 445–459. Najjar, S.S. 2005. The Importance of Lower-Bound Capacities in Geotechnical Reliability Assessments. Ph.D. Dissertation, The University of Texas at Austin, 347 pp. OTRC. 2006. Reliability of Mooring Systems for Floating Production Systems. Final Report for Minerals Management Service, Offshore Technology Research Center, College Station, Texas, 90 pp. OTRC. 2008. Mudslides During Hurricane Ivan and an Assessment of the Potential for Future Mudslides in the Gulf of Mexico. Final Project Report, Offshore Technology Research Center. Prepared for US Minerals Management Service. 190 pp. Stahl, B., Aune, S., Gebara, J.M. and Cornell, C.A. 1998. Acceptance criteria for offshore platforms. Proceedings of Conference on Offshore Mechanics and Arctic Engineering. OMAE98-1463. Tang, W.H. and Gilbert, R.B. 1993. Case study of offshore pile system reliability. Proceedings of Offshore Technology Conference. 677-686. USBR. 2003. Guidelines for Achieving Public Protection in Dam Safety Decision Making. Dam Safety Office, United States Bureau of Reclamation. Denver, Colorado. USNRC. 1975. Reactor Safety Study: An Assessment of Accident Risks in US Commercial Nuclear Power Plants. United States Nuclear Regulatory Commission, NUREG75/014. Washington, D. C. Tufte, E.R. 1990. Envisioning Information, Graphics Press. Cheshire, Connecticut.

200 © 2011 by Taylor & Francis Group, LLC

2 Geohazards and gas hydrates

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Neotectonic deformation of northwestern Australia: Implications for oil and gas development J.V. Hengesh, K. Wyrwoll & B.B. Whitney The University of Western Australia, Perth, Australia

ABSTRACT: Although Western Australia is commonly viewed as a Stable Continental Region with low rates of earthquake activity, geological and geomorphological evidence indicates that active tectonic processes are occurring: (1) the north coast is accommodating crustal flexure due to the collision with the Banda Arc; (2) the central west coast exhibits evidence of active fold growth and reverse movement on reactivated normal faults; and (3) the Murchison region has clear evidence of Quaternary tectonic deformation, repeated surface rupturing events, and a record of two large magnitude historical earthquakes. The evidence of Neotectonic deformation in northwestern Australia indicates that a number of seismic sources are present and these sources have the potential to produce moderate to large magnitude earthquakes such as the Mw 7.1 Meeberrie event. Future seismic hazard assessments should implement refined seismic source models that treat distinct seismic sources and the epistemic uncertainty associated with those sources. These seismic sources should be considered in the selection and engineering of seabed infrastructure such as manifolds, flowlines, export pipelines, and anchorage systems, as well as onshore LNG plants and port facilities. 1

INTRODUCTION

Western Australia is commonly viewed as a “Stable Continental Region” (SCR). It is largely composed of Archean age terranes, such as the Yilgarn and Pilbara cratons, Proterozoic and Phanerozoic age basins such as the Kimberly, Canning, Carnarvon, and Eucla Basins, and intervening deformed belts such as the Albany Frazier, Capricorn, King Leopold, and Halls Creek orogens. There is no orogenesis (active mountain building) occurring in WesternAustralia (WA) and the landscape is severely weathered with deep regolith attesting to long term stability on a regional scale (Anand and Paine, 2002). However, the occurrence of large magnitude historical earthquakes such as the 1885 ML 6.6 Mt. Narryer, 1941 ML 7.1 Meeberrie, and 1967 ML 6.7 Meckering events, and geomorphic evidence of crustal deformation and fault scarps from previous earthquakes indicate that parts of WA are being actively deformed. EPRI (1994) established criteria to define SCRs, which include: (1) evidence for no tectonic activity younger than early Cretaceous (∼100 million years before present {Ma}); (2) no deformed forelands or orogenic belts younger than Cretaceous (∼65 Ma); (3) no anorogenic intrusions younger than Cretaceous; and (4), no rifting or significant extension younger than Paleogene (∼35 Ma). SCRs are divided into domains composed of extended and non-extended crust, and domains underlain by extended crust are further divided into continental margins and failed rifts. According to EPRI (1994), SCRs that are underlain by extended crust have greater seismogenic potential than

those underlain by non-extended crust.Although Western Australia broadly meets these criteria, regional tectonic warping as evidenced by coastal submergence in the north, folding, the presence of numerous fault scarps, and the occurrence of several large magnitude earthquakes suggest that tectonic deformation is occurring across parts of WA. Active crustal dynamic processes occurring in SCR’s remain an enigma within the earth sciences. Understanding these processes is important for characterizing the location, severity, and frequency of earthquake occurrence, and assessing potential triggers for submarine landslides, liquefaction, and site response for structural design of both offshore and onshore facilities. 2

The northern margin of the Australian Plate is involved in a complex collision with the Sunda and Philippine Sea plates (Figure 1). Relative motion of the Australian, Sunda, and Philippine Sea plates is constrained from repeated geodetic surveys that use Global Positioning System (GPS) satellites to measure the precise positions of survey points located throughout a region. The repeated surveys provide direct measurements of the rates and directions of motion of points on different tectonic plates. Reoccupation of GPS sites in the Pacific, Australia, Indonesia, and Southeast Asia between 1991 and 2003 indicate that: (1) theAustralian Plate is moving along an azimuth of 015◦ and is converging with the Sunda Plate/Banda Arc at a rate of

203 © 2011 by Taylor & Francis Group, LLC

REGIONAL TECTONIC SETTING

Figure 1. Regional tectonic setting showing major tectonic structures and relative motion vectors of tectonic plates.

67 to 75 mm/yr relative to a fixed Eurasian reference frame (Figure 1) (Bock et al., 2003; Nugroho et al., 2009); (2) the Philippine Sea plate is moving northwestward at a rate of approximately 110 mm/yr relative to Eurasia (DeMets, et al., 1994); and (3) the Sunda Plate is moving east-northeastward at a rate of about 7 to 11 mm/yr relative to Eurasia (Simons et al., 2007). The northern boundary of the Australian Plate follows the Sunda Arc subduction zone and the Banda Arc collision zone (Figure 1). The Australian Plate consists of two main parts including Australian continental crust and oceanic crust of the Indian Ocean, and the differences in crustal type control the nature of processes occurring along the plate boundary. Subduction, in the past, has occurred along the entire Sunda and Banda plate boundary, but now is limited to that part of the plate boundary where oceanic crust of the Indian Ocean is colliding with continental crust of the Sunda Plate (McCaffrey and Nabelek, 1984). Here, the thinner, denser oceanic crust is subducted northward beneath the thicker, less dense continental crust. Subduction extends from the Andaman Islands in the northwest to approximately 120–121◦ east longitude near the island of Flores (Audrey-Charles, 1975; Karig et al., 1987). East of this location, the oceanic crust of the Indian Ocean already has been consumed and subduction of oceanic lithosphere has ceased (Silver et al., 1983; Genrich et al., 1996). The plate boundary from Flores to East Timor is now characterized by collision of the Australian continental crust with fragments of the former island arc (e.g. Sumba, Rote, and Timor islands) and accretion of those fragments to the Australian Plate. The main deformation front now occurs along the northern side of the island arc on a system of south-dipping north-verging reverse faults that are referred to as the Bali-Flores and Wetar thrusts (McCaffrey and Nabelek, 1984). The formation of a south dipping thrust system on the north side of the island arc and similarities in motion vectors for both Australia and Timor (Genrich et al., 1996), indicates that the former island arc is being accreted to the

Australian Plate and the subduction zone has reversed polarity (compared to the Sunda Arc to the west). The collision of the Australian continental crust along the southern Banda Arc has caused profound changes in the style of deformation including cessation of north-directed oceanic subduction, accretion of the former island arc, and reversal of subduction polarity along the Flores and Wetar thrusts (east of 120◦ east). The collision of the Australian continental crust with the South Banda Arc also is causing warping and deformation of northern and northwestern Australia. Therefore, although the main active plate boundary lies on the north side of the Banda Arc, northern Australia is responding to the effects of the collision. We speculate that the transition from oceanic subduction to continental collision at approximately 120◦ east longitude is generating stresses in the Australian crust that are a source of strain energy for earthquakes observed in northern WA. Furthermore, although the main deformation front now lies on the north side of Timor, structures along the Timor trough may still be active earthquake sources. 3 TECTONIC DEFORMATION IN A STABLE CONTINENTAL REGION 3.1 Tectonic flexure Northeast directed convergence of the Australian Plate with the Banda Arc is causing downward flexure of the continental lithosphere in response to the collision. The flexure is most evident along the Kimberly and has formed an anomalously wide continental shelf and sinuous shoreline morphology consistent with crustal subsidence and coastal submergence. The continental shelf, shown on Figure 2, is over 500-km wide (at the position of East Timor) and decreases to less than 100-km wide near Exmouth (Cape Range). From Exmouth southward along the West Australian coast the continental shelf is typically between 40- and 100-km wide.

204 © 2011 by Taylor & Francis Group, LLC

Figure 2. Digital elevation model of the northwest Australian continental shelf. Note wide shelf between Australia and Timor and narrowing of the shelf to the southwest. Elevation data from Geosceince Australia (2002).

elevations to sea-level high and low-stands. Figure 3 shows a strong correlation of surface elevations to the sea-level curve and using the maximum ages predicts minimum subsidence rates of 0.22 to 0.25 mm/yr for the shelf area south of the Timor trough (Table 1 and Figure 4). The hinge line of this flexure lies south of Broome, where coastal morphology changes from wave erosional (stable) to sinuous (drowning), and last interglacial (Marine Isotope Stage 5e) shoreline deposits (circa 120–130 ka) are present to the south and appear to be absent to the north. If the last interglacial deposits are now submerged the minimum coastal subsidence rate in the vicinity of Broome would be approximately 0.05 mm/yr. This indicates that the tectonic flexure associated with Australia’s collision with the Banda Arc diminishes from north to south and reaches zero north of the Cape Range where coastal uplift is observed and Last interglacial marine units are at their expected heights (Kendrick et al., 1991). The tectonic flexure documented by subsidence of geomorphic surfaces indicates that the continental lithosphere of northwestern Australia is being actively deformed due to the collision with the Banda Arc. This deformation will cause strain in the crust, and therefore the area of tectonic flexure may have greater seismogenic potential than areas not undergoing flexure.

Figure 3. Sea-level prediction diagram showing correlation between surface elevations and sea-level stages. From top to bottom, geomorphic surface elevations are −25 m, −40 m, −50 m, −70 m, −85 m, and −125 m. Elevation from Van Andel & Veevers, 1965. Sea-level curve from Lambeck and Chappell, 2001. Upper and lower curves indicate uncertainty.

The anomalously wide shelf in the north is inferred to be the result of marine erosion during progressive tectonic subsidence. Van Andel & Veevers (1965) recognized a series of submergent shoals, shelfs and terraces and postulated that these could have formed through a combination of tectonic subsidence and sealevel fluctuations. However, at the time, the theory of plate tectonics had not yet emerged, data on sea-level fluctuations were sparse, and thus it was not possible to document the nature and rate of deformation. Van Andel & Veevers (1965) recognized six main submergent surfaces at elevations of −120 to 140 m (at the shelf break), −85 m, −70 m, −50 m, and −40 m (along the continental shelf), and −25 m at a series of submergent atolls on the shelf break (Karmt Shoals) (Figure 2).The geomorphic surface elevations are plotted on a sea-level prediction plot (Figure 3), which is tentatively used to correlate the geomorphic surface

3.2 Folding and faulting in the Cape Range The 320-km long section of coast from the Cape Range to Cape Cuvier exhibits evidence of Neogene uplift and folding.The folding and uplift have strong regional expression, giving rise to the Cape Range, Cape Rough and Giralia Ranges, as well as the Lake MacCleod basin. Each of these ranges is mapped as an anticlinal

205 © 2011 by Taylor & Francis Group, LLC

Table 1.

Correlation of geomorphic surface elevations and ages.

OIS stage

Sea-level (m)

Surface elevation (m)

Subsidence (m)

Age min. (ybp)

Age max. (ybp)

5e 5d 5b 3 3 2

3 −16 −30 −57 −75 −118

−25 −40 −50 −70 −85 −125

28 24 20 13 10 7

118000 106000 82000 54000 31000 19000

123000 108000 92000 60000 43000 28000

OIS = Oxygen Isotope Stage; ybp = years before present; m = metre

Figure 4. Subsidence rate diagram showing net subsidence and age values used in the rate calculation, and linear regression.

fold with intervening west-dipping reverse reactivated normal faults (Myers and Hawking, 1998). Evidence of uplift is spatially associated with the anticlinal structures. A series of at least four marine terraces occur along the west coast of the 120-km long N-NE trending Cape Range anticline. The presence of these elevated shoreline deposits indicates long-term emergence of the coast and active fold growth. The growth of the Cape Range anticline implies crustal shortening and movement along associated thrust faults. Coastal uplift also is expressed in the anomalous height of Pleistocene marine deposits exposed in coastal sections west of Lake MacLeod. Uplift of this section of coast appears associated with anticlinal structures, which has prevented a number of rivers from reaching the coast, and resulted in formation of the Lake MacLeod evaporate basin. In the Lake McLeod area, folding of Miocene and younger sedimentary deposits has formed the Gnargoo Range and diverted the Lyndon and Minilya rivers around the resulting anticlines (Figure 5). Development of these supercedent streams may indicate Quaternary uplift rates in excess of incision rates and can provide constraints on rates of fold growth and associated fault slip rates.

Figure 5. Neogene folding at Lake MacCleod. Black lines are fold axes. Streams flow around the noses of folds (supercedent) indicating fold growth controlled stream position. Modified from GSWA, 1985.

The crustal shortening and fold growth is likely associated with reverse movement along former extensional structures and is more pronounced in this region than elsewhere in the Carnarvon Basin. The geomorphic evidence for reactivation of former extensional structures indicates that these are potential seismogenic sources that should be explicitly considered in seismic hazard assessments. These structures follow the continental shelf break and cross much of the Exmouth Plateau (Myers and Hawking, 1998) and therefore may be near-field sources of ground shaking for offshore facilities. 3.3

The Mt. Narryer fault zone in central WA is located near the northwest margin of the Yilgarn craton and the eastern margin of the Southern Carnarvon Basin

206 © 2011 by Taylor & Francis Group, LLC

Surface faulting and historical earthquakes

Figure 6. Digital terrain model (30 m digital elevation data reproduced by permission of the Western Australian Land Information Authority, 2010) showing oblique view to northeast along the Mt. Narryer fault scarp at the Roderick River. Arrows point to fault scarp and line shows position of topographic profile on Figure 7.

Figure 7. Topographic profile across the Mt. Narryer fault showing offset of alluvial valley surface (2009 data reproduced by permission of Western Australian Land Information Authority, 2010).

(Williams, 1979). The fault is approximately 120-km long and strikes in a northeast direction. Historical reports of earthquake strong ground shaking suggest that the fault zone may have produced the 1885 ML 6.6 Mt. Narryer earthquake (Clark, 2006). The epicenter for the 1941 ML 7.1 earthquakes also is located near the fault zone and so it is likely that the Mt. Narryer fault zone has produced two historical large magnitude earthquakes. The 1941 ML 7.1 Meeberrie earthquake is the largest earthquake to have been recorded in Australia. The Mt. Narryer fault zone includes at least four left-stepping en-echelon fault segments. From north to south the segment lengths are 11 km, 33 km, 40 km, and 35 km. The northern fault segments are expressed by strong vegetation alignments and fault scarps on the order of 1 to 1.5 m high (Clark, 2006).The linear nature and subvertical dip of the northern scarps suggests a significant strike slip component of motion. The two southern segments of the fault zone are expressed by a west-side up reverse sense of displacement and have formed east facing scarps across the Roderick and Sanford river alluvial valley deposits (Williams et al., 1983; Myers, 1997; Clark, 2006). Analysis of imagery and digital terrain models indicate that the scarps across the Sanford and Roderick rivers have

captured and diverted active stream flow, formed sag ponds, and impounded Lake Wooleen.The alluvial surfaces in both valleys are uplifted, warped and incised (Figures 6 and 7). Folding in the hanging wall of the fault zone has caused uplift and abandonment of the main river channel and incision of the river through the fold. Where the river cuts through the fold the channel pattern changes from braided to incised. Scarp heights of 3 to 8 m (Figure 7) suggest that the fault has experienced multiple surface rupturing events in Quaternary time. Preliminary analysis of drainage patterns and stream profiles west of the Mt. Narryer fault zone suggest that additional fault scarps may be present. 4

Although Western Australia is commonly viewed as a Stable Continental Region with low rates of earthquake activity, geological and geomorphological evidence indicates that active tectonic processes are occurring: (1) the north coast is accommodating crustal flexure due to the collision with the Banda Arc; (2) the central west coast exhibits evidence of active fold growth and movement on reactivated normal

207 © 2011 by Taylor & Francis Group, LLC

IMPLICATIONS FOR OIL AND GAS DEVELOPMENT

faults; and (3) the Murchison region has clear evidence of Quaternary tectonic deformation, repeated surface rupturing events, and a record of two large magnitude historical earthquakes. The Mt. Narryer fault may be an analogy for the types of earthquakes that can occur on the reactivated normal faults along the Cape Range and Exmouth Plateau. In this case, a magnitude 7 or greater earthquake is a possible scenario for faults within 50 to 150 km of major offshore production facilities and onshore processing plants. Depending on the activity rates applied to these potential seismic sources, the incorporation of specific fault sources (in seismic source models for the region) could result in large contributions to the overall ground motion hazard at a site. The presence of active seismic sources also is an important consideration for evaluation of surface fault rupture hazards, selection of time histories used in site-specific site-response analysis, analysis of slope instability, and assessment of liquefaction potential. These seismic sources provide triggering mechanisms for instability and permanent ground deformation and should be considered in the selection and engineering of seabed infrastructure such as manifolds, flowlines, export pipelines, and anchorage systems, as well as onshore LNG plants and port facilities. REFERENCES Anand, R. R. and M. Paine, 2002. Regolith geology of the Yilgarn Craton, Western Australia: implications for exploration, Australian Journal of Earth Sciences 49, 3–162. Audley-Charles. M.G., 1975. The Sumba fracture – A major discontinuity between eastern and western Indonesia, Tectonophysics, 26, 213-228. Bock, Y., Prawirodirdjo, L., Genrich, J.F., Stevens, C.W., McCaffrey, R., Subarya, C., Puntodewo, S.S.O., and E. Calais, 2003. Crustal motion in Indonesia from Global Positioning System measurements. Journal of Geophysical Research 108 (B8), 2367. Clark, D. J., 2003. Reconnaissance of recent fault scarps in the Mt Narryer region, W. Australia, Minerals & Geohazards Div. Earthquake Hazard & Neotectonics Group, Canberra. DeMets, C., Gordon, R.G., Argus, D.F., and S. Stein, 1994. Effect of recent revisions to the geomagnetic reversal time scale on estimates of current plate motions. Geophysical Research Letters 21, 2191–2194. Genrich, J.F., Bock, Y., McCaffrey, R., Calais, E., Stevens, C.W., and C. Subarya, 1996. Accretion of the southern Banda arc to the Australian plate margin determined by

Global Positioning System measurements. Tectonics 15, 288–295. EPRI, 1994. The earthquakes of stable continental regions. Volume1: Assessment of large earthquake potential. Report prepared for Electric Power Research Institute by Johnston, A. C., Coppersmith, K. J., Kanter, L. R. and C. A. Cornell. Geological Survey of Western Australia, 1985. Geology of the Carnarvon Basin, 1:1,000,000, Bulletin 133, Plate 1. Geoscience Australia, 2002, Australian bathymetry and topography grid, CDROM. Karig, D.E., Barber, A.J., Charlton, T.R., Klemperer, S.E., and D.M. Hussong, 1987. Nature and distribution of deformation across the Banda Arc–Australian collision zone at Timor, Geological Society of America Bulletin 93, 18–32. Kendrick, G.W., Wyrwoll, K.-H. and B.J. Szabo, 1991. Pliocene-Pleistocene coastal events and history along the western margin ofAustralia. Quaternary Science Reviews, 10, 419-439. Lambeck, K. and J. Chappell, 2001. Sea level change through the last glacial cycle, Science 292, 679. Landgate, 2009, 30 m digital elevation data, geospatial data CDROM; www.landgate.wa.gov.au. McCaffrey, R, and J. Nabelek, 1984. The geometry of back arc thrusting along the eastern Sunda are, Indonesia: Constraints from earthquake and gravity data, Journal of Geophysical Research, 89, 6171–6179. Myers J.S., 1997. Byro, WA, Sheet SG 50-10 (2nd edition): Western Australia Geological Survey 1: 250,000 Series. Myers, J.S. and R.M. Hocking, 1998. Geological map of Western Australia, 1:2,500,000 (13th ed.), W. Australia Geological Survey. Nugroho, H., Harris, R., Lestariya, A.W., and B. Maruf, 2009. Plate boundary reorganization in the active Banda Arc–continent collision: Insights from new GPS measurements, Tectonophysics, 479, 52–65. Silver, B.A., D.R. Reed, R McCaffrey, and Y. Joyodiwiryo, 1983. Back arc thrusting in the eastern Sunda are, Indonesia: A consequence of arc continent collision, Journal of Geophysical Research, 88, 7429–7448. Simons, W. J. F., Socquet, A., Vigny, C., Ambrosius, B.A.C., Abu, S.H., Promthong, C., Subarya, Sarsito, D.A., Matheussen, S., Morgan, P. and W. Spakman, 2007. A decade of GPS in Southeast Asia: resolving Sundaland motion and boundaries. Journal of Geophysical Research 112, B06420. Van Andel, T.H. and J.J. Veevers, 1965, Submarine morphology of the Sahul Shelf, Northwestern Australia, Geological Society of America Bulletin, v. 76, p. 695–700. Williams, I. R., 1979. Recent fault scarps in the Mount Narryer area, Byro 1:250,000 sheet.: Western Australia. Geological Survey. Annual Report 1978, v. 51–55. Williams, I. R., Walker I. M., Hocking R. M., and S. J. Williams, 1983. Byro, W. Australia. W. Australian Geological Survey, 1:250,000 Geological Series Explanatory notes, p. 25p.

208 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Deepwater Angola part I: Geohazard mitigation A.J. Hill & J.G. Southgate BP Exploration, Sunbury-on-Thames, UK

P.R. Fish Halcrow Group Ltd., Birmingham, UK

S. Thomas Fugro Geoconsulting Ltd., Wallingford, UK

ABSTRACT: BP Angola and its equity partners have embarked upon a rolling programme of subsea developments tied back to multiple FPSOs with future gas export to an LNG plant onshore. The programme concept has meant that BP’s Geohazard Assessment Team has been able to build and refine ground models for each development area. The key geohazards have been evaluated at the appropriate time within the project cycle allowing facilities engineers, drilling and subsurface teams to plan wells and subsea layouts that avoid or mitigate against geohazards where necessary. This paper provides specific examples of geohazards that commonly occur in deep water offshore Angola such as pockmarks, shallow gas and gas hydrates, faults, and seabed and sub-seabed slope instability. It also discusses some atypical features which have been encountered by BP and how they have been investigated.

1

INTRODUCTION

1.1 Background BP’s proposed programme of developments offshore Angola has provided an excellent opportunity for a phased assessment of geohazards and construction of a regional ground model to manage geotechnical risk. The ability to synchronise shallow geophysical and geotechnical data collection with the key stages in a project development (which BP terms APPRAISE, SELECT, DEFINE, EXECUTE and OPERATE) provides the best value to projects (Fig. 1). This is particularly true in relatively remote regions where mobilisation of survey vessels is normally associated with a long lead time. A programme of developments allows, for example, opportunistic data collection at one site whilst carrying out a reconnaissance level investigation at another site nearby. Similarly, a more mature project can benefit from more targeted supplementary information acquired in conjunction with a reconnaissance investigation in an adjacent area. The advantages of a phased approach are well understood by engineers and geoscientists (Campbell 1984). However, opportunities for implementing such a plan are often limited for a variety of commercial, political and logistical reasons. This paper describes how the phased approach adopted by BP Angola has identified key geohazards and provided an evolving ground model at a resolution that matches the project requirements as they develop.

209 © 2011 by Taylor & Francis Group, LLC

Figure 1. Geotechnical Engineering and Geohazard Mitigation.

BP Angola has formalised the documentation of this process in terms of Geotechnical Engineering and Geohazard Mitigation (GGM) reports which are produced at key milestones; when new data are acquired

or the proposed infrastructure changes significantly. The GGM reports act as baseline information which reflect the project team’s collective state of knowledge of the shallow sub-surface geotechnical conditions and geohazards at a given time. The reports are used to ensure that contractors bidding for work do so on an equal footing and that subsequent designs are consistent. Ultimately the reports provide a reference for seabed-related problems throughout the life of the development through to decommissioning. Figure 1 illustrates the relationship between the project schedule, geo-data collection and the production of GGM reports. Figure 2. Typical pockmark field (hillshaded DEM).

2

GEOHAZARD MANAGEMENT

BP Angola’s overall strategy for management of geohazards is to avoid significant geohazards; if this is not cost effective or practical, the strategy is to implement design mitigations or accept the risks where they are fully understood and tolerable. Ground conditions which are less predictable using a regional model should also be avoided. However, if that is not possible or cost effective, the design must accommodate a wider range of soil conditions. In the case of high-risk structures, the design may need to be supported by additional location-specific data. As well as describing the seabed conditions close to, and at a distance from, locations of geotechnical investigations, a ground model also facilitates the selection of the right scope and tools and techniques to be employed for subsequent surveys and site investigations.

3 3.1

Table 1. Likelihood of activity

Scale of Score activity

Relict Periodically active Recently active Probably active Active

1 2

FEATURE MAPPING AND CHARACTERISATION Deepwater Angola geohazards

This paper provides specific examples of geohazards that commonly occur in deep water offshore Angola such as pockmarks, shallow gas and gas hydrates, faults, and seabed and sub-seabed slope instability. It also discusses some atypical features which have been encountered by BP and how they have been investigated. Examples of these are: a narrow trough feature, which is an extreme example of one of the many linear bedforms; the presence of unusually stiff clays near to the seabed; and the products of hydrocarbon migration which in some cases have reached the seabed. More detailed commentary on the engineering challenges that these geohazards pose is provided in a related paper at this conference (Hill et al. 2010). Some geohazards have had a direct influence on field layout and engineering design and some, so far, have not. There is an inter-relationship between some geohazards, such as the association of pockmarks with faults and salt diapirs, which act as flow-paths for fluid expulsion; the position of gas hydrates and the

210 © 2011 by Taylor & Francis Group, LLC

Pockmark hazard severity classification.

3 4 5

No seepage Micro-seepage Gentle bubbling Vigorous bubbling Significant outbursts Eruptions of gas and sediment

Hazard Score *Result level 0 1 2

<6

Low

3

6–8

Medium

4

>8

High

5

*Result is a multiple of likelihood and scale of activity scores

thermal conductivity of salt; and salt diapirism and the presence of stiff clays near the seabed. The following sections provide a brief overview of some key geohazards offshore Angola and how the risks they pose have been managed in a timely manner throughout the programme of subsea developments. 3.2

Pockmarks

Pockmarks are conical depressions in the seabed formed by fluid expulsion which may be hundreds of metres wide and tens of metres deep (Fig. 2). They are commonly found offshore Angola where potential hazards include expulsion of corrosive fluids and slope instability (Judd and Hovland 2009). In order to rationalise the constraint they impose on facilities, a method has been developed that assesses the hazard severity of individual pockmarks leading to specific guidance for hazard mitigation (Table 1). From the geospecialists that have reviewed pockmarks offshore Angola for BP, the consensus view for placement of seabed facilities nearby is to adopt a nominal 100 m stand-off from the rim (defined by the 5◦ slope angle ‘contour’) of pockmarks of medium and high hazard severity and that low hazard pockmarks

Figure 4. Fault types and relationships to salt.

sediments. This has been proven using CPTs fitted with a temperature element and is one explanation for an elevated base of the GHSZ close to the salt diapir. Figure 3. Relationship between pockmarks, gas and faults.

3.4 are just topographic constraints. Medium and high hazard pockmarks differ in the level additional assessment required to locate structures in stand-off zones. Additional soils data collection is likely to be required for high hazard pockmarks. The pockmark assessment has demonstrated that understanding their relationship with shallow gas and faults is central to determining their impact. High hazard pockmarks not only have an underlying gas pocket, but also a migration pathway along a fault or other plane of weakness such as the edge of a salt diapir (Fig. 3). 3.3 Shallow gas and gas hydrate Shallow gas and gas hydrate occurrences offshore Angola are attributed to a number of mechanisms: gas trapped beneath gas hydrate within thick PlioPleistocene sag basins, gas migrating from reservoir sources via faults or fractures intersecting structural traps, or gas trapped in sandy channel deposits within Miocene sediments. Irrespective of their sources, all shallow gas and hydrates are potential threats to drilling operations and to seabed founded structures, and therefore require systematic mapping. Detailed mapping of bottom simulating reflectors, believed to represent the base of the gas hydrate stability zone (GHSZ), have shown they are found at variable depths below seabed. This variation is thought to relate to local variations in environmental factors such as increases in water salinity, changes to the geothermal gradient around salt diapirs, and chemical composition of the shallow gases forming the hydrate. Recognition of these local factors emphasises the need for detailed mapping, rather than reliance on modelled relationships between base of GHSZ and water depth. One feature of the presence of salt (discussed later) is its relatively high thermal conductivity which elevates the geothermal gradient in the surrounding

211 © 2011 by Taylor & Francis Group, LLC

Faults and seismic hazard

Faulting evident offshore Angola can be broadly classified into two main types: very shallow polygonal faults, common in thick Plio-Pleistocene clays and caused by dewatering (Cartwright and Lonergan 1996); and faults driven by the extensional regime imposed by salt movement which generally extend no more than 7 km deep. Salt-related faults are classified depending on their relationship to the underlying salt: domino faults sole out on the ductile surface of the salt and are formed by lateral salt movement; flap faults are extensional and occur at the margins between subsiding basins and rising salt diapirs; and keystone faults, which are associated with anticlinal folding from salt uplift (Fig. 4). BP’s concessions are some distance from plate boundaries and consequently no faults are considered to be of tectonic origin (Fugro West, 2008). Measured displacements and calculated displacement rates of less than 1 mm/year (Fig. 5) suggest that there will be no significant movements across these faults during the field life. This is due to very low rates of sedimentation and slow movement of salt. It was concluded in the early SELECT phase that there was very low risk of fault displacement impacting seabed structures, but that there remains a residual risk that human intervention (such as reservoir depletion) may trigger movement along pre-existing faults. The seismic hazard is from background and onshore sources. Seismic hazard analyses have estimated peak ground accelerations of 0.015 g and 0.15 g for 200year and 2,900 year return periods respectively. It is concluded that seismic hazard is not likely to be significant but key structures should be assessed against the full design spectrum according to the period of the structure and foundation. It was concluded that for the development areas in question, no stand-off zone was needed with respect to faults. The reactivation of existing faults is considered unlikely and generation of new faults even more so in the project lifetime.

Figure 6. Mass transport deposits and steeply dipping beds above salt diapir (chirp data).

Figure 5. Fault movement rates.

3.5

Seabed slopes and buried mass transport deposits

Active landslides have not been widely observed on the seabed, except for small failures on the oversteepened flanks of salt diapirs, but seismic data indicate that Mass Transport Deposits (MTDs) are commonly observed in the sedimentary sequence of sag basins between salt diapirs (Fig. 6). This observation suggests that the very gradual uplift of the salt leads to over-steepening, and eventual failure, of slopes. Furthermore, it indicates that slope failure cannot be discounted over the life of field, especially if slopes are loaded by pipelines or other seabed facilities. Consequently site-specific slope stability assessment has been adopted (Hill et al. 2010). MTDs pose a drilling hazard associated with wellbore instability and require assessment on a well-specific basis. Figure 6 also shows steeply dipping beds of older and stiffer clays pushed towards seabed surface by rising salt, which may cause unexpectedly strong materials to be encountered by shallow foundations. 3.6

Narrow trough and other bedforms

Bedforms in a variety of morphologies are very commonly seen in ultra deep water offshore Angola, but they are not generally considered to be hazardous because of their low amplitude and extremely slow rate of migration. One exception is a particularly large bedform (up to 250 m wide, 30 m deep and over 15 km long) and has been separately described as a narrow trough. This feature was identified as a significant geohazard risk driver to the development because it bisects two potential drill centres. Avoidance would have meant a pipeline reroute of over 10 kilometres. There was also

Figure 7. Narrow trough conceptual model.

a concern the feature may result in enhanced potential for erosion or sedimentation and slope instability (Fig. 7). Opportunistic sampling and deployment of seabed current monitoring equipment meant that these concerns could be addressed in the project’s SELECT phase. Sedimentation rates calculated from geochronological tests (14 C and optically-stimulated luminescence) on the flanks and thalweg of the narrow trough were low. Data from both locations indicate no significant active erosion but slightly higher rates of deposition on the flanks of the trough relative to its centre (0.2 mm/yr on flank and 0.08 mm/yr in thalweg). Current meter data show that bottom currents associated with trough are also low (0.04 m/s), with little potential for scour. The more common bedforms are also static features, which are not expected to change in size or migrate over the life of field. They do not require a stand-off

212 © 2011 by Taylor & Francis Group, LLC

Figure 8. Conformable and cross-cutting HIRs and seabed manifestation as mounds.

zone and are represented as topographic features only which may pose flow assurance constraints. However, the size of the narrow trough is exceptional, meaning it has potential slope instability and flowline spanning issues. Based on these data, it is concluded that the narrow trough is a static feature acting as a topographic constraint to facilities, and its main impact relates to potential slope instability (Hill et al. 2010). 3.7 Hydrocarbon migration Seabed photographs taken during environmental surveys show dark-coloured hydrocarbon extrusions and associated paler-coloured suspected carbonates. The features were also mapped from multibeam echosounder (MBES) data where they were shown to be widespread, and up to 20 m across and 2 m high. Analysis of AUV (chirp) sub-bottom profiler data shows that the subsurface extent of the features is more extensive than their seabed expression, and a High [acoustic] Impedance Reflector (HIR) has been mapped across the development area. There were examples where the HIR is conformable with the bedding and others where it cuts across the beds (Fig. 8). These features were first observed during RoV surveys shortly before they were also identified on side-scan sonar, multibeam echosounder and subbottom profiler data. The reconnaissance geotechnical investigation was then able to target these features using gravity core and seabed CPT systems. Gravity core samples were extremely high strength, and CPT refusal occurred at the intersection of some, but not all, HIRs. More detailed investigation of these features was then scheduled using additional RoV surveys and geotechnical drilling during a supplementary investigation. This was another example of the phased approach to building a ground model where detailed geotechnical investigations were informed by earlier reconnaissance-level investigations.

All of this work was completed prior to DEFINE which prevented it becoming a potentially costly critical-path issue later in the development cycle. Subseabed occurrences of carbonate-rich claystones have been reported by other operators elsewhere offshore Angola, but not as extensively or with such visible seabed manifestations. Following analysis of samples recovered from geotechnical boreholes, it is thought that the HIR is indicative of the stratigraphic position where rising hydrocarbons have interacted with the in situ sediments over geological time resulting in the precipitation of carbonate. In some cases the hydrocarbon reaches the seabed and forms extruded mounds. The majority of samples across the HIR show a predominance of carbonate, mainly in the form of thin, hard carbonate-rich claystone beds or nodules with some intra-bed voids. The hard carbonate-rich claystone beds or nodules are typically only a few centimetres thick and nearly all the carbonate has at least a small component of remnant hydrocarbon associated with it. In some examples there is still a significant proportion of hydrocarbon associated with the hard carbonates. These hydrocarbons have been recovered in the geotechnical samples as either asphalt or as thick oil. Seabed extrusions of asphalt and carbonate mounds are static features which may cause abrasion of flowline coatings, and therefore constrain routing. The corridor width needed for installation of flowlines is typically a few tens of metres and therefore individual mounds have been mapped giving field planners the opportunity to pass through an area of mounds while avoiding individual features. 3.8

213 © 2011 by Taylor & Francis Group, LLC

Salt diapirism

Salt diapirism is not considered a geohazard in itself. However, salt movement (which is predominantly upwards in the shallower water areas and has a more lateral component in the deeper water) can impact the field layout and foundation designs. Salt uplift and subsequent erosion of the seabed causes stiff to very stiff clays of older strata to be thrust upwards such that clays with undrained shear strengths of greater than 150 kPa have been sampled within 1 m of the seabed. Again, reconnaissance geotechnical investigations provided an early indication of these conditions and supplementary soils data combined with the engineering-quality AUV data enabled the extent and characteristics of these materials to be well constrained. More details on how salt diapirism influences the geotechnical zoning, (known as soil provinces) of the shallow subsurface are provided in Hill et al. (2010). The steep slopes formed by this uplift may be over 20 degrees, which may constrain pipeline routing and can give rise to slope instability, particularly in those areas where recent sedimentation forms a drape over the older materials. The in situ stress regime is also altered by the salt movement, giving rise to tension

cracks and faulting in the most active areas and reorientation of the principal stress directions elsewhere. This can influence the radial stresses providing support to piled foundations. 4

CONCLUSIONS

This paper describes a phased approach to geohazard and geotechnical characterisation used by BP Angola to develop a regional ground model for optimising field layouts and engineering wells and facilities. This strategy has been shown to be highly effective with significant advantages over less systematic methods. It has allowed the project team to identify the potentially high-impact geohazards early in the development cycle; to focus on the geohazards that are most relevant for the planned developments; and to avoid rework and threats to the project schedule by achieving geohazard-tolerant field layouts early. Specifically, this approach has: enabled pockmarks to be ranked according to their severity; to understand the threat posed by shallow gas; discount the impact of faults and seismic hazards; quantify the risk of slope failure; allowed the narrow trough to be classified as a topographic constraint rather than a dynamic feature; provided an understanding of the nature of hydrocarbons and their by-products in the shallow surface; and to identify clearly the areas of atypical soils due to the effects of salt diapirism. Some of the geohazards described in this paper are relatively rare in the developed areas offshore Angola, but others are ubiquitous. As development reaches ever-increasing water depths the authors believe that for similarly geohazardous areas, the systematic approach adopted by BP Angola for evaluating the impact of geohazards is required. There remains however, a need for the approach to managing geohazards to be tailored to project and site specifics.

5

This paper represents the views of the authors and not necessarily those of the companies they represent. The conclusions relate to specific aspects of some of the geohazards present offshore Angola hopefully of interest to the geotechnical community. The authors wish to thank BPAngola, their development partners and Sonangol for permission to publish this material. They are also grateful for the critical reviews, guidance and stimulating conversation provided by Trevor Evans, Mike Sweeney and Mike Fiske.

214 © 2011 by Taylor & Francis Group, LLC

CLOSING REMARKS AND ACKNOWLEDGMENTS

REFERENCES Campbell, A. J. 1984. Predicting Offshore Soil Conditions. Proceedings Annual Offshore Technology Conference Houston, Texas, 7–9 May, OTC paper Number 4692. Cartwright, J.A. & Lonergan, L. 1996.Volumetric contraction during the compaction of mudrocks. Basin Research 8: 323–331. Dendani, H., Colliat J.L., Puech,A. and Nauroy, JF 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences Keynote Paper, Proceedings 2nd International Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Evans, T.G. 2010. A Systematic Approach to Offshore Engineering for Multiple-Project Developments in Geohazardous Areas, Keynote Paper, Proceedings 2nd International Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Fugro West (2008) Seismic Hazard Analysis Block 31, Offshore Angola, Issued to BP Angola December 2008. Doc No. 3193.034 Hill, A. J., Evans, T.G., Mackenzie, B & Thompson, G, 2010. Deepwater Angola Part II: Geotechnical Challenges, Proceedings 2nd International Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Judd, A. & Hovland, M 2009. Seabed Fluid Flow. Cambridge: Cambridge University Press.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Deepwater Angola part II: Geotechnical challenges A.J. Hill & T.G. Evans BP Exploration, Sunbury-on-Thames, UK

B. Mackenzie Fugro Geoconsulting Ltd., Wallingford, UK

G. Thompson Senergy Survey & GeoEngineering, Bath, UK

ABSTRACT: BP Angola and its equity partners have embarked upon a rolling programme of subsea developments tied back to multiple FPSOs with future gas export to an LNG plant onshore. This paper discusses the geotechnical challenges related to geohazards encountered in those development areas and the benefits of a properly constructed ground model in regions that are geologically complex. Examples include: the stability of seabed slopes in and around salt diapirs; and the behaviour of deep foundations in stiff clays up-thrust by salt and in soils modified by hydrocarbon migration.

1

INTRODUCTION

encountered offshore Angola and how they have been managed by BP and its partners.

1.1 Background Hill et al. (2010) and Evans (2010) describe the geohazards identified by BP offshore Angola and the value of constructing a predictive ground model as an effective way of mitigating those hazards. This paper describes how the ground model has been used to provide guidance for geotechnical engineering purposes, with particular reference to suction caissons, driven pile foundations and seabed slope stability. There are many other geotechnical challenges common to deepwater developments in more typical normally consolidated clay environments, but this paper is restricted to those key foundation design issues associated with geohazards. The ground model approach seeks to adequately define a 3D block of soil in a proposed development area. If the ground model is constructed properly, much of the geophysical and geotechnical data acquisition can be done early in a project cycle. It also provides flexibility for changes in field layouts and limits the amount of geotechnical investigation required after the seabed facilities architecture is frozen. Timing the investment in surveys, site investigations and subsequent analysis is critical. The key here is to recognise the complexity of a site (or conversely the simplicity, in some areas) at the right stage in the development cycle. The goal is to produce a ground model that is fit-for-purpose: neither overly complex, nor insufficiently resolute to design reliable foundations, anchors and flowlines. This paper describes the impact of some of the more challenging geotechnical conditions

2 2.1

General

The geotechnical conditions offshore Angola in freefield areas (i.e. those outside the influence of geohazards) are typical of deepwater hemipelagic environments. With the exception of a slightly stronger “crust” within 1m of the seabed, the clays are generally lightly over consolidated to normally consolidated with a shear strength gradient of 1–1.5 kPa/m. The submerged unit weights of the clays are very low (typically only 2–3 kN/m3 in the upper 20 m). As part of the ground modelling process, the shallow subsurface is divided into soil provinces which describe areas or zones within which the geotechnical conditions are interpreted to be broadly similar for engineering purposes (Fig. 1). The criteria for defining the boundaries that separate soil provinces are based on foundation engineering considerations and specifically the inferred geotechnical conditions within 35 m of the seabed. The idealised boundaries between soil provinces are defined by projecting the intersection points of the 35 m depth contour and key seismostratigraphic horizons (Fig. 2). The selected seismostratigraphic horizons were chosen because they represent potentially significant changes in geotechnical conditions which could impact the suitability of suction-installed caissons, currently the most common deep foundation type for BP Angola’s

215 © 2011 by Taylor & Francis Group, LLC

GEOTECHNICAL CONDITIONS

Figure 1. Soil province map.

Figure 3. Undrained shear strength in each soil province.

occupy the sag basins between salt diapirs and topographic highs created by movement of the underlying salt. Soils in the sag basins are generally uniformly normally to lightly over consolidated with physical characteristics that are reasonably predictable over large areas with a sufficiently advanced ground model. Conversely, soils on the topographic highs that have been uplifted by salt can exhibit significant variability, ranging from very soft to hard clays with possible presence of sand layers. The soils become increasingly less predictable as the influence of salt increases, as shown by the widening undrained shear strength design envelopes in Figure 3.

Figure 2. Definition of soil provinces.

developments. The 35 m depth was chosen because this represents the maximum expected suction caisson penetration with a small margin for tolerances. Each soil province is characterised by a layered sequence of geotechnical units with associated characteristic soil parameters. The soil provinces represent a simplified and idealised model for what are complex conditions. Although the simplified models show the boundaries between soil provinces as sharp breaks, they actually represent transition zones. During the early stages of characterising a development area, the soil provinces are largely based on the interpretation of the 3D exploration geophysical data but as higher resolution (engineering quality) geophysical and geotechnical site investigation data are acquired the soil provinces are refined and engineering guidance developed. 2.2

Impact of salt diapirism on soil provinces

In some parts of BP Angola’s development areas, the soil conditions are heavily influenced by salt diapirism. Salt movement, both uplift and lateral spread, influences the seabed morphology and features, the stress regimes and the engineering properties of the soils. Soil provinces have broadly been defined by areas that

2.3

The results of sub-bottom geophysical (chirp) profiling revealed an anomalous semi-continuous High [acoustic] Impedance Reflector (HIR) in some areas of the ground model. This HIR was within the depth of interest for foundations and anchors and was considered sufficiently important to justify additional investigation. Physical samples, in situ tests of the seabed and ROV surveys, improved the understanding of the geotechnical significance of the HIR. Most recently, geotechnical boreholes have been performed to examine the sub-surface properties of this feature and have developed our understanding even further.

216 © 2011 by Taylor & Francis Group, LLC

Impact of hydrocarbon migration

Figure 4. Geohazard mitigation process.

Figure 5. Foundation guidance – soil provinces.

When only geophysical data and information from seabed testing were available, advice to subsea engineering teams was necessarily cautious, and mitigation by avoidance was the first option. With the benefit of the new data, it was concluded that the HIR need not necessarily be avoided if driven piles were used rather than suction caissons. However, it would need to be proven that the in-place performance of foundations could be reliably predicted in the materials beneath the HIR. This is discussed further in Section 3.

comprise largely predictable normally consolidated to lightly over consolidated clays, the guidance was that suction-installed and driven piles may be used and that additional structure-specific soils data may not be required. The advice for Soil Province 3 was that suction-installed caissons and driven piles may be used, although structure-specific samples and/or CPTs would be required for design optimisation. In Soil Province 4, suction-installed caissons were not recommended because the older, stiffer and more variable soils may be close to the seabed. The earlier recommendation for driven piles in Soil Province 4 was that they may be used, but that structure-specific soils data would be needed. At the time of writing, a series of geotechnical boreholes has just been completed. The guidance for placement of caissons and piles is being refined but is likely to conclude that driven piles may be used in all soil units and even in some places where the pile may encounter a High Impedance Reflector (see Section 2.3). It is also likely that suction caissons can be used even where there are over consolidated clays near to the seabed, but only where HIRs are absent.

3

ENGINEERING GUIDANCE

3.1 Field layout planning Experience has shown that one of the most effective ways of communicating the rationale behind geohazard mitigation is by use of simple flowcharts such as the one shown in Figure 4. It presents the stepwise guidance for mitigation by avoidance or mitigation by design (with or without additional data collection) with respect to first geohazards (Stage 1) and then geotechnical conditions (Stage 2). The guidance is modified for different seabed structures and foundation types.

3.3 3.2 Suction caissons and piles near salt As our understanding of the effects of salt diapirsm on soil properties develops, engineering guidance for facilities planning and foundation design is evolving to reflect that. The effect of this is that the guidance provided to the facilities engineers becomes less generic, allowing greater flexibility for layouts and the use of different foundation types. Initially the uncertainties in soil properties of some of the older materials led to a conservative strategy of avoidance. This meant that it could not be relied upon that foundations could be installed through certain seismostratigraphic horizons. Now that the ground model is better-calibrated with deeper geotechnical samples, avoidance is no longer the only option. Figure 5 shows the foundation guidance when geotechnical data from only piston cores and seabed CPTs were available. For Soil Provinces 1 and 2, which

217 © 2011 by Taylor & Francis Group, LLC

Suction caissons and piles in soils modified by hydrocarbons

High Impedance Reflectors (HIRs) were initially thought to pose a number of constraints for facilities and for the suitability of the deep foundation options being considered. The nature of this geohazard was investigated to a limited extent during a reconnaissance site investigation. However the geotechnical seabed tools specified were not designed to penetrate strong materials to sufficient depth to provide data for deep foundations. Geotechnical boreholes integrated with downhole wireline logging were performed at strategic locations (selected using the chirp sub-bottom data) to calibrate the ground model and refine engineering guidance relative to the HIRs. The HIRs have seabed manifestations which are uneven mounds up to 2 m in height, closely spaced in some areas and as discrete features elsewhere. Subsurface HIRs range from one to a series of tabular

which may be problematic for pile driveability, and suction caisson installation. The HIRs are characterized by high CPT cone resistance values or, where sampled, carbonate-rich claystone. Such layers are fully expected to offer increased penetration resistance to driven piles or suction-installed caissons. In the study, HIR hardness and thickness were varied parametrically within credible bounds that were informed by the site investigation and a geological/geochemical review of the material. This led to a maximum modelled thickness and hardness of 2.0 m and 90 MPa respectively. In this context, ‘hardness’ is represented as equivalent CPT cone resistance. Pile driveability results showed a strong dependency on the depth at which the HIR is encountered. For example, assuming the lowest expected background soil strength profile typical of Soil Province 4, an 84-inch pile can be driven through a 0.6 m thick, 90 MPa hard HIR with a Menck MHU 270T hammer, if encountered at 20 m depth. However, if the same pile encounters the same HIR at 70 m depth, it cannot be driven through the layer. This prediction reflects the increasing amount of driving energy taken out of the pile by the soil as it penetrates, such that when driven to 70 m, there is insufficient remaining energy reaching the tip to advance it through the HIR. Predicted driveable combinations of HIR thickness and hardness can be represented graphically (Fig. 6). For suction caisson installation, the study was extended to consider inclination of the HIR, and it was shown that significant problems may be experienced should the caisson encounter an inclined or localised HIR. For example, should a 5.0 m diameter caisson encounter an inclined, moderate strength HIR of 10 MPa at a depth 1.0 m below seafloor, a HIR thickness of as little as 5 cm may cause the caisson to tip during installation. However, if a similarly hard HIR is encountered at 5.0 m depth, then the additional fixity of the caisson at this deeper penetration means it can tolerate a greater HIR thickness before tipping, in this case 20 cm or greater.

horizons conformable with the bedding, to those which cross-cut bedding and are more chaotic in nature. There are not always direct correlations between subsurface and surface features, and therefore seabed mapping cannot be used to determine the presence or absence of the sub-surface HIR. The materials forming the HIR are inferred to be varying degrees of the host soil, gas, oil and asphalt, and carbonate-rich cemented hard clay, and are found as both continuous horizons and discontinuous nodules. Geophysical surveys and geotechnical investigations revealed variable characteristics and properties, ranging from laterally continuous indurated features (causing CPT refusal), to undulating features laterally discontinuous over short distances.The hard layers targeted and encountered during the geotechnical site investigations were between 0.1 m and 0.7 m thick. These hard materials are thought to be authigenic carbonates formed by two primary geological and physicohemical processes: 1. Depositional carbonate formation in the older sediments as alternating layers of relatively carbonaterich and low carbonate sediments, deposited and progressively buried. Diagenesis concentrated the carbonate into layers and cementing the soils into tabular structures to varying degrees depending upon the extent of the diagenesis; 2. Post-depositional carbonate formation associated with the migration and intrusion of hydrocarbonrich fluids into the host soils. This brought about suitable geochemical conditions to preferentially precipitate carbonate within the host sediment and develop the chaotic and laterally discontinuous cemented features observed. These conditions are thought to be driven to some extent by the relatively dynamic hydraulic environment in the vicinity of the salt diapirs. The HIRs pose a number of uncertainties for piled foundations, both for installation and in-place performance. Where these features occur, the soils are either significantly altered through geochemical processes and/or affected by hydrocarbons, modifying their engineering behaviour. Guidance on what types of foundations can be installed where depends on the foundation type and size, and the relative depth and characteristics of the HIR at any given location. The guidance provided prior to the detailed investigation of the HIRs in the boreholes was: all foundation types should avoid seabed manifestations of HIRs; suction-installed caissons should avoid penetrating any subsurface HIR; and driven piles should avoid penetrating any subsurface HIRs unless there is sufficient confidence in the ground model at the proposed pile location. This may require more detailed interrogation of the geophysical data and/or acquisition of additional geotechnical information. An engineering study was designed to enable combinations of thickness, hardness and depth of HIRs

3.4

Among the geotechnical challenges was the possibility of seabed slope instability, and the potential impact this may have on development infrastructure. As discussed above, the seabed morphology in the development area has been strongly influenced by underlying salt movement. This has caused locally steep terrain and topographic highs and has also contributed to the generation of a steep-sided, narrow trough feature (Fig. 7). Further details of this feature are provided in Hill et al. (2010). While these underlying salt processes are considered static within a life-of-field timescale, the legacy of a locally steep terrain together with very soft seabed soils means that zones of the seabed may be susceptible to shallow instability. A study was therefore commissioned in order to identify routing and placement options within the

218 © 2011 by Taylor & Francis Group, LLC

Slope stability

Figure 7. Seabed renderings showing topographic features. Figure 6. Graphical representation of driveable combinations of HIR thickness and hardness, soil province 4 soil strength profile, Menck MHU 270T hammer, 84” pile.

development area which would help minimise the exposure of the development infrastructure to shallow slope failures. The narrow trough is of particular importance to the project, as pipelines running between two drill centres will be required either to detour around it, or to traverse it. The latter option represents a value opportunity in terms of reduced material and installation cost, but this must be balanced against the possible increased risk of damage from steep trough-side instability. The slope instability study was performed in two phases, and a description of the main study elements within each phase is presented below. In the first phase, a soil model was developed to map the variation of shallow soil strength over the proposed development area, based on a correlation between observed soil strength and interpreted seismic horizon depths. This allowed a spatially-resolute and continuous soil strength model, which avoids the mapped strength discontinuities otherwise associated with a purely soil province-based model. A GIS-based numerical modeling exercise was used to perform a site-wide slope stability analysis. The analysis was premised on infinite slope theory incorporating a shallow slab-type failure mode. Although this was deemed to be the most appropriate geomechanical model, the validity of this assumption was tested by performing comparative analyses based on a deeper rotational failure mode. The GIS-based infinite slope analysis was also informed by seabed slope derived from AUV high resolution bathymetry, as well as soil submerged unit weight and the influence of external loading from the on-bottom weight of various facilities. This first phase yielded a deterministic estimate of instability potential, in the form of estimated Factor

219 © 2011 by Taylor & Francis Group, LLC

of Safety (FoS), mapped over the study area. In order for the project to understand and manage the instability risk more fully it was decide to advance this deterministic assessment into a probabilistic stability assessment (PSA) to provide estimates of annual probability of slope failure. The first step in the PSA was to quantify the bias and uncertainty in the predictive soil strength model, as well as in the other stability calculation inputs. For soil strength, which is arguably the most influential parameter, this uncertainty was investigated by comparing model-predicted values with actual values measured from geotechnical borehole data or interpreted from seabed in situ tests. A statistical examination of predicted versus measured values at discrete locations gives a probabilistic strength distribution, which is essentially a measure of the ability of the regional soil model to predict the equivalent strength at any location had it been measured using conventional geotechnical methods (Fig. 8). For routine foundation design, the data in Figure 8 would be used to modify the probability density function of the predicted operational undrained shear strength to achieve the same level of design reliability that would be obtained using location-specific measured soils data (Gilbert et al. 1999). However, the data in Figure 8 were used directly in Monte Carlo simulations for initial slope stability screening purposes. The infinite slope equation was again used, but instead of using fixed, deterministic input values, they were chosen at random in each trial from the predefined measured strength (or other input parameter) distributions. The second phase also considered earthquake loading, the effect of which was modelled as an additional, quasi-static downslope load.The PSA allowed annual slope failure probability, under both gravitational-only and earthquake loading to be mapped over the study area (Fig. 9).

4

CONCLUSIONS

This paper has shown that a properly constructed ground model can be used to make informed decisions about the preferred location of foundations and how initially conservative assumptions can be changed with increasing resolution of the ground model. One of the most important applications of this model for offshore Angola was to inform facilities engineers of acceptable locations for installation of foundations and anchors in sediments affected by salt and hydrocarbon intrusion. The same ground model has also been used to decide on the routing of flowlines to lessen the impact of certain geohazards, such as slope stability. 5

Figure 8. Frequency distribution of ratio of measured (su(m)) to predicted (su(c)) shear strength.

CLOSING REMARKS AND ACKNOWLEDGMENTS

This paper represents the views of the authors and not necessarily those of the companies they represent. The authors wish to thank BP Angola, their development partners and Sonangol for permission to publish this material. They are also grateful for the critical reviews, guidance and stimulating conversation provided by Mike Sweeney and other members of the Exploration and Production Technology team at BP. REFERENCES

Figure 9. GIS map showing annual probability of slope failure.

A key outcome of the overall study is that, whereas the deterministic analysis suggested it would be difficult to traverse the narrow trough feature whilst maintaining a FoS magnitude considered as ‘high’ by conventional standards, the PSA suggests that more direct traversing routes exist where the seabed has a very low estimated probability of failure. Also included in the second phase was an assessment of the impact of landslide-related soil movement on installed pipelines and flowlines, and on the residual seabed topography (freespan potential). For a given slope failure probability, the engineering consequence in terms of induced pipeline stress and deformation could then be anticipated.

220 © 2011 by Taylor & Francis Group, LLC

Dendani, H., Colliat J.L., Puech, A. & Nauroy, JF 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences Keynote Paper, Proceedings 2ndInternational Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Evans, T.G. 2010. A Systematic Approach to Offshore Engineering for Multiple-Project Developments in Geohazardous Areas. Keynote paper, 2nd International Symposium Frontiers in Offshore Geotechnics, Perth, November 2010. Hill, A. J., Southgate, J.G, M., Fish, P. & Thomas, S. 2010. Deepwater Angola Part I: Geohazard Mitigation, 2nd International Symposium Frontiers in Offshore Geotechnics, Perth, November 2010. Gilbert, R., Gambino, S. & Dupin R. 1999. Reliability-based approach for foundation design without site-specific soil borings, Offshore Technology Conference, Houston, Texas, OTC 10927.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Shallow gas hazard linked to worldwide delta environments S. Kortekaas, E. Sens & B. Sarata Fugro Engineers B.V., Leidschendam, The Netherlands

ABSTRACT: Shallow gas hazard assessments are normally based on geophysical data, but when these are limited or unavailable, the geological setting can be used to predict whether shallow gas may be present. A GIS hazard map was created using geological settings, because they control local formation of biogenic gas, which is the most common gas (as defined by origin) in shallow sediments. The highest probability of shallow gas occurrence is in shallow marine settings with high terrestrial influx, deltas in particular. Therefore, a delta classification system based on shallow gas hazard was created and was verified using known gas occurrences. Together the hazard map and delta classification system can be an early screening tool for shallow gas hazard to provide input for risk assessments regarding open-hole or riser-less offshore drilling operations, even when no other relevant data are available.

1

INTRODUCTION

Shallow gas, which occurs at depths less than 1000 m below seafloor (Floodgate & Judd 1992), may pose a hazard to offshore open-hole or riser-less drilling operations, such as geotechnical drilling or drilling of the tophole section of oil and gas wells. For this reason, shallow gas hazard assessments are routinely performed prior to geotechnical drilling (Kortekaas & Peuchen 2008). Normally, hazard assessments are based on indications of gas accumulations in geophysical data. When limited or no geophysical data are available, the geological setting can be used to predict whether shallow gas is present. Information on the geological setting may also be instructive in a shallow gas hazard assessment because it can help the user assess the possibility that there is non-pressurised gas or gas in solution. Neither may be visible on seismic data, but both can pose a hazard to geotechnical drilling. Swabbing pressures produced by retrieval of a sample tool or drill string may cause exsolution of gas and induce gas flow or a gas kick (Kortekaas & Peuchen 2008). In this paper, relationships between depositional environment and the probability of shallow gas occurrence are discussed. This information was used to create a map of the probability of shallow gas accumulations based on depositional environment. The map could be used as an early screening tool when planning open-hole or riser-less drilling in these areas. For example, a user might be a project planner whose first act would be to locate the proposed drilling site on the map. If the proposed site lies within an area with a low probability of shallow gas hazard, the planner could relatively confidently proceed with drilling as planned with standard safety equipment and preventative procedures. However, if the site is in a zone of

intermediate or high probability of shallow gas, his/her second act would be to request further information and a detailed risk assessment.A detailed assessment could indicate whether drilling a pilot hole and other mitigation measures are sufficient (Kortekaas & Peuchen 2008), or the drilling site should be relocated.

2

There are two different types of shallow gas as defined by origin. The first is thermogenic gas, which forms at depth under high temperatures and pressures. It may be present in the shallow subsurface when it has migrated up from a deeper reservoir (Floodgate & Judd 1992). The second is biogenic gas, which forms at shallow depths through bacterial activity. Biogenic gas is by far the most common gas in shallow sediments (Lin et al. 2004). Thermogenic gas can migrate upward along natural pathways, through porous strata or along faults, or along leaking wells. Thermogenic gas may pose a hazard to drilling operations, especially when drilling close to existing wells. Nevertheless, it was excluded from this study because the focus was present-day depositional environments, which govern conditions of the formation of only local, biogenic gas. Biogenic gas formation requires a sufficient supply of organic matter and a rapid sedimentation rate to bury organic material before it is oxidised. The gas accumulates only when it can migrate in a free gas phase (Rice 1993). This occurs when the concentration in the pore fluid exceeds gas solubility, or when gas exsolves due to reduction of hydrostatic pressure, which could be caused by erosion of the seabed or a fall in relative sea level.

221 © 2011 by Taylor & Francis Group, LLC

DEPOSITIONAL ENVIRONMENT AND SHALLOW GAS

The preservation of shallow gas accumulations requires a reservoir, a seal and gas. In the shallow subsurface, reservoirs are most commonly formed by coarser-grained materials such as sand, and seals by fine-grained sediments such as clay. For simplicity, we refer to potential reservoir and seal materials as, respectively, sand and clay in this paper. Given the requirements described above, it can be concluded that shallow biogenic gas is most likely to occur in environments with sand and clay deposits, an influx of organic material and a rapid sedimentation rate. The combination of these factors is most probable on shallow ocean shelves, where a terrestrial influx of sand and organic material can be expected, and, in particular, in delta environments. 3

HAZARD MAP AND DELTA CLASSIFICATION

First, bathymetric data were loaded into a Geographic Information System (GIS). A map created from these data displays the initial, coarse hazard zonation by bathymetry. Most of the ocean was defined as deep sea, where either the water depth exceeds 1500 m, or the water depth is between 1000 m and 1500 m and the seabed slope is less than 1◦ . The continental slope and rise were grouped together for the purposes of the shallow gas hazard map. The slope and rise were defined as the area between 600 m and 1000 m water depth and the area between 150 m and 1500 m water depth where the slope exceeds 1◦ . The continental shelf is the zone between the continental slope and the shoreline. The water depth is generally 0 m to 150 m, but can reach 550 m. In this study, the shelf was defined as the area from the coastline to 150 m water depth and the zone between 150 m and 600 m water depth where the slope is less than 1◦ . Shallow shelf areas where terrestrial influx is high are more likely to contain sand, clay and gas than the deep sea. Given that these conditions are most common in deltas, information about 230 deltas was collected and placed in GIS. The information included: annual river discharge (Hovius 1998, Meybeck & Ragu 1995, Milliman et al. 1995, Orton & Reading 1993, Syvitski et al. 2005) and annual sediment load before damming (Meybeck & Ragu 1995, Milliman & Syvitski 1992); the area of the drainage basin (Beusen et al. 2005, Hovius 1998, Meybeck & Ragu 1995, Milliman et al. 1995, Syvitski et al. 2005); the net slope of the drainage area (calculated from topographic information); the climate in the delta and the climate in the drainage area (both inferred from global climate maps (The Times atlas of the world (1999)); and the presence of known oil and gas fields. As stated above, sand and clay deposits are the most likely combination of materials to provide a reservoir and seal for the accumulation and preservation of shallow gas. However, it is difficult to infer the ratios of sand and clay deposition in deltas from available

datasets. Existing delta classifications are not suitable to predict the presence of sand and clay deposits (e.g. Galloway 1975, Postma 1990, Orton & Reading 1993). Most delta deposits are likely to contain sand and clay, therefore, this analysis has focussed on the factors critical for gas formation: sedimentation rate and the supply of organic material to the delta. To assess the effect of these two factors on the probability of shallow gas presence in a given delta, two datasets were selected for the delta classification. 1. The annual sediment load (before human influence) was assumed to approximate the influx to the delta and to be the most important factor determining the sedimentation rate in the delta. A high sedimentation rate is critical for the formation of shallow gas accumulations because it facilitates burial of organic material before it can be oxidised. 2. The climate in the drainage area affects the amount of organic matter that is transported to the delta. A greater influx of organic matter to deltas of warmer, more humid drainage basins would be expected because these areas would be more biologically productive. Deltas with annual sediment loads greater than 0.1 million tons per year were examined, thus 199 deltas were included. The deltas were categorised by the present-day climate type dominant in their drainage basins (temperate humid, hot humid, cold humid and cold dry). Next, the deltas within each of the four climate types were classified by annual sediment load, also into four groups (Table 1). The deltas examined in this study are represented in Figure 1 by circles. Larger circles denote greater sediment loads. Please note that rather than create a fifth group with only one delta, the Ganges Delta was included in the group with the greatest sediment load though its annual load exceeds a trillion tons per year. To verify the classification, known occurrences of shallow gas from literature (Fleischer 2001, GarciaGarcia et al. 2007) and from Fugro experience were used and are represented by stars in Figure 1. Most published locations of shallow gas occurrences are known from seismic surveys, degassing seabed (observed visually and through acoustic surveys), gas escaping from samples, and from features such as pockmarks on the seabed. The sources (biogenic or thermogenic) Table 1. Number of deltas included in this study, divided by annual sediment load and dominant climate in their drainage basins. *For simplicity, the Ganges Delta is included in this group though the annual sediment load is greater than a trillion tons per year.

222 © 2011 by Taylor & Francis Group, LLC

Sediment load (106 ton/year)

Hot humid

Temperate humid

Cold humid

Cold dry

0.1 to 1 1 to 10 10 to 100 100 to 1000*

9 13 21 13

21 37 38 8

5 4 7 0

8 8 6 1

Figure 1. Locations of deltas examined in this study (circles) and reported shallow gas (stars). Circle size reflects relative sediment load.

of most of these shallow gas occurrences are known. Only biogenic gas occurrences were included in the verification of the delta classification. In contrast, the sources of most of the Fugro shallow gas occurrences are unknown. Therefore, it is possible that gas reported as biogenic gas in some deltas was, in fact, thermogenic gas that leaked up from deeper hydrocarbon reservoirs. Figure 2 shows the deltas subdivided by climate of their drainage basins and classified according to their annual sediment loads. For each sediment load class, the proportion of deltas in which shallow gas has been encountered is indicated. 4

RESULTS AND DISCUSSION

4.1 Delta classification Comparison of the results of the delta classification with known gas occurrences shows that shallow gas is most likely to occur in deltas with drainage areas in temperate humid and hot humid climates (Table 2). Greater biological production could be expected in such regions, and thus, by extension, more organic material transported by fluvial and overland drainage to these deltas. The probability of gas presence is generally higher in deltas of river systems with high annual sediment loads, which in general can be correlated to the large delta systems. A high annual sediment load correlates to a high sedimentation rate in the delta and thus a greater probability that organic material is buried and available for gas formation. Annual sediment load of a river is directly related to the size of the drainage basin: the larger the drainage area, the higher the sediment load (Milliman & Syvitski 1992). 4.2 GIS and the hazard map A GIS map created from selected data displays the initial, coarse hazard zonation by bathymetry, which

223 © 2011 by Taylor & Francis Group, LLC

Figure 2. The number of deltas with (black) and without (grey) reported shallow gas by annual sediment load, displayed by climate in the deltas’ drainage areas. Note that vertical scales differ.

essentially correlates with proximity to the coast and expected terrestrial input. A detail from the map is shown in Figure 3. The map includes deltas investigated in this study, annotated by sediment load and the climate in the drainage basin, both of which indicate the likelihood of shallow gas accumulation. Locations of known gas hydrates also appear on the map (Mazurenko & Soloviev 2003). Gas hydrates are not directly hazardous to geotechnical drilling; however, they may indicate the presence of free gas below the gas hydrate stability zone. This map can be used as a first screening tool of the shallow gas hazard in an area where drilling is proposed. 4.3

Although large amounts of organic material are not expected to be transported from cold dry drainage basins, gas was encountered in the large deltas with drainage basins in such climates. These may be older deltas, where in the past, climate conditions were different (i.e. warmer and/or more humid) and more favourable for shallow gas formation. It is important to note that palaeodeltas were not considered individually in this study. Nonetheless, as they are generally located on the shelf, their influence on shallow gas hazard probability is indirectly incorporated in the coarse hazard zonation by bathymetry. The probability of encountering shallow gas accumulations is also affected by wave and current energy. In high energy environments, sediment reworking prevents quick burial of organic material and deposition of clay, which is usually necessary to form seals. This has not yet been investigated in detail.

Data limitations

Known shallow gas occurrences (stars in Figure 1) appear to be concentrated in the northern hemisphere, especially around Europe and North America. The distribution is most likely more even than it appears. The apparent concentration in the northern hemisphere likely reflects the fact that more work has been done in these areas. Table 2.

Studied deltas by climate in drainage area.

Climate

Number of deltas

Deltas with reported shallow gas (%)

Hot humid Temperate humid Cold humid Cold dry All deltas

56 104 16 23 199

23 31.5 12.5 4.5 24.5

5

CONCLUSIONS

To generate and accumulate shallow gas in marine sediments, organic material, sand, clay and rapid deposition rates are usually necessary. All these are most commonly found in shallow seas (on the ocean shelves), and in delta environments, in particular. A classification of deltas by climate in the drainage area and annual sediment influx (river load before human influence) provides information about shallow gas hazards. The highest probability of shallow gas presence is in deltas with drainage basins in temperate humid and hot humid climates and in deltas, which

Figure 3. Part of the global GIS map, showing bathymetry zones, deep sea, continental slope/rise and continental shelf in different shades of blue. Climate of the delta drainage basin is indicated by the colour of the circle and the size corresponds to sediment load. Stars represent locations of reported shallow gas (black) and gas hydrates (yellow).

224 © 2011 by Taylor & Francis Group, LLC

receive high annual sediment loads. A GIS map of shallow gas hazard was created from the results of this study. This map can be used as a first screening tool for shallow gas hazard when planning geotechnical drilling operations, to provide input for risk assessments and to inform decisions regarding offshore drilling operations, even in areas for which no other relevant data are available. ACKNOWLEDGEMENTS The authors gratefully acknowledge Fugro’s commitment to improving safe and sustainable practice and supporting the work presented in this paper. Thanks are given to Poppe de Boer and Joek Peuchen for their valuable input. We also thank two anonymous reviewers for their comments, which helped to improve this paper. The opinions expressed in this paper are those of the authors and they are not necessarily shared by Fugro. REFERENCES Beusen, A.H.W., Dekkers, A.L.M., Bouwman, A.F., Ludwig, W. & Harrison, J. 2005. Estimation of global river transport of sediments and associated particulate C, N, and P. Global Biogeochemical Cycles 19(4): GB4S05.1GB4S05.17. Fleischer, P. Orsi, T.H., Richardson, M.D. & Anderson, A.L. 2001. Distribution of free gas in marine sediments: a global overview. Geo-Marine Letters 21(2): 103–122. Floodgate, G.D. & Judd, A.G. 1992. The origins of shallow gas. Continental Shelf Research 12(10): 1145–1156. Galloway, W.E. 1975. Process framework for describing the morphologic and stratigraphic evolution of deltaic depositional systems. In Broussard, M.L. (ed.), Deltas: models for exploration: 87–98. Houston: Houston Geological Society. García-García, A., Tesi, T., Orange, D., Lorenson, T., Miserocchi, S., Langone, L., Herbert, I. & Dougherty, J. 2007. Understanding shallow gas occurrences in the Gulf of Lions. Geo-Marine Letters 27(2-4): 143–154. Hovius, N. 1998. Controls on sediment supply by large rivers. In K.W. Shanley & P.J. McCabe (eds.), Relative role

of eustasy, climate, and tectonism in continental rocks: 3–16. SEPM Special publication 59. Tulsa: Society for Sedimentary Geology. Kortekaas, S. & Peuchen, J. 2008. Measured swabbing pressures and implications for shallow gas blow-out. In OTC.08; Proceedings 2008 Offshore Technology Conference, 5–8 May 2008. Houston, Texas, USA. OTC Paper 19280. Houston: Offshore Technology Conference. Lin, C.M., Gu, L.X., Li, G.Y., Zhao, Y.Y. & Jiang, W.S. 2004. Geology and formation mechanism of late Quaternary shallow biogenic gas reservoirs in the Hangzou Bay area, eastern China. AAPG Bulletin 88(5): 613–625. Meybeck, M. & Ragu,A. 1995. River discharge to the oceans: an assessment of suspended solids, major ions and nutrients. Environmental Information and Assessment Report. Nairobi: United Nations Environment Programme. Mazurenko, L.L. & Soloviev, V.A. 2003. Worldwide distribution of deep-water fluid venting and potential occurrences of gas hydrate accumulations. Geo-Marine Letters 23(3–4): 162–176. Milliman, J.D. & Syvitski, J.P.M. 1992. Geomorphic/tectonic control of sediment discharge to the ocean: the importance of small mountainous rivers. Journal of Geology 100(5): 525–544. Milliman, J.D. Rutkowski, C. & Meybeck, M. 1995. River discharge to the sea: a global river index (GLORI). LandOcean Interactions in the Coastal Zone Reports and Studies. Den Burg: LOICZ Core Project Office. Orton, G.J. & Reading, H.G. 1993. Variability of deltaic processes in terms of sediment supply, with particular emphasis on grain size. Sedimentology 40(3): 475–512. Postma, G. 1990. Depositional architecture and facies of river and fan deltas: a synthesis. In: A. Colella & D.B. Prior (eds.), Coarse-grained deltas: 13–27. Special Publications of the International Association of Sedimentologists 10. Oxford: Blackwell. Rice, D.D. 1993. Biogenic gas: controls, habitats, and resource potential. In D.G. Howell (ed.), The Future of Energy Gases: 583–606. U.S. Geological Survey Professional Paper 1570. Washington: United Sates Government Printing Office. Syvitski, J.P.M., Kettner,A. J., Correggiari,A. & Nelson, B.W. 2005. Distributary channels and their impact on sediment dispersal. Marine Geology 222–223: 75–94. The Times atlas of the world. 1999. Tenth Comprehensive Edition. Crown Publishers, New York.

225 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Analysis of submarine flow slides in fine silty sand P.V. Lade The Catholic University of America, Washington, D.C., U.S.A.

J.A. Yamamuro Oregon State University, Corvallis, OR, U.S.A.

ABSTRACT: The mechanism of instability in granular soils is briefly explained, and its requirement as forerunner for liquefaction of level or sloping ground is described. Tests on loose silty sand indicate a ‘reverse’ behavior with respect to confining pressure and as opposed to the behavior of loose, clean sands. Strong correlations between fines content, compressibility and liquefaction potential are found for these soils. A procedure for analysis and evaluation of static liquefaction of slopes of fine sand and silt such as submarine slopes is presented. It involves determination of the region of instability in stress space in which potential liquefaction may be initiated and determination of the state of stress in the slope. The instability line and the region in which potential liquefaction may be initiated can be determined from consolidated-undrained triaxial compression tests, and a method of finding the state of stress is developed to predict the zone of potential liquefaction in simple slopes. 1

INTRODUCTION

Recent experiments involving loosely deposited silty sands dispute the assumption that clean sands always behave similar to silty sands. The tests on loose silty sand indicate a ‘reverse’ behavior with respect to confining pressure and this violates the basic assumption that loose, silty sands behave similar to loose, clean sands. There is a strong correlation between fines content, compressibility and liquefaction potential of these soils. A procedure for analysis and evaluation of static liquefaction of slopes of fine sand and silt such as submarine slopes, mine tailings, and spoil heaps is presented. This procedure involves determination of the region of instability in stress space in which potential liquefaction may be initiated for the soil in question and determination of the state of stress in the slope. It is explained how the instability line and the region in which potential liquefaction may be initiated can be determined from consolidated-undrained triaxial compression tests, and a straightforward method of finding the state of stress is employed to predict the region of potential liquefaction in simple slopes.

2

REVERSE BEHAVIOR OF SILTY SAND AT LOW CONFINING PRESSURES

Figure 1. Drained triaxial tests on 12% relative density silty Nevada sand at confining pressures from 26 to 600 kPa.

Within a range of low confining pressures, drained triaxial tests on very loose silty sand with high compressibility show negligible effect of magnitude of confining pressure on the contractive volume change, as shown in Figure 1 (Yamamuro and Lade 1997). The corresponding undrained tests in the same range of

confining pressures show development of essentially equal pore pressures, as indicated in Figure 2, and consequently the effective confining pressures reach zero faster with decreasing initial consolidation pressures. Thus, the lower the initial consolidation pressure

© 2011 by Taylor & Francis Group, LLC

227

Figure 2. Effective confining pressures during shearing from tests on 12% relative density silty Nevada sand.

Figure 4. Effects of fines content and void ratio on volume compressibility and static liquefaction in experiments on silty Nevada sand.

temporary liquefaction. This second region is characterized by an initial peak stress difference, followed by a decline. As shearing continues the stress path crosses the phase transformation line into the region of dilation and pore pressure decline, resulting in stress differences increasing to much higher magnitudes than the initial peak. In this region the specimens show increasing dilatancy with increasing initial consolidation pressure, contrary to conventional sand behavior. The following two regions of temporary instability and instability are those recognized from conventional sand behavior. Figure 3. Four distinctly different general types of undrained effective stress paths for loose silty sands: static liquefaction, temporary liquefaction, temporary instability, and instability shown in p’-q diagram.

the faster liquefaction conditions are reached in the specimens. This clearly shows that static liquefaction is a low-pressure phenomenon. The effect of increasing the confining pressure is to increase the resistance to liquefaction. This behavior is contrary to observed behavior for conventional undrained tests on clean sands (Seed & Lee 1967). This ‘reverse’behavior observed for very loose silty sand at low confining pressures is accompanied by an inflection in the instability line, as seen (exaggerated) in the schematic diagram in Fig. 3 (Lade and Yamamuro 1997). Four distinctly different types of effective stress paths with corresponding behavior patterns are shown. Static liquefaction occurs at the lowest pressures, and it is characterized by large pore pressure developments that result in zero effective confining pressure and zero stress difference at low axial strains. In this range, the maximum effective friction angle increases with increasing effective confining pressure and it continues to increase through the following region of © 2011 by Taylor & Francis Group, LLC

3

COMPRESSIBILITY AS A MEASURE OF LIQUEFACTION POTENTIAL

This pattern of ‘reverse’ sand behavior is entirely controlled by the very high compressibility of the very loose silty sand. The compressibility is in turn controlled by the amount of fines present in the sand. Figure 4 shows a correlation between the fines content, void ratio, volume compressibility and static liquefaction observed in experiments on Nevada sand (Yamamuro and Lade 1998). The “wall” between the region of stable behavior and the liquefaction regime corresponds to almost constant compressibility. Thus, if the compressibility is higher than this wall, i.e. in the approximate range from (1.4–2.2) · 10−5 (1/kPa), then liquefaction can occur under undrained conditions. Similar results were obtained by Lade et al. (2009) from experiments on Ottawa sand mixed with Loch Raven silt. For this silty sand liquefaction may occur under undrained conditions for compressibilities higher than (1.2–1.6) · 10−5 (1/kPa), which are very similar to the compressibilities for Nevada sand. The development of pore pressures under undrained conditions is directly related to the compressibility of the soil, and loose silty sands exhibit significant volumetric contraction at low pressures. Yamamuro and

228

Lade (1998) proposed to use volumetric compressibility as an alternative indicator of liquefaction potential for silty sands. Volumetric compressibility does not require determination of void ratio and fines content, and it can be measured in many different ways. For example, the last increment in isotropic compression before undrained shearing in a triaxial compression test may be used to obtain the volumetric compressibility. It may also be obtained from a standard oedometer test at the appropriate low stress magnitude, where the liquefaction potential is highest. Furthermore, it may be obtained from an in-situ test on the intact soil by inserting a screw-plate to relatively shallow depth and performing a plate load test to determine the vertical compressibility at the relevant field location. Such an in-situ test also captures the very important effect of the soil fabric or structure (Wood et al. 2008; Yamamuro et al 2008) and avoids the difficult to impossible task of having to recover intact samples of the loose, silty sand for testing in the laboratory. Alternatively, a pressuremeter test may be employed to determine the compressibility in the horizontal direction. Whether the vertical or the horizontal compressibility is more relevant for indication of liquefaction potential remains to be seen. In addition, the actual boundary conditions in the field are important for determining the liquefaction potential in the field. From knowledge of constitutive modeling of soils, it is logic that the volumetric compressibility is one of the significant factors that control the development of pore pressures under undrained conditions. The fact that this property of a loose, silty sand deposit may be determined in-situ by a screw-plate test (vertical compressibility) or by a pressuremeter test (horizontal compressibility) at relatively shallow depths may make determination of liquefaction potential relatively easy. Besides, the fact that all significant factors that influence the volumetric compressibility are already present in the field deposit further increases the importance of such in-situ tests in indicating liquefaction potential.

4 ANALYSIS PROCEDURE FOR STATIC LIQUEFACTION To analyze a slope for its potential for static liquefaction, the states of stress everywhere in the slope are compared with the states of stress in the region of potential liquefaction. According to the recently discovered ‘reverse’ behavior, the critical region reaches all the way down to the stress origin, where true liquefaction occurs. In the following step of the instability method, the state of stress in the slope is superimposed on the stress diagram to find out if any stress states overlap with the region of potential instability. In such overlapping regions, any point is a point of potential instability, and instability will develop under undrained conditions if a suitable trigger mechanism is activated to initiate the © 2011 by Taylor & Francis Group, LLC

Figure 5. Location of region of potential instability and subsequent liquefaction in p’-q diagram.

instability. If the region of potential instability reaches down to the stress origin, then the ground may liquefy.

5

EXPERIMENTS TO DETERMINE THE LIQUEFACTION REGION IN STRESS SPACE

The region of potential instability is located between the instability line and the failure line, and it reaches down to the stress origin for liquefaction to occur, as indicated schematically in Figure 5. The instability line connects the tops of the yield surfaces, and it is a straight line through the stress origin (Lade 1992). The effective stress paths from undrained tests on very loose compressible soil essentially trace the yield surfaces, and determination of the top of one such yield surface is, in principle, sufficient to define the instability line. Thus, the instability line may be characterized by its inclination, similar to a friction angle. However, it should be understood that this inclination cannot be used in a manner similar to a conventional friction angle. For analysis of liquefaction potential, it is necessary to determine the peak shear stress, Sp for which liquefaction can occur, i.e. for which the effective stress path reaches down to the stress origin where true liquefaction occurs, in order to locate the triangular region of potential liquefaction, as indicated in Figure 5. The key difficulty in performance of the undrained tests consists of depositing a silty sand in the laboratory with a structure similar that in the field. Many methods of soil deposition are available for laboratory creation of loose sand deposits, and their effects on the behavior have been investigated in detail (e.g. Ishihara 1993; Vaid et al. 1999, Wood et al 2008; Yamamuro et al. 2008). Each method of deposition produces its own sand fabric and behavior, and these behaviors are very different for different deposition methods. However, it is not known at this time which method is most likely to produce a sand fabric similar to that in the field, simply because the fabric in field deposits of fine silty sand is not known and has not been sufficiently characterized to allow reproduction by an artificial laboratory method. While the difficulty lies in obtaining undisturbed samples of field deposits

229

instability, the addition of the shear stresses may cause the gently inclined slope to have stress states that, on some planes, will reach into the region of potential instability. To determine the required slope inclination for the Mohr circle to become tangent to the instability line, as shown in Figure 6(a), an expression may be developed on the basis of (1) the expressions for the principal stresses from the Mohr circle shown in Figure 6(a), (2) the Mohr-Coulomb failure criterion for sand used with the instability line inclined at φi , (3) the value of K0 = 1-sinφ. The resulting expression for the slope inclination, α, becomes:

Figure 6. (a) Mohr diagram with indication of shear and normal stresses limiting the region of potential instability, and (b) Volume of soil for calculation of stresses along plane parallel with sloping surface.

Thus, for slopes with inclinations smaller than α, the Mohr circle will not reach into the region of potential instability and such slopes will not become unstable under static conditions. For slopes with inclinations greater than α, many planes in the slope will be located in the region of potential instability. For such conditions the zone of potential instability in the submarine slope is also limited by the peak shear stress, Sp , indicated in Figure 5. Thus, for locations with shear stresses less than Sp in the slope, the state of stress is in the region of potential instability. The analysis required to determine this region is simply an infinite slope stability analysis, indicated in Figure 6(b), from which it may be determined that:

to study in the laboratory, such intact samples have been obtained by the freezing method and tested in the laboratory (e.g. Yoshimi et al. 1984, 1989; Vaid et al. 1999). Vaid et al. (1999) recommend using the water pluviation method, because they found that it produces results that are comparable with those on in-situ frozen undisturbed sand specimens from three different sites.

6

ZONE OF POTENTIAL LIQUEFACTION IN SLOPING GROUND

The states of stress in the region of potential liquefaction, shown in Figure 5, can be identified so as to outline a zone in sloping ground, such as in a submarine slope, within which liquefaction may be triggered. Beginning with a slope with moderate inclination, α, the state of stress in the slope may be obtained as shown in Figure 6(a). The state of stress in level ground is obtained from the value of K0 . Thus, σ h ’= K0 ·σ v ’, and the corresponding Mohr’s circle is shown in Figure 6(a). For a gently sloping ground surface, the normal stresses on vertical and horizontal planes may be approximated by those obtained from the K0 stress state. The shear stresses acting on these planes may be obtained from τ = σ v ’·tanα. The corresponding states of stress are represented by the larger Mohr circle in Figure 6(a). This circle represents a reasonable approximation to the real states of stress in the gently inclined slope. While the Mohr circle for the K0 stress state may not reach into the region of potential © 2011 by Taylor & Francis Group, LLC

in which h is the vertical depth below the sloping surface, b is the length considered along the sloping surface, and γ b is the buoyant unit weight of the silty sand. The Mohr diagram in Figure 6(a) shows that the peak shear stress, Sp = (σ 1 −σ 3 )/2, may be expressed in terms of τ in Equation (4):

The vertical depth, hi , to which the zone of potential instability reaches down from the sloping surface may then be determined from the expression in Equation (5):

Thus, the zone in the submarine slope in which instability may be initiated reaches from the surface and vertically down to a depth of hi . Once the

230

Note that the resulting zone of potential liquefaction reaches all the way up to the sloping surface and is not bounded by a zone of dilation and stable behavior, as was the zone of instability previously determined (Lade 1992). The reason is the difference between the ‘reverse’ behavior of loose, silty sand and the normal behavior observed for undrained tests on clean sand, as described above. Figure 7. Submarine slope of silty sand with indication of zone of potential instability and subsequent liquefaction.

7

slope inclination increases above α as determined from Equation (1), then the value of hi determines the extent of the zone below the surface of the slope, as indicated in Figure 7. It should be noted that the maximum possible value of the slope inclination is α = ϕ, at which the slope is barely stable against conventional slope stability, which involves comparisons of shear stresses and shear strengths. Therefore, as long as the soil remains drained, it will remain stable. But a small disturbance will cause the silty sand, which has relatively low permeability, to react in an undrained manner and will subsequently cause the slope to become unstable under essentially static conditions. The small disturbance may cause any point within the zone of potential instability, shown in Figure 7, to respond in an undrained manner and trigger the instability and subsequent liquefaction of the slope. A few example calculations are performed for a submarine slope consisting of Nevada sand with 6% fines and a relative density of 12% (Yamamuro and Lade 1997). The friction angle for this silty sand is 33◦ , the instability angle for the isotropically consolidated silty sand is 17◦ , the maximum peak shear stress for a specimen that liquefies is Sp = qmax /2 = 50/2 = 25 kPa, the void ratio after consolidation is approximately e = 0.8, and the buoyant unit weight is therefore γb = (Gs -1)·γ w /(1 + e) = 0.92 g/cm3 . The value of K0 = 1-sinφ = 0.455, which corresponds to a mobilized friction angle of 22◦ . Since this value is greater than the angle of instability ϕi = 17◦ , the sand is potentially unstable under a level ground surface, and all that is required for this soil to become unstable and subsequently liquefy is a trigger to initiate the instability. However, as explained in the Discussion section below, the instability line will be steeper for a K0 consolidated soil than for an isotropically consolidated soil, and in practice it will always be located with greater slope than the K0 -line. Thus, assume for the sake of demonstration that the appropriate instability line is inclined at ϕi = 25◦ . The minimum slope inclination required for potential instability is calculated from Equation (1) to be α = 8.1◦ . At that slope angle the vertical depth of the zone of potential instability may be calculated from Equation (6) to be 17.6 m. As the slope inclination increases above the value of α = 8.1◦ , the instability zone in the slope becomes less deep and at α = ϕ = 33◦ , the vertical depth is hi = 5.4 m. © 2011 by Taylor & Francis Group, LLC

DISCUSSION

The zone of potential instability and subsequent liquefaction comprises a volume of soil parallel with the sloping surface, indicated by the shaded zone in Figure 7, in which any point is on the verge of unstable behavior, and once the instability has been initiated, this is the zone that liquefies first. However, once liquefaction has been initiated, the rapidly increasing pore pressures will cause the pore water to move towards locations with lower pressures, i.e. out of the slope and deeper into the slope. Here the increasing pore pressures will overcome the tendency of the silty sand at greater depth to dilate (according to the ‘reverse’ behavior pattern) and render the deeper dilating soil unstable as well (Lade et al. 1993). Thus, the unstable zone will reach into the slope and tend to create what may look like a concave cavity, as indicated in Figure 7. The difference between conventional slope failure and slope liquefaction is that the conventional slope failure results in a slumping surface in which the soil moves down slope by some small distance, while slope liquefaction produces a scarp with no remaining soil, because the liquefied soil runs far away as a liquid, as observed in the Nerlerk berm failures (Lade 1993).

8

CONCLUSION

A summary of the ‘reverse’ behavior discovered for loose deposits of fine silty sands at low confining pressures is presented. The high compressibility of the loose, silty sands plays an important role in their liquefaction potentials, and it is proposed that this may be identified from in-situ measurements by screw plate or pressuremeter tests on the intact soil in the field. This has the advantage of producing test results for intact soil with the in-situ particle fabric, and it avoids the problems surrounding the recovery of intact samples of soils that are notoriously difficult to sample. Laboratory testing procedures for appropriately deposited specimens to determine the liquefaction region in stress space are reviewed, and an analysis procedure is developed for potential instability and liquefaction of sloping ground resulting in submarine flow slides.

231

REFERENCES Ishihara, K. 1993. Liquefaction and flow failure during earthquakes. Geotechnique, 43(3): 351–415.

Lade, P.V. 1992. Static instability and liquefaction of loose fine sandy slopes. Journal of Geotechnical Engineering, ASCE, 118(1): 51–71. Lade, P.V. 1993 “Initiation of static instability in the submarine Nerlerk berm”, Canadian Geotechnical Journal, 30(5): 895–904. Lade, P.V., Bopp, P.A. & Peters, J.F. 1993. Instability of dilating sand. Mechanics of Materials, Elsevier, 16: 249–264. Lade, P.V. & Yamamuro, J.A. 1997. Effects of non-plastic fines on static liquefaction of sands.Canadian Geotechnical Journal, 34(6): 918–928. Lade, P.V.,Yamamuro, J.A. & Liggio, C.D., Jr. 2009. Effects of fines content on void ratio, compressibility, and static liquefaction of silty sand, Geomechanics and Engineering, Techno-Press, 1(1): 1–15. Seed, H.B. & Lee, K.L. 1967. Undrained strength characteristics of Cohesionless soils. Journal of the Soil Mechanics and Foundations Division, ASCE, 93(SM6): 333–360. Vaid, Y.P., Sivathayalan, S. & Stedman, D. 1999. Influence of specimen-reconstituting method on the undrained response of sand. Geotechnical Testing Journal, ASTM, 22(3): 187–195.

© 2011 by Taylor & Francis Group, LLC

Wood, F.M., Yamamuro, J.A. & Lade, P.V. 2008. Effect of depositional method on the undrained response of silty sand. Canadian Geotechnical Journal, 45(11): 1525–1537. Yamamuro, J.A. & Lade, P.V. 1997. Static liquefaction of very loose sands. Canadian Geotechnical Journal, 34(6): 905–917. Yamamuro, J.A. & Lade, P.V. 1998. Steady state concepts and static liquefaction of silty sands. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 124(9): 868–877. Yamamuro, J.A., Wood, F.M. & Lade, P.V. 2008. Effect of depositional method on the microstructure of silty sand. Canadian Geotechnical Journal, 45(11): 1538–1555. Yoshimi,Y., Tokimatsu, K. Kanoko, O. & Makihara,Y. (1984). Undrained cyclic shear strength of a Niigata sand. Soils and Foundations, 24(4): 131–145. Yoshimi, Y., Tolimatsu, K. & Hosaka, Y. (1989). Evaluation of liquefaction resistance of clean sands based on high-quality undisturbed samples. Soils and Foundations, 29(1): 93–104.

232

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Hydrate dissociation around oil exploration infrastructure A.K. Sultaniya, J.A. Priest & C.R.I. Clayton University of Southampton, United Kingdom

ABSTRACT: The widespread occurrence of gas hydrates in marine sediments has raised concerns of potential risks to oil production infrastructure, should the hydrate dissociate when pumping hot oil from deeper strata through riser pipes. Hydrate dissociation changes the ice-like structure of hydrate back to its constituent components of gas and water, possibly increasing pore pressure, reducing effective stress, and altering the stiffness and strength of the sediment. To understand sediment behaviour during hydrate dissociation well-controlled laboratory tests on hydrate-bearing sands were conducted at the University of Southampton to mimic temperature in sediment around an oil riser pipe. Measurements were conducted throughout the dissociation stages to determine effective stress and sediment stiffness changes, as well as the associated inherent damping, during undrained and drained conditions. The results will be subsequently used within a FE model to assess the performance and stability of oil exploration infrastructure during hydrate dissociation.

1

INTRODUCTION

Gas hydrates are naturally occurring metastable compounds composed of gaseous molecules encapsulated within a water matrix to form an ice-like structure. The most common gas found within gas hydrates is methane, although gases like CO2 , H2 S, ethane etc are also found in naturally occurring gas hydrates. Gas hydrates exist where there is an ample supply of gas within the sediment, combined with high pressure and/or low temperature conditions. In nature these conditions exist within oceanic sediments on continental margins and deep within sediments in arctic regions below the permafrost. As gas hydrates are metastable they dissociate if temperature or pressure conditions are sufficiently altered. This can change gas hydrate from an ice-like structure back to its constituent parts of gas and water. This will lead to changes in the physical properties of hydrate-bearing sediments. In forming gas hydrates the gas is able to achieve a denser packing (1 m3 of methane gas hydrate contains 164 m3 of methane at Standard Temperature and Pressure (Sloan, 1998)) than it could occupy in its gaseous state, so dissociation may result in an increase in pore pressure in the sediment. Oil and gas exploration activities have begun to extend to significant water depths (greater than 1000 m) where gas hydrates are known to exist. Pumping of hot oil or gas through the hydrate-bearing sediments may dissociate hydrate, resulting in heave or subsidence around oil/gas wells, depending upon the permeability of the adjacent layers. This can lead to casing failure or at the extreme platform subsidence.

© 2011 by Taylor & Francis Group, LLC

This phenomenon has been observed in oil drilling activities by many researchers (Makogon, 1981; Nimblett et al., 2005; Tan et al., 2005). Hydrate dissociation, induced through changes in sea level or rise in ocean bottom temperature, has been associated with large seafloor failure in the geologic past causing devastating tsunamis and widespread flooding (Carpenter, 1981; Kayen & Lee, 1991; Padden et al., 2001). Numerical modeling of sediment behaviour can be used to asses the impact of drilling activities on sediment behaviour. Understanding and mitigating the risk of drilling through hydrate-bearing sediments, using numerical models, can only be effective if detailed knowledge of sediment behaviour during dissociation is available. At present the influence of hydrate dissociation on sediment behaviour is not well understood. Gas hydrate dissociation characteristics can be derived either through field testing, laboratory testing on recovered or testing laboratory prepared gas hydrate samples. However, since gas hydrate exists in deep-water oceanic sediments, or deep permafrost sediments, it is often impractical to perform field testing and difficult to obtain undisturbed in-situ samples for testing (Priest et al., 2008). This paper reports on a series of tests conducted on methane hydratebearing sands to determine the physical properties of these sands during formation and dissociation of gas hydrate. Physical properties included measurements of the small strain stiffness as well as the respective damping ratios under low frequency conditions relevant to seismic geological testing. Factors such as the stress conditions during formation and dissociation were also investigated

233

2

LABORATORY APPARATUS

There are a number of different laboratory techniques that have been utilized to determine the physical properties of soils. One such apparatus that is routinely used is the resonant column, which allows estimates of the small strain stiffness of soils and its respective material damping. Generally resonant column testing is performed under low stress conditions (1 MPa) of pressure and at room temperature. But, methane hydrate exists under restricted thermobaric conditions and so pressures (up to 20 MPa) and low temperatures (down to −20◦ C) are required to allow effective formation of hydrate in the laboratory (Stern et al., 1996). To that end the Gas Hydrate Resonant Column (GHRC) apparatus was developed at the University of Southampton (Priest, 2004). The drive mechanism and operating principles are based on a ‘Stokoe’ resonant column, and was modified to apply flexural vibration to the sand (to derive longitudinal wave velocity) in addition to torsional vibration (shear wave) at small strain (<10-6). For a fuller description of the apparatus refer to (Clayton et al., 2005). In a standard Stokoe’ resonant column, a cylindrical specimen is fixed on a base pedestal with the other end connected to the drive mechanism via a top cap (Fig 1). Torsional vibrations are induced by applying a sinusoidal voltage to the coils to create a magnetic field which attracts the magnets attached to the drive mechanism. Flexural vibration is induced by only applying voltage to two opposing coil sets (Cascante et al., 1998). The magnitude of vibration is measured by an accelerometer attached to the drive plate. Figure 2 shows the response of a 140 mm high by 70 mm diameter cylindrical specimen of Leighton Buzzard sand during a frequency sweep. The resonant frequency of the system (specimen and drive mechanism) identified from the maximum voltage amplitude, can easily be identified from the response curve. From this resonant frequency the shear modulus of the specimen, Gmax can be calculated. The same technique is used to derive the flexural Young’s modulus, Eflex during flexural vibrations. Material damping for each vibration mode is calculated using the free vibration decay method by turning off the power to the drive coils when the system is vibrating at its resonant frequency. Using an open circuit during free vibration decay prevents back e.m.f. being generated in the coils (Wang et al., 2003) which can lead to an overestimate of damping. 3

SAMPLE PREPARATION

All soil specimens used in the testing were prepared using Leighton Buzzard sand (Grade-E). This is uniform sub-angular sand with a nominal grain size between 90–150 µm (85% by weight). Its specific gravity was calculated at 2.65. Minimum and maximum densities were 1331 kg/m3 and 1624 kg/m3 respectively. Minimum and maximum void ratio obtained for this sand was 0.633 and 0.933, respectively (Priest et al., 2005).

© 2011 by Taylor & Francis Group, LLC

234

Figure 1. General layout of a ‘Stokoe’ resonant column apparatus. Flexural vibration is also shown.

Figure 2. The dynamic response of a Leighton Buzzard sand with 10% moisture content during frequency sweep.

Gas hydrates within the sand specimens were formed using the ‘excess gas’ method was utilized (Priest et al., 2009). A known mass of de-aired water is added to a known mass of air-dried sand. During formation of hydrate all free water is converted into hydrate from which the volume of hydrate can be quantified. Specimens were formed using the moist tamping method into a mold in 8 equal layers to form a dense specimen. The void ratio of the sample was calculated from the mass of sand and dimensions of the specimen; from these values, and the mass of water added, the hydrate saturation of the pore space can be calculated. Once the specimen is formed and sealed in a butyl membrane, a vacuum of 25 kPa was applied to allow the mold to be removed and the various apparatus components (drive mechanism, thermistors and axial transducers) to be attached. Once complete a cell pressure of 250 kPa was applied to the specimen (point A, Figure 3). Methane gas back pressure was then applied to the specimen and raised to 6 MPa whilst maintaining 250 kPa effective stress (point B). Once at the desired pressures the specimen was allowed to rest until the pore pressure and axial displacement readings stabilized. The temperature of the GHRC was then lowered

Table 1. Physical properties of laboratory prepared specimen used in the testing. Sample Properties

10H-DD

10H-DU

Hydrate saturation* (% of pore space) Moisture content (%) Void ratio Dry density (kg/m3 ) Effective stress during hydrate formation (kPa)

10.63

10.32

2.98 0.756 1509

2.93 0.775 1493

250

2000

*calculated from added water

Figure 3. Different testing stages represented with arrow in methane hydrate phase boundary curve.

to 1◦ C (point C), to induce hydrate formation. Temperature and pressure were maintained for at least 48 hours to allow complete hydrate formation; this is assessed by monitoring changes in stiffness from resonant column measurements undertaken. After formation the hydrate was dissociated by raising the temperature (point D) to investigate dissociation characteristics of the sediment. Two different tests were undertaken with a target hydrate saturation of 10% of the pore space. In one test hydrate was formed whilst under an effective stress of 250 kPa and dissociated under the same effective stress by increasing cell pressure (to compensate for increase in pore pressure due to dissociation). This can be idealized as a drained case where the pore pressures are able to dissipate and overall effective stress remains unchanged. In the second test, hydrate was formed under 2000 kPa effective stress and dissociated under a similar initial starting effective stress, however in this test the cell pressure was not raised and so effective stress reduced with dissociation, therefore being idealized as an undrained test.

Figure 4. Change in stiffness with time during hydrate formation. Full hydrate formation assumed when stiffness does not change with time.

All RC tests on the specimen were conducted at strain levels below the elastic threshold, (γ <10−5 ), defined as the point when the calculated modulus (G, Eflex ) is independent of strain (Saxena et al., 1998). Material damping was calculated from the free vibration decay curve obtained at each resonant frequency during the testing. 5

RESULTS AND DISCUSSION

4 TESTING PROCEDURE Resonant column (RC) tests were conducted using both torsional and flexural vibration. Small strain resonant column tests were conducted at regular intervals during hydrate formation and dissociation stages. Full hydrate formation was assumed to have occurred when no change in specimen stiffness was observed; for both tests this occurred ∼44 hrs after specimen was lowered into the hydrate stability field. After full hydrate formation, resonant column tests were conducted during loading and unloading of the specimen. This was achieved by raising the effective stress on the specimen from 250 kPa to 2000kPa in 250 kPa steps and then reducing effective stress back down to its initial value. Each increment was maintained until axial displacements, as measured by the transducer, were negligible (∼30 minutes).

© 2011 by Taylor & Francis Group, LLC

As discussed two specimens of dense sand containing ∼10% water within their pore space were subjected to hydrate formation and dissociation whilst at the same time undergoing resonant column testing to determine the small strain shear modulus, Gmax and small strain flexural Young’s modulus, Eflex of the specimens and their respective damping ratios. Each specimen was subjected to different initial effective stresses during the formation and dissociation stages. Initial properties of the soil specimens tested are presented in table 1. Figure 4 shows the calculated Gmax and Eflex values derived from RC tests during gas hydrate formation, along with the specimen temperature during this stage. It can be seen that both Gmax and Eflex increase rapidly during the early stages of hydrate formation due to the cementing effects of the gas hydrate. The temperature

235

Figure 7. Relationship between temperature and small strain modulus of the specimen with time during dissociation.

Figure 5. Relationship between Gmax with isotropic confining pressure, σ
.

Figure 6. Relationship between Eflex with isotropic confining pressure, σ
.

data shows a spike in specimen temperature during the initial stages of formation, due to the exothermic nature of hydrate formation. The top of the temperature spike (∼3 hrs) corresponds to ∼32% hydrate formation. As hydrate forms at the gas/water interface a rind of hydrate is formed encapsulating the water with further hydrate growth by diffusion of gas through this rind. This reduces the rate of formation of hydrate, and so the heat liberated is insufficient to counteract the loss of heat caused by lowering the system temperature. When 50% of the hydrate is formed (equivalent to 5% hydrate saturation) an increase of 30% in specimen stiffness is observed. This is somewhat less than the total increase in stiffness for a specimen with 5% hydrate saturation observed by Clayton et al. (2005), which had a Gmax and Eflex of 1.5GPa and 2.4GPa respectively. This difference in stiffness is due to how the hydrate interacts with the sand grains and is discussed later. Figure 5 & 6 shows the calculated values of Gmax and Eflex as a function of isotropic effective confining pressure, σ
for the two specimens listed in Table 1. Also included is data for loose and dense sand specimens for the same Leighton Buzzard sand with 0% hydrate saturation (Priest et al., 2005). The results show that the inclusion of the hydrate within the pore © 2011 by Taylor & Francis Group, LLC

space increases the stiffness by ten fold, due to the cementing effect of gas hydrate at the grain contacts (Waite et al., 2004; Priest et al., 2005), when compared to the non-hydrate bearing dense sand. Figure 5 & 6 also shows that the stiffness of hydrate bearing sands is independent of the effective stress applied during the formation. This is unsurprising given the magnitude of stiffness increase caused by adding hydrate to the pore space compared to the changes in stiffness due to grain compliance for the non-hydrate bearing specimens as a function of effective stress. As was seen in Figure 4, gas hydrate formation is an exothermic reaction; however during dissociation the reaction is endothermic. Thus as the temperature of the system is raised to initiate hydrate dissociation a change in gradient for the specimen temperature is observed at ∼8.9◦ C, as can be seen in Figure 7. Although this corresponds to the location on the hydrate stability curve for a gas pressure of 6 MPa (see Figure 3), minor volumes of hydrate had already dissociated (2.9%). This small volume of hydrate dissociation may result from a lag temperature between the centre of the specimen where the thermistor is located and the outside of the specimen. The volume of hydrate dissociated is calculated from the rise in pore pressure during dissociation using the Peng-Robinson equation. The rise in pore pressure is related to the volume of methane gas released by the hydrate and can be calculated knowing the volume of the pore space in the specimen. The true volume of voids is corrected to account for the changing fraction of hydrate and water that occurs. The parameters used in the Peng-Robinson equation are given in Table 2. Figure 7 also shows Gmax and Eflex calculated during hydrate dissociation. It can be seen that a sharp reduction in specimen stiffness occurs for a small reduction in hydrate saturation, with specimen stiffness reducing by up to 80% for only a 15% change in hydrate (equivalent to a reduction from 10% to 8.5% total hydrate saturation). This can be more clearly seen in Figure 8, which shows the change in Gmax and Eflex with respect to the volume of hydrate dissociated. It can be seen that for ∼5% reduction in hydrate saturation stiffness

236

Table 2. Physical properties of methane and methane hydrate are used for the calculations. Critical temperature of methane (◦ K) Critical pressure of methane (MPa) Accentic factor of methane Density of methane hydrate (kg/m3 ) Methane mole fraction dissolved in water

190.6 4.656 0.0108 917 0.0012

Figure 10. Relationship between effective stress and pore pressure with temperature of the specimen during dissociation.

Figure 8. Variation of small strain modulus of the specimen with percentage of hydrate dissociation.

Figure 9. Idealized hydrate formation and dissociation at grain contacts. (a) Hydrate formation, (b) Hydrate dissociation.

reduces by about 50% while at ∼10% dissociation ∼70% loss occurs. Comparing these values with those from hydrate formation, it can be seen that stiffness is much more sensitive to changes in hydrate saturation during dissociation. In the ‘excess gas’ method water initially resides at grain contacts. As hydrate forms at the water-gas interface the hydrate grows inwards into the contact and forms a bridge between the grain contacts which slowly increase in depth until all the hydrate is formed (Fig. 9a). Thus early stiffness of the specimen is dependent on the buckling resistance of the hydrate bridge. During dissociation, the hydrate shrinks inwards across the whole of its surface area. Therefore the cementing effect of the hydrate is lost almost immediately (fig. 9b) becoming a ‘frame filling’ component. Figure 10 shows the change in effective stress for specimen 10H-DU during temperature rise with all © 2011 by Taylor & Francis Group, LLC

237

ports locked off. As the specimen temperature rises a minor change in effective stress occurs (due to minor differences between the ‘equation of state’ parameters for methane and nitrogen). As the hydrate dissociates a rapid increase in pore pressure occurs causing a large reduction in effective stress. For reference the increase in pore pressure resulting from the expansion of free methane gas (calculated using Peng-Robinson’s equation of state (Peng & Robinson, 1976)) is also shown. Similar results have also been observed for carbon dioxide hydrate bearing sands (Wu et al., 2008). Thus in deep ocean sediments where undrained conditions may exist dissociation may cause large reduction in sediment strength through reduction in effective stress, regardless of the interaction between the hydrate and the sediment (cementing or frame filling behaviour). Using the excess gas method the hydrate acts as a cement, however in nature normal gas hydrates may have many different growth morphologies and the identification of hydrate, or hydrate dissociation may be more difficult to interpret through remote geophysical sensing which relies on seismic wave velocity. However, use of seismic wave attenuation (or damping) may be more susceptible to hydrate presence. Figure 11 shows the torsional, Ds and the flexural, Df damping for hydrate bearing sands during hydrate dissociation. The results show that the damping ratio increases as the hydrate start to dissociate. Although not shown a similar behaviour was also seen during formation. Maximum damping was observed when between 5–20% of the hydrate was dissociating which corresponded with the large drop in stiffness. These initial results suggest that measurement of seismic wave attenuation may be useful in determining hydrate dissociation around a well.

6

CONCLUSIONS

Laboratory experiments were conducted to investigate changes in small strain stiffness and respective damping, of Leighton Buzzard sand during dissociation

REFERENCES

Figure 11. Variation of material damping of the specimen as a percentage of hydrate dissociation.

of methane hydrate. Results show that the presence of methane hydrate in sand considerably increases both Gmax and Eflex when the ‘excess gas’ method was used to form the hydrate. The stiffness was found to be independent of both the initial effective stress (when hydrate was formed) and changes in applied effective stress (after formation), with the cementing effect of the hydrate masking the influence of grain contact compliance. Dissociation of hydrate causes a rapid reduction in stiffness of the sediment for minor changes in hydrate saturation; in contrast to that during formation where a more gradual increase in stiffness occurs. This is due to the immediate loss of cementation at the grain contacts during dissociation. Generally, numerical models which consider the effects of hydrate dissociation around a well (FreijAyoub et al., 2007) assume that stiffness is a linear function of hydrate dissociation. However, the results presented herein show an exponential reduction in stiffness during dissociation. Therefore, the inclusion of this non-linear behaviour is required to more accurately model the behaviour of sediments around an oil riser pipeline Hydrate formed using the ‘excess gas’ method produce cementation of the sediment, which may not be pertinent to all hydrate bearing sediments. However, hydrate dissociation can cause significant increase in pore pressure in sediments where the strata remain undrained. This will lead to significant reduction in effective stress and sediment strength regardless of whether the hydrate is cementing or not and may lead to heave around the oil/gas wells or internal shear failure of the sediment. Results of material damping showed an increase during dissociation which reached a maximum when between 5–20% of the hydrate had dissociated. This suggests that observations of material damping maybe more successful as a method of assessing the dissociation of hydrate bearing sediments; although further results are required to correctly correlate this assumption. © 2011 by Taylor & Francis Group, LLC

Carpenter, G. (1981), ‘Coincident sediment slump/clathrate complexes on the US Atlantic continental slope’, GeoMarine Letters, 1(1), 29–32. Cascante, G., Santamarina, C. & Yassir, N. (1998), ‘Flexural excitation in a standard torsional-resonant column device’, Canadian Geotechnical Journal, 35, 478–490. Clayton, C., Priest, J. & Best, A. (2005), ‘The effect of disseminated methane hydrate on the dynamic stiffness and damping of a sand’, Geotechnique 55(6), 423–434. Freij-Ayoub, R., Tan, C., Clennell, B., Tohidi, B. & Yang, J. (2007), ‘A wellbore stability model for hydrate bearing sediments’, Journal of Petroleum Science and Engineering, 57(1-2), 209–220. Kayen, R. & Lee, H. (1991), ‘Pleistocene slope instability of gas hydrate-laden sediment on the beaufort sea margin’, Marine Georesources & Geotechnology 10(1), 125–141. Makogon, Y. (1981), Hydrates of natural gas, PennWell Publishing, Tulsa, Oklahoma. Nimblett, J., Shipp, R., Strijbos, F. (2005), Gas hydrate as a drilling hazard: Examples from global deepwater settings, in ‘Offshore Technology Conference’, Houston, TX, May 2–5. Padden, M., Weissert, H. & De Rafelis, M. (2001), ‘Evidence for late Jurassic release of methane from gas hydrate’, Geology, 29(3), 223. Peng, D. & Robinson, D. (1976), ‘A new two-constant equation of state’, Industrial & Engineering Chemistry Fundamentals, 15(1), 59–64. Priest, J. (2004), The effect of methane gas hydrate on the dynamic testing of cohesive sand, PhD thesis, University of Southampton, Southampton. Priest, J., Best, A and Clayton, C. (2005), ‘A laboratory investigation into the seismic velocities of methane gas hydrate-bearing sand’. Journal of Geophysical ResearchSolid Earth, 110(B4), B04102. Priest, J., Kingston, E. & Clayton, C. (2008), The structure of hydrate bearing fine grained marine sediments, in ‘6th International Conference on Gas Hydrate’, Vancouver, British Columbia, CANADA. Priest, J., Rees, E. & Clayton, C. (2009), ‘Influence of gas hydrate morphology on the seismic velocities of sands’, Journal of Geophysical Research-Solid Earth, 114(B11), B11205. Saxena, S., Avramidis, A. & Reddy, K. (1988), ‘Dynamic moduli and damping ratios for cemented sands at low strains’, Canadian Geotechnical Journal, 25(2), 353–368. Sloan, E. D. (1998), Clathrate Hydrates of Natural Gases, New York: Marcel Dekker. Stern, L., Kirby, S. & Durham, W. (1996), ‘Peculiarities of methane clathrate hydrate formation and solid-state deformation, including possible superheating of water ice’, Science, 273(5283), 1843. Tan, C., Freij-Ayoub, R., Clennell, M., Tohidi, B. & Yang, J. (2005), Managing wellbore instability risk in gas hydratebearing sediments, in ‘SPE Asia Pacific Oil and Gas Conference and Exhibition’. Waite, W., Winters, W. & Mason, D. (2004), ‘Methane hydrate formation in partially water-saturated Ottawa sand’, American Mineralogist, 89(8–9), 1202–1207. Wang, Y., Cascante, G. & Santamarina, J. (2003), ‘Resonant column testing: the inherent counter emf effect’, ASTM Geotechnical Testing Journal, 26(3), 342–352. Wu, L., Grozic, J. & Eng, P. (2008), ‘Laboratory analysis of carbon dioxide hydrate-bearing sands’, Journal of Geotechnical and Geoenvironmental Engineering, 134, 547.

238

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

An investigation of past mass movement events in the West Nile Delta S. Thomas, L. Bell & K. Ticehurst Fugro GeoConsulting, Wallingford, UK

P.S. Dimmock BP Exploration and Production Technology, Sunbury, UK

ABSTRACT: BP Egypt and its equity partner RWE Dea are studying the potential for deepwater natural gas developments in the West Nile Delta (WND), Offshore Egypt. These developments face numerous potential geohazards which are being managed using a 3D geological and geotechnical ground model. This model, together with an appreciation of sedimentary process and environmental controls on the delta, has aided the understanding of geohazard type, magnitude and frequency. This paper demonstrates the integral place of detailed sedimentological and ichnological logging of specially acquired long piston cores at the heart of the investigation into the frequency and magnitude of submarine mass movements. It is shown that the intelligent integration of the results from the specialist geohazard core logging with ultra high resolution AUV CHIRP data can significantly modify the perception of type, size and number of submarine mass movements that have occurred in the proposed development area.

1

BACKGROUND

BP Egypt and its equity partner RWE Dea are studying the potential for deepwater natural gas developments in the West Nile Delta (WND), approximately 50 km offshore Egypt in water depths between 100 m and 1100 m. The area presents numerous geohazard challenges to development activities, including seabed slope instability, turbidity currents, mud volcano activity, pockmarks, fault displacement, seismicity, bioherms, seabed erosion, tsunamis, drilling hazards, variable soils and topographic constraints (Evans et al., 2007). This paper provides some insights in to the approach applied by the integrated study team to the investigation of the historical frequency, type and size of mass movements. This definition is required in order to quantify the potential hazard and assess geohazard risk and mitigation strategies. A key component of this strategy is the detailed sedimentological and ichnological logging of specially acquired long piston cores and the subsequent integration of these findings with the existing 3D ground and evolutionary model. Figure 1. West Nile Delta development areas showing location of data used in Figures 3 to 6.

2 THE GROUND MODEL An area wide program of data acquisition in the West Nile Delta has been performed which includes 3D Exploration seismic, 2D ultra high resolution (UHR) seismic and ultra high resolution CHIRP profiler data acquired from an Autonomous Underwater Vehicle (AUV). These surveys, combined with optimized © 2011 by Taylor & Francis Group, LLC

phases of geotechnical sampling and testing, form the basis of the development of the ground model in the WND (Evans et al., 2007). The acquisition of ultra high resolution AUV CHIRP data provided the necessary resolution in the shallow section (up to 50 m

239

below seafloor) for use in geohazard and geotechnical studies. Region wide seismic horizons were identified and interpreted which group the sediment column into distinct seismostratigraphic units and enable the correlation of units across the whole of the development area. Factual morphological and interpretive geomorphological mapping was performed to capture the distribution and extent of seabed features that are interpreted as being related to mass movement processes. Individual buried mass movement events were further identified and mapped from the AUV CHIRP data using acoustic character and geometry to define the sub bottom or seismic geomorphology. Experience in the WND has shown mass movement deposits to be characterized by a generally structureless acoustic character and an irregular basal contact with the underlying sediments (Figure 6). The seismostratigraphic units interpreted to be mass movement deposits are mapped to determine thickness, length, width and volume. These data are vital for use in geomechanical modelling of landslides, debris flows and runout culminating in turbidity currents.

3 APPROACH TO CORE ACQUISITION In order to calibrate the interpretations derived from the seismic data, cores need to be acquired at locations where the seismostratigraphic model identifies mass movement deposits. Sampling may be performed either using a long piston corer or downhole drilling and sampling equipment depending on the depth of the features to be sampled and tested. The cores should be preferentially located on cross lines of the AUV CHIRP in order to maximize the correlation of the core derived data and interpretations of the seismic data. To improve the confidence in interpretation and correlation of the data the core samples must be specifically acquired for detailed geohazard core logging. Where cores are sub-sampled for geotechnical testing, significant sections of the stratigraphic record are removed allowing for whole event sequences to be missed, adding uncertainties to the derivations of event frequency and magnitudes (Thomas et al., 2010). This approach has been adopted in the WND and long piston cores were specifically acquired at locations where mass movement deposits were identified from geomorphological and geophysical data interpretation. These cores were targeted at locations demonstrating different geomorphological features, including pull apart windows formed by a slab slide (Figure 3), spreading failure (Figure 4), mass movement runout (Figure 5) and a debris apron at the base of a steep scarp slope (Figure 6). It should be noted that geohazard specific cores were acquired as part of geotechnical site investigations designed to characterize the soil conditions regionally across the development area and calibrate the ground model. At most locations continuous seabed cone penetration tests (CPT) were acquired to 20–40 m penetration below seabed. These tests provide © 2011 by Taylor & Francis Group, LLC

a geotechnical stratigraphy for the cores, which complements the stratigraphic information from the logging by providing a geotechnical understanding of both the in situ and disturbed sediments. In addition to the CPT, a box core was acquired to supplement the sampling of the youngest seabed sediments, which is considered crucial for dating the most recent events. 4

GEOHAZARD CORE LOGGING

The cores acquired in the WND were carefully logged in order to identify sedimentological and ichnological evidence for mass movement deposits identified from the geophysical data and detailed geomorphological mapping. The core logging process identifies key lithological features including colour, primary and secondary soil type, discontinuities, minor constituents and contacts. Based on unique lithological features of the sediment package a facies classification can be developed. Sediments of a particular facies have similar sedimentological, geotechnical and ichnological properties which can be used to identify the depositional environment. A key outcome of this classification process is the differentiation between primary depositional process, for example hemipelagic deposition, and post depositional events, such as mass movements. These post depositional processes can be captured by the addition of facies modifiers, which demonstrate that the primary sedimentary structure, such as well bedded hemipelagite, can be clearly recognized in spite of post depositional failure. The facies classification applied in the WND indicates the presence of landslide, debris flow, mudflow and turbidity current deposits interbedded with hemipelagic sediments (Figure 6). The relationship between the facies is used to develop an event stratigraphy which precisely delineates the type, number and thickness of hemipelagic sediment packages and intervening mass movement deposits. 5

BIOSTRATIGARPHIC ANAYSES AND GEOCHRONOLOGICAL TESTING

It is important to highlight that it is not possible to obtain meaningful dates for the actual mass movement events inferred from the facies identified in the cores. This stems from the uncertainties in stratigraphic position and true origin of the materials extracted from the mass movement deposits for age dating. Instead, the age of the event is constrained by obtaining age dates from in situ sediments above and below the disturbed sediments. In the case of the WND, hemipelagites, which are inferred to represent the normal mode of in situ sediment accumulation, have been used for age dating purposes. In order to obtain meaningful estimates for the ages of the mass movement events interpreted from the ground model and inferred precisely from the facies identified in the cores, a selective program of

240

biostratigraphic, radiometric and luminescence testing is required. The tests are carefully targeted on appropriate material from the sediment whose depositional origin and stratigraphic position is accurately and precisely known from the geohazard core logging. The biostratigraphic techniques include micropaleontology, primarily the planktonic and benthic foraminiferal assemblages, and palynology, primarily terrestrial pollen assemblages. These analyses provide corroboration of the depositional environment interpreted from the facies and confirms the origin of the in situ hemipelagite, the primary host sediments for age dating purposes. The depositional environment then provides a proxy correlation to the global and regional, climate stratigraphy and the evolutionary history of the River Nile and its catchment. The analysis of the land derived pollen assemblages proved particularly useful in providing information on the origin of the sediment components and hence insights in to the depositional processes that formed the sediments. The ichnological character of the sediments is investigated to further define the potential history and extent of bioturbation within the selected sediment so that this can be avoided in the sub sampling process. Where the bioturbation is well preserved, it can be classified and in some instances provide a proxy qualitative and relative age dating technique as part of the facies based approach. On completion of this screening process, materials for radiometric and luminescence dating are extracted from the selected sediments. In the WND, an optimized program of radiometric dating techniques including 14 C, 210 Pb, 137 Cs, and 87 Sr/86 Sr along with optically stimulated luminescence (OSL) have been performed to constrain the ages of specific events, but also to calibrate the seismostratigraphic framework of the 3D ground model. The primary dating techniques used were 14 C for the sediments which are up to about 45 ka BP (thousand years before present) and OSL for those up to about 250 ka BP. The sediments estimated to have been deposited in the last few hundred years were further corroborated by obtaining a small number of 210 Pb and 137 Cs dates. Those estimated to be in excess of 45 ka BP and 250 ka BP, the current limits of 14 C and OSL respectively, were dated using 87 Sr/86 Sr ratio analysis. The approach adopted here provides for overlap of age dating techniques in the sedimentary column to permit the analysis of age variance between the techniques. A coherent set of results was obtained from the various techniques indicating that a robust and fit for purpose chronological framework has been established for the study area.

6

DATA INTEGRATION AND CASE STUDIES

The process of data integration is iterative with each stage developing from the findings of previous stages. Figure 2 summarizes the approach adopted in the WND, from the regional ground model through the

© 2011 by Taylor & Francis Group, LLC

Figure 2. Flowchart summarising the data integration approach adopted in the WND.

geohazard core logging and to a site specific facies classification. The next step of the process is to carry out a further iteration of data integration and interpretation to investigate the significance of individual events to refine the regional ground and evolutionary model. This section provides an insight in to this process using four case studies (Figures 3 to 6). The core in Figure 3 is located in a pull-apart window formed by a surface slab slide, failure. The pull-apart window in this case refers to the area vacated by the slab of sediment that has moved down slope exposing older sediments at seabed. In order to get an indication of the possible age of this event it is necessary to obtain a date in the first hemipelagite sediments that were deposited on the recently exposed seabed. The AUV CHIRP data indicated the presence of a thin accumulation of apparently conformable sediments on the failure surface at this location. The logging of the geohazard core indicates that immediately above the exposed failure surface there is a thin sequence of mass movement deposits, which are inferred to represent failures from the steep slopes along the newly formed headwall scarp resulting from the slab slide failure. These deposits are overlain by hemipelagites representing the post failure resumption of the normal pattern of in situ sediment accumulation. These sediments form a conformable sequence of sediment types or facies, which in themselves provide a proxy indication of age from correlation to the wider ground and evolutionary model. This is known as a facies association. The in situ hemipelagite sediments were dated using 14 C. In addition, the undisturbed sediments below the failure surface were also dated to constrain the age of the event and together indicate that the failure occurred between 8,700 a BP and 3,300 a BP. Evidence from other dates for the hemipelagite facies association

241

Figure 5. Example integration of data sets showing mass movement runout area.

Figure 3. Example integration of data sets showing pull-apart window formed by a slab slide.

Figure 6. Example integration of data sets within a debris apron at the base of a steep scarp slope.

Figure 4. Example integration of data sets showing spreading failure.

found here suggest that the failure most probably occurred towards the younger end of the age range. The core in Figure 4 is located in a surface, retrogressive spreading landslide failure. The AUV CHIRP data shows an acoustically structureless seismic unit apparently without any identifiable stratified sediment accumulation, commonly called drape, overlying it. Based on the interpretation of the seismic data alone, it would be considered reasonable to infer a very recent event. However, the core logging clearly showed that stratified hemipelagite sediments were in fact present overlying the structureless seismic unit. The sediment facies association identified at this location appeared to be intact and the date indicated that the event was older than 2,300 a BP. The facies association in itself provided a proxy age indication derived from the understanding of sediment accumulation across the WND. This is still a relatively recent event in an engineering timescale, but not as young as the seabed © 2011 by Taylor & Francis Group, LLC

geomorphology and seismic data alone would have initially suggested. The core in Figure 5 is located in an area that is inferred to have experienced past mass movement runout events originating on the scarps to the east and south where localized slopes exceed 20◦ . The AUV CHIRP data show three acoustically structureless seismic units separated by thin acoustically well bedded units, which are inferred to be mass movement deposits and hemipelagites respectively. Age dates were obtained from the undisturbed sediments and by inference from the regional facies model. Based on this interpretation, it would be credible to infer three large events: 1) >9,200 a BP; 2) between 7,000 and 6,500 a BP; 3) between 860 and 500 a BP (Figure 5). The core logging results indicate that the mass movement deposit 2 can be divided in to at least two smaller events. No hemipelagite was observed between the mass movement events, which indicate that they may have occurred close in time or even been

242

associated in some way. For example, a slope failure on the scarps could produce the initial runout event seen in the core and then subsequent failure of the newly exposed steep slopes could occur generating another runout event. Alternatively, the intervening in situ sediment accumulation could have been removed by the subsequent event leading to an erroneous understanding of the timing of the mass movement events. The understanding of the most likely mass movement process and chronology operating in the area and their setting within the ground and evolutionary model is essential to the correct interpretation of the facies association. The core in Figure 6 is located in a debris apron at the foot of the Taurus Scarp formed by runout from mass movements originating on the steep slopes of the scarp. The geomorphological mapping of the seafloor shows a series of lobes from individual runout events that have coalesced to form the wedge-shaped apron along the scarp. Again, the AUV CHIRP data shows a number of acoustically structureless seismic units that are locally separated by thin acoustically well bedded units at depth. In the uppermost part of the section, the acoustically structureless unit is separated by a single seismic horizon. Based on these data, it could be inferred that the event stratigraphy at this location comprised two large events that occurred relatively close in time and two older, smaller events more separated in time. However, the results of the detailed core logging and geochronological testing showed that the thickest acoustically structureless unit could be divided in to seven much smaller mass movement events which are inferred to have occurred between 8,200 and 4,200 a BP. These findings significantly modified the perception of the type, size and number of mass movement events in this part of the development area. Specifically, the number of events increased while the size of each event decreased significantly. Based on this understanding, it was possible in some cases to map individual lobes and better model the mass movement process that formed them. This synthesis provides a best estimate for frequency and magnitude data for input into the geohazard assessment and geotechnical mitigation studies described in more detail by Evans et al. (2007). These examples clearly illustrate the need and benefit of thorough and systematic integration of geophysical, geomorphological, lithological and geotechnical data as described and illustrated in Figure 2. 7

CONCLUSIONS

Exploration seismic, including reprocessed 3D, data can be invaluable in the initial identification of mass movement deposits within an area. The limitations in resolution of 3D data allow mapping of large scale mass movement deposits only. This does not provide an accurate assessment of magnitude and frequency of mass movement events for use in geohazard risk assessments. © 2011 by Taylor & Francis Group, LLC

The identification and mapping of smaller scale mass movement deposits, and high resolution mapping of the larger deposits, requires the acquisition of 2D UHR and AUV CHIRP data. Interpretation of the AUV CHIRP data is essential to provide the necessary resolution to accurately identify deposits of past mass movement events. A comprehensive understanding of magnitude, frequency and depositional process is, however, limited by the sole use of geophysical data. It has been observed in the WND, that mass movement deposits identified from AUV CHIRP data alone may overestimate the magnitude of an event and consequently understate the frequency. The use of AUV CHIRP data allows the targeting of strategically located long piston cores and is critical for performing a site specific and accurate assessment of frequency and magnitude of past mass movement events. It is important to ensure the long piston cores are sited on cross lines of the AUV CHIRP survey to enhance the confidence in integration of the data. Specialist geohazard core logging of long piston cores identifies key sedimentological features, which through the development of a sedimentological facies model, facilitates the interpretation of depositional processes. This event stratigraphy clarifies the type and number of events, as well as a finer resolution to the thickness of mass movement deposits initially interpreted from the geophysical data alone. For example, in cores sampled in the WND, a single mass movement deposit interpreted from the geophysical data was observed to consist of a complex stratigraphy of hemipelagites interbedded with debrites, slide deposits, turbidites and mud flow deposits. It should be borne in mind that only through the integration of the complete event stratigraphy with the geophysical data and geomorphological interpretation can the magnitude, spatial extent and distribution of the mass movement deposits in the area be fully understood. In order to achieve an accurate assessment of the frequency of events at a location focused biostratigraphic analyses and geochronological testing are required. These techniques can be used to develop a temporal framework for deposition of past mass movement events. It is essential to ensure interpretations from the geohazard core logging are used to target the testing on sediments with a known depositional process to ensure the success of the geochronological testing program. The synthesis of the available data can be used to inform and focus geohazard risk assessments. In turn, mitigation studies, including runout impact studies such as for example Niedoroda et al. (2000), Bruschi et al. (2006), Parker et al. (2008), Zakeri, (2010) should be performed. These studies aid in identifying criteria for avoidance or development, through geohazard resistant design, at a specific location. ACKNOWLEDGEMENTS The authors acknowledge the many contributions made by members of BP Egypt, EPT and GAT in

243

carrying out the work detailed in this paper. Moreover, the authors acknowledge the specialist studies that have been carried out by StrataData Limited. Thanks also go to Fugro for providing the time and materials with which to write this paper. Finally, the authors are grateful to BP Egypt, RWE Dea and Egyptian Natural Gas Holding Company (EGAS) for their support. REFERENCES Bruschi, R., Bughi, S., Spinasse, M., Torselletti, E. & Vitali, L. 2006. Impact of debris flows and turbidity currents on seafloor structures. Nor J Geol 89: 317–337 Evans, T., Usher, N. & Moore, R. 2007. Management of Geotechnical and Geohazard Risks in the West Nile Delta. Proceedings of the 6th International Offshore Site Investigation and Geotechnics Conference: Confronting New Challenges and Sharing Knowledge, London, UK, 11–13 September 2007 Niedoroda, AW., Reed, C.W., Parsons, B.S., Breza, J., Forristall, G.Z. & Mullee, J.E. 2000. Developing engineering

© 2011 by Taylor & Francis Group, LLC

design criteria for mass gravity flows in deepsea slope environments. Proceedings of Annual Offshore Technology Conference, OTC Paper Number 12069 Parker, E.J., Traverso, C., Moore, R., Evans, T. & Usher, N. 2008. Evaluation of landslide impact on deepwater submarine pipelines. Proceedings of Annual Offshore Technology Conference, Houston, Texas, 5–8 May 2008, OTC Paper Number 19459 Thomas, S., Clare, M. & Hooper, J. 2010. Constraining geohazards to the Past: Impact Assessment of Submarine Mass Movements on Seabed Developments. In D.C. Mosher et al. (eds.), Submarine Mass Movements and Their Consequences; Advances in Natural and Technological Hazards research, Vol. 28. Springer Science and Business Media B.V. Zakeri, A. 2010. Estimating Drag Forces on Suspended and Laid-on-Seafloor Pipelines Caused by Clay-Rich Submarine Debris Flow Impact. In D.C. Mosher et al. (eds.), Submarine Mass Movements and Their Consequences; Advances in Natural and Technological Hazards research, Vol. 28. Springer Science and Business Media B.V.

244

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Deformation of seabed due to exploitation of methane hydrate reservoir J. Yoneda, M. Hyodo, Y. Nakata & N. Yoshimoto Yamaguchi University Graduate school of Science and Engineering, Yamaguchi, Japan

R. Orense University of Auckland, Auckland, New Zealand

ABSTRACT: Due to recent investigations, methane hydrate is expected to become a possible future energy resource. Both thermal recovery methods and depressurization methods have been suggested and developed for exploiting the methane hydrates from reservoirs that exist in deep ocean floors. Using both methods, methane hydrates in the ground are dissociated to release the methane gas. During this process, the mechanical strength of the sediments may change as temperature and pore water pressure in the ground change in the vicinity of the production well. In this study, the change in soil strength is evaluated using finite element analysis. As a result of a decrease in hydraulic pressure, consolidation will occur with a corresponding decrease in the mechanical strength of the ground caused by methane hydrate decomposition, which in turn causes the sea floor around the production well to generate wedge shaped distribution of shear strains. 1

2

INTRODUCTION

A project to extract methane hydrate (herein after referred to as MH), which is expected to be an energy resource in the next generation, from the deep sea-bed has been advanced (Research Consortium for Methane Hydrate Resources in Japan). MH is composed of a methane molecule and water molecules and exists as an ice-like crystal under low-temperature and highpressure condition. In the MH extraction project, a well is drilled into the sea floor from a marine platform. Then, fluids in the well are either heated or depressurized to induce MH dissociation and the dissolution of methane gas is collected in-situ. During MH production, there are concerns about the settlement of the seabed and the possibility that landslides will occur due to change in effective stress induced by drilling, water movement due to depressurization, dissociation of MH, methane gas generation and thermal change, which are all inter-connected. In addition, rebound of the ground and possible landslides caused by the reduction in effective stress accompanying the water pressure recovery after the end of production are also important issues. Therefore, an analysis of soil behavior is required to estimate the resulting ground deformation as a result of the extraction process The authors (2005) have performed triaxial compression tests on MH bearing-sand and undisturbed soils under high-pressure and low-temperature, and have developed a constitutive model which can express deformation behavior (2008). In this paper, ground deformation in the vicinity of MH exploitation is predicted using a soil-water-gas-heat coupled finite element method with the developed elasto-plastic constitutive model. © 2011 by Taylor & Francis Group, LLC

2.1

ELASTO-PLASTIC CONSTITUTIVE MODEL FOR MH-BEARING SAND Basic features of MH model

The main features of the constitutive model are described as follows. Figure 1 shows the equivalent yield function of the proposed model in mean stress p
-deviator stress q space. It is based on the modified Cam-clay model. The plastic strain increment is defined by an associative flow rule. To express the elasto-plastic deformation within the yield surface, the subloading surface model (Hashiguchi and Ueno

245

Figure 1. Constitutive model of MH-bearing sand on p
-q space.

1977) is introduced for expressing plastic deformation at inside of yield surface. The cementing action due to the presence of MH is introduced as internal stress component, pint . The initial internal stress, pint , is determined from MH saturation. The expression for the subloading surface is given by Equation (1).

3.2

Equation of continuity for water

Equation of continuity for water is expressed as follows.

where the left side refers to the change in water density and water content per unit time. On the right side, qW is the amount of inflow of water from the boundary, and m ˙ W is the amount of water generation due to MH dissociation. The dissociate equation of the methane hydrate used the expression that Kim et al. (1995) suggested.

where p0
is yield stress for methane hydrate-bearing sand, p∗0
is yield stress for host sand, R is similarity ratio between normal yield and subloading surfaces, M is stress ratio at critical state.

3.3

Equation of continuity for gas

Equation of continuity for gas is expressed as follows.

2.2 Internal stress pint and evolution law From past research, the relationship between damage due to internal stress and total work, MH saturation, temperature and pressure had been confirmed (Yoneda et al. 2007). Thus, dpint is assumed by introducing the internal stress on the non-elastic work dWp in the modified Cam-clay model as follows:

where the left side indicates the change in gas density and gas content per unit time. On the right side, qG is amount of inflow of gas from boundary, and m ˙ G is the amount of gas generation due to MH dissociation. 3.4

where ζ is a coefficient that converts temperature and water pressure dependency expressed by L and SMH into pint0 . The parameter L is defined (see Hyodo et al. 2008) which relates the initial temperature /water pressure conditions and the stable boundary of MH.

Law of the conservation of energy

The law of the conservation of energy within the ground is expressed as follows.

In order to express deformation of the ground, movement of water, generation of gas, decomposition of MH and energy movement that are assumed to take place during development, soil-water-gas-heat coupled FEM is conducted. At the first, it is supposed that the ground is elastic, and the governing equations are formulated and implemented through finite element method.

In the above equation, the left side consists of energy change per unit weight per volume. On the other hand, the first term of the equation on the right-hand side is the transportation of water and gas, the second term is the thermal conductivity, the third term is the thermal energy produced by deformation, and the last term is the energy change due to MH dissociation. In the above-mentioned equation, T is temperature, cα is specific heat, vαi is velocity, Kα is thermal conductivity, and β is coefficient of thermal expansion.

3.1

3.5

3

GOVERNING EQUATIONS OF SIMULATION

Equilibrium equation

Equilibrium equation for soil skeleton is expressed as follows using the effective stress principle of Bishop (1959):

where σ
is effective stress, SW is water saturation ratio, PW is pore water pressure, SG is gas saturation ratio, PG is gas pressure, and γ is body force. It is assumed that the ratio of water pressure and gas pressure gives an effective stress which is equal to each saturation ratio in the unsaturated zone. © 2011 by Taylor & Francis Group, LLC

Discrete formation of governing equations

The governing equations are discretized using the method of weighted residuals, and finite difference method is used for the time discretization. In addition, the infiltration characteristics in an unsaturated zone is assumed to follow the soil-water characteristic curve model of van Genuchten(1980), and the coefficient of permeability of water and gas in the unsaturated ground is based on Mualem model(1976). In the said model, the coefficient of the unsaturated permeability of water is determined by assuming the parameter m = 1-1/n which is given in the van Genuchten model to simplify the formulation.

246

Table 2.

Parameters for FEM analysis.

Parameter for FEM analysis

Figure 2. Finite element mesh and boundary conditions for model of seabed. Table 1.

Conditions of each simulation.

Production method

Case of symulation

Condition of depressurization source

Case 1

Vertical well (Width of Depressurization 40 m) Point well Horizontal well (Width of Depressurization 20 m) Horizontal well (Width of Depressurization 40 m) Horizontal well (Width of Depressurization 60 m) Horizontal well (Width of Depressurization 100 m) Horizontal well (Width of Depressurization 200 m)

Depressurization

Case 2 Case 3

Hydrostatic

Case 4

pressure→4(MPa)

Case 5 Case 6 Case 7

Upper & Lower layer

kN/m3 kN/m3 kN/m3 m/sec m/sec

1.0 × 105 0.33 25 10 9.12 1.0 × 10−6 1.0 × 10−4

1.5 × 105 0.33 25 10

/K cal/m·K·sec

0.4 0.2 (SMH = 50%) 5.0 × 10−6 0.14

5.0 × 10−6 0.14

cal/m·K·sec

0.5

0.5

cal/m·K·sec

0.1

0.1

cal/m·K·sec

0.1

cal/N·K cal/N·K cal/N·K cal/N·K g/mol

2.5 × 104 1.0 × 105 1.0 × 104 1.0 × 105 16

2.5 × 104 1.0 × 105 1.0 × 104

g/mol

18

18

g/mol

119.5

119.5

kPa·m3 / K·mol

8.314

8.314

Symbol

Parameter

Unit

E ν ρs ρw ρH kw kG n nH

Elastic modulus Poisson’s ratio Unit weight of soil Unit weight of water Unit weight of hydrate Permeability Air permeability Initial porosity Volumetric ratio of MH Expansion coefficient Thermal conductivity of water Thermal conductivity of soil Thermal conductivity of gas Thermal conductivity of MH Specific heat of water Specific heat of soil Specific heat of gas Specific heat of MH Molecular weight of methane gas Molecular weight of water Molecular weight of methane hydrate Gas constant

kPa

A KtW KtS KtG KtH CvW CvS CvG CvH MG MW MH R

4

MH concentrated layer

1.0 × 10−5 1.0 × 10−4 0.4 0

16

DETAILS OF SIMULATION

A simulation was performed assuming MH production using the depressurization method employed in the Nankai Trough region in Japan. The analysis was performed assuming an elasto-plastic plane strain problem. Finite element mesh and boundary conditions for model of seabed is shown on Figure 2. The model, consists of 400 elements and 1279 nodes, has an area of 200 × 500 m, and is located at 800 m water depth. It is assumed that the layer 100 m∼150 m from the ground surface as the MH-bearing layer with a MH saturation level of 50%. Seven different cases were analyzed depending on the condition of the depressurization source, as indicated in Table 1. In Case 1, the water pressure at a certain region on the left side of the model (Nodes 373, 435, 497 and 559 of the MH-bearing sedimentary layers) was depressurized over 27 hours from hydrostatic pressure to 4 MPa, and methane gas is produced by maintaining the pressure at the source of depressurization for the next eight years. Furthermore, the depressurization source is assumed to have free boundaries after that for eight years in order to simulate water pressure recovery. In Case 2, depressurization is performed only at Node 435 (to a represent point well), and Case 3 to Case 7 consider depressurization at all horizontal nodes of contact extending to a pre-defined width from Node 435; these are indicated as width of © 2011 by Taylor & Francis Group, LLC

247

depressurization in Table.1. After that, the recovery of hydraulic pressure is simulated in the same way as in Case 1. The boundary conditions for the simulation are shown in Figure 2. 5

RESULTS OF MH PRODUCTION BY USING DEPRESSURIZATION METHOD

Figure 3 shows the effect of production on the distribution of water pressure of the model of case 1. It is seen that the reduction in the pressure is centered at the source of depressurization. On the other hand, Figure 4 shows the decrease in temperature through the model of case 1 resulting from MH decomposition. It is observed that there is no significant difference in the vicinity of the region where the methane hydrate had been decomposed as compared to the initial condition. The ground deformation of case 1, magnified 50 times, is illustrated in Figure 5. It can be seen that compression of the model is concentrated near the source of the depressurization and the top of the model just above the production well showed maximum settlement. This is due to the consolidation caused by increase in effective stress and decomposition of MH. Figure 6 shows the distribution of MH saturation ratios with the decomposition of MH. The region spreads by about

Table 3.

Parameters for the constitutive model.

Parameter for the constitutive model

Symbol Parameter λ

κ

pi ei M u

α

ξ χ ς

MH Upper & concentrated Lower Unit layer layer

slope of the normally 0.146 consolidated line in
e-logp space slope of the 0.0016 overconsolidated line in e-logp
space initial mean MPa 1.2 effective stress void ratio at mean 0.973 effective stress pi stress ratio at critical state 1.2 material constant in 10 evaluation of similarityratio R material constant to 10 evaluate λ and κ of MH bonding sand material constant for yield 10 stress p
0 relative to decreasing of 5 internal stress material constant to 0.05 evaluate pint

0.146

Figure 6. Saturation of methane hydrate (Case1).

0.0016

1.2 0.973 1.2 10

Figure 7. Shear strain at the seabed (Case1). ·

· · ·

Figure 8. Displacement of the top of the seabed (Case1).

Figure 3. Water pressure of the model (Case1).

Figure 4. Temperature of the model (Case1).

Figure 5. Deformation of the model (Case1).

10 m wide from the production well; however, MH decomposition becomes constant after that because the condition of water pressure and temperature is stable for MH. © 2011 by Taylor & Francis Group, LLC

Figure 7 shows the shear strain distribution at the seabed. Note that shear strains occur at the boundary of MH-bearing layer and the upper/lower layer like a wedge as a result of differential settlement. The relationship between the settlement at various points in the seabed and elapsed time is depicted in Figure 8. The trend shows that the settlements continue after depressurization for 27 hours due to the delay caused by drainage, and it became constant after 3 days. Figure 9 shows the relationship between MH saturation ratio and elapsed time, with emphasis on the elements located near the depressurization source (the enlargement shows node location). Elements 141 and 161 showed immediate response when MH was decomposed while Element 121 showed late response because of the difference in geothermal energy. Decomposition advanced after that and was completed after 8 days at the location of Elements 141 and 161. Figure 10 illustrates the effective stress path for each element. Each element showed an increase in effective stress with decrease in water pressure due to depressurization. During production, the deviator stress in each element did not exceed the limit stress ratio, and the volume change tendency showed only compression. Figure 11 depicts the effective stress path of the elements located at the boundary between MH

248

Figure 9. Relationship between methane hydrate saturation and time.

Figure 12. Settlement of seabed in each case after 8 years.

Figure 10. Effective stress path of each element.

Figure 13. Maximum displacement of each case.

Figure 11. Deviator stress at each element.

concentrated layer and the upper layer which showed maximum shear strain (the enlargement show node locations). An element at the well region shows the highest deviator stress, with the tendency of increasing deviator stress and decreasing effective stress with increasing distance from the well. The settlements at the seabed for each case investigated herein are shown in Figure 12. The settlements increase with the width of depressurization and they are concentrated at the top of the well. These caused the pumping discharge per unit time to increase by having increased horizontal distance in the depressurization source to undergo consolidation. Figure 13 shows © 2011 by Taylor & Francis Group, LLC

249

Figure 14. Settlement of seabed in each case after water pressure recovered.

maximum settlement of each case. It is understood to settle the amount of the maximum settlement on the seabed according to an increase in the width of depressurization gradually. Because consolidations caused by the depressurization were spread widely with a longer distance of the depressurization source, and deform vertical axis remarkably as a one dimensional problem. Figure 14 shows the settlement for each case after the production of MH stopped and simulate the recovery of hydraulic pressure. The results show the plastic

decreasing effective stress while the water pressure was being recovered. It was expected that the deformation of MHconcentrated layer due to application of depressurization method varies greatly according to the width of the depressurization source. However, for in-situ ground, clay layers, difference in permeability, etc. would influence the depressurization region in the subsurface. Therefore, modeling real ground conditions accurately is an important area for future research. ACKNOWLEDGEMENTS Figure 15. Effective stress paths of elements where shear stresses are concentrated.

strains caused by using the depressurization method. Figure 15 (Case 1) shows the effective stress paths for each element located in the horizontal direction from the production well up to about 100 m in depth from the ground surface (region where the shear strains are most concentrated). For each element, the effective stress increases initially with depressurization of water pressure. The deviator stress of each element does not exceed the limit stress ratio. Therefore the dilatancy behavior shows only compression. After that, the deviator stress at each element decreases with decrease in the effective stress when the hydraulic pressure is recovered. The elements which are in the neighborhood of the depressurization source showed the highest deviator stress.

6

CONCLUSIONS

In this paper, soil-water-gas-heat coupled finite element analysis was employed, incorporating the elastoplastic constitutive model developed by the authors, to predict the deformation of the ground around regions of MH exploitation. It was found that volumetric strain became the greatest near the source of depressurization, and it became clear that the greatest shear strain occurred at the boundary of the MH-bearing layer and the upper layer. In addition, the deviator stress that occurred during depressurization decreased with

© 2011 by Taylor & Francis Group, LLC

This research has been conducted under the auspices of the MH21 Research Consortium. Thanks are due to our sponsors and partners. REFERENCES Bishop, A. W. 1959. The principle of the effective stress, Teknisk Ukeblad 106(1):859–863. Hashiguchi, K. & Ueno, M. 1977. Plastic constitutive law of granular materials. Constitutive equations of soils, JSSMFE, 73–82. Hyodo, M., Nakata, Y., Yoshimoto, N. & Ebinuma, T. 2005. Basic research on the mechanical behavior of methane hydrate-sediments mixture. Soils and Foundations 45(1): 75–85. Hyodo, M., Nakata, Y., Orense, R.P., Yoshimoto, N. & Yoneda, J. 2008. Elastoplastic constitutive equation for methane hydrate-supported sand in deep seabed. Proc. of the 14th International Symposium on Plasticity 2008: 349–351. Kim, H.C., Bishinoi, P.R., Heidemann, R.A. & Rizvi, S.S.H. 1985. Kinetics of methane hydrate decomposition, Chem. Eng. Sci. 42: 1645–1653. Mualem, Y. 1976. A new model for predicting the hydraulic conductivity of unsaturated porous media.Water Resour.Res 12: 513–522. Research Consortium for Methane Hydrate Resources in Japan: http://www.mh21japan.gr.jp. Van Genuchten, M. T. 1980. A closed-form equation for predicting the hydraulic conductivity of unsaturated soils. Soil Science Society American Journal 44: 892–898. Yoneda, J. Hyodo, M., Nakata, Y., Yoshimoto, N. & Ebinuma, T. 2007. Mechanical property of gas hydrate sediment at deep seabed in triaxial compression test. Ground Engineering 25: 113–122.

250

3 In situ site characterisation and pore pressure measurement

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils D. Borel, A. Puech & S. Po Fugro Offshore Geotechnics, Nanterre, France

ABSTRACT: The design of suction caissons and the determination of soil-pipe interactions in deep water require an accurate assessment of the intact and residual undrained shear strength of the sediments within the upper 30 metres of penetration. Particular care is necessary near seabed for flowline stability assessment where extremely soft soils are encountered. Deep water investigations are routinely based on continuous and/or discrete in-situ testing methods complemented by high quality sampling and laboratory testing. The increasing customer demand for fast track engineering parameters and the availability of new tools push to review the organisation of geotechnical deep water site investigations. A revised strategy is proposed which is aimed at obtaining a preliminary but reliable set of engineering parameters from direct interpretation of in-situ tests.

1

INTRODUCTION

Measuring accurate soil parameters, in particular shear strength profiles, remains a challenge in deep water environments. The sediments are often characterised by extremely low values of undrained shear strength at seabed, increasing linearly with depth. The development of deepwater hydrocarbon reservoirs started in the mid 90s and has boomed since the early 2000s. From the early days, geotechnical techniques and tools were improved and adapted in order to closely fulfill the new needs. The main evolution drivers during this period were 1/ to improve the duration of the operations and decrease the investigation costs, 2/ to adapt the equipment to the deep water pressures and 3/ to adapt the in-situ measuring sensors to the very low resistances of the soils. Kolk & Wegerif (2005) present an overview of the techniques and tools together with the main challenges for the offshore geotechnical investigation. Recently, new equipment, developed specifically for near seabed soil characterization, expanded the range of available tools. In parallel, the oil and gas industry continuously shortens the field development cycles and requires contractors to adapt their services to the project planning. Among all geotechnical tasks leading to the definition of soil design parameters, the laboratory testing phase can hardly be shortened.This phase, which often lasts a few months for deep water projects, provides indispensable data for the determination of final soil parameters and will still remain essential in respect to the present state of engineering practice. Recent experience has shown that the geotechnical contractor can effectively support the planning shortening and provide reliable sets of design parameters

© 2011 by Taylor & Francis Group, LLC

at short notice. An efficient data integration process however requires the in-situ campaign to be rationally planned and interactively conducted with client as explained further in this paper. In the following, the focus is placed on the determination of shear strength parameters which are considered as most critical for the geotechnical design of seabed pipelines and caisson type foundation structures. The proposed procedure is illustrated through examples coming from a site investigation offshore West Africa where the client imposed a fast-track delivery of soil design parameters for preliminary foundation design. The soil investigation program was tailored in close cooperation with the client in order to be able to provide the engineering team with accurate preliminary estimates of undrained shear strength profiles within a time frame of two weeks after completion of the field work. Deep water areas affected by geohazards which are likely to significantly modify the stratigraphic environment and therefore the behaviour of the sediments are not addressed in this paper. 2

CAISSON FOUNDATIONS

Suction caisson foundations are extensively used in the development of deep water hydrocarbon reservoirs for anchoring floating structures or riser towers and for supporting seabed structures, e.g. well manifolds, PLETs or PLEMs. Embedment ratios (embedded length to diameter) of the caissons vary from 0.5 to 5 with maximum penetrations of 20 to 30 m. The data required for the geotechnical design of these caissons necessitates investigations of typically 25 to

253

35 m penetration. Geotechnical engineering of caisson foundations is based on a precise knowledge of key soil properties: – undrained shear strength and resistance anisotropy for capacity calculations, – sensitivity to remoulding due to penetration of the caisson’s wall, – shear strength degradation under cycling loading, – thixotropy effect for evaluation of regain in capacity with time after installation. 2.1

Present practice of geotechnical site investigations for caisson type foundations

– cyclic triaxial and cyclic DSS tests; – thixotropy tests. These laboratory tests provide specific parameters directly used in the most recent design methods as recommended in engineering codes and standards of the offshore industry. The monotonic undrained shear strengths measured through laboratory testing also provide references for calibrating the N-factors needed to transform resistance measurements of CPT, T-bar or Ball probe penetrometers into undrained shear strengths. 2.2

Geotechnical investigations are generally based on an association of in-situ testing and high quality sampling. Continuous penetrometer tests are conducted with seabed modules down to 20–40 m penetration. The Cone Penetration Test (CPTU) technique is the most extensively used whereas full flow penetrometers (T-bar and Ball probe) have been introduced and are recognized to offer great potential. Penetrometer measurements allow continuous undrained shear strength profiling and identification of any heterogeneity which could alter the capacity of the foundation or compromise its installation. In addition to continuous measurements, discrete measurements of in-situ undrained shear strength can be carried out using the in-situ vane (VST). Samples of the highest possible quality standard are required for advanced laboratory testing. Downhole push sampling and/or piston gravity coring can be performed for that purpose but the preference often goes to truly stationary piston corers due the very effective times of operation. The STACOR®piston sampler provides continuous large samples of 100 mm diameter and 20 m length in average. It has been extensively used offshore West Africa for deep water investigations. Quality of STACOR®samples, evaluated with disturbance index e/eo and Lunne et al. (1998) disturbance criterion, proved that these samples are of class 1 (very good to excellent) or 2 (good to fair) (Borel et al., 2002). Offshore laboratory testing is generally limited to basic identification tests due to the very soft and sensitive nature of the material. All advanced laboratory tests are performed onshore and a typical laboratory program includes:

The nature and quantity of the tests to be performed often represent a relatively long laboratory testing period of several months, which is sometimes hardly compatible with project development planning. In this case a good appreciation of the main engineering parameters can be obtained shortly after the geotechnical campaign, provided the scope of work has been appropriately developed. The proposed site investigation strategy to obtain fast-track engineering parameters is based on the following:

– identification testing; – multi-sensors core logging and/or X-ray radiography; – intact and remolded laboratory vane tests and unconsolidated undrained triaxial tests for sensitivity assessment; – Ko consolidated triaxial tests; – consolidated triaxial tests in extension (CAUE) and compression (CAUC) and direct simple shear tests (DSS) for determination of undrained shear strength in extension, compression and shearing modes; © 2011 by Taylor & Francis Group, LLC

Proposed site investigation strategy for fast-track design

254

– investigate all preselected locations with either one single CPT test (30 m or deeper) or one CPT plus one high quality piston sampling and assess site variability; – based on above results, select a limited number of locations where “clusters” of in-situ tests are performed including one CPT, typically to 35–40 m, one Vane Shear Test (to 25 m or deeper) and one T-bar test with cyclic measurements; – perform one high quality sampling down to at least 15–20 m at the same “cluster” locations; – all in-situ measurements at the “cluster” locations are cross-correlated and respective Nk factors (Nk CPT/VST and Nk T-bar/VST ) are calculated, taking the VST as reference; – the above Nk factors are used to derive continuous undrained shear strength profiles at all other locations with single CPT. The application of the procedure is illustrated below. Figure 1 presents the results of 13 CPTs carried out on a relatively homogeneous site in the Gulf of Guinea, offshore West Africa. The variation of the net cone tip resistance qn can be considered as linearly increasing with depth below 4 m of penetration. A statistical analysis discloses a standard variation of 0.308 MPa which can be considered as a good estimation of the natural variability of the resistance over the whole site. Six in-situ vane tests have also been carried out across the same site at close proximity to CPT locations (Figure 2). The standard deviation of SuVST over the same range of penetration (4 to 25 m) is equal to 2.25 kPa. The results of both types of in-situ tests have been correlated at the cluster locations and the NkVST

Figure 1. CPT test results (13 locations) and statistical parameters.

Figure 3. Correlation between in-situ directly measured (VST) and CPT / T-bar deduced undrained shear strengths.

Figure 2. In-situ vane test results and statistical parameters.

Figure 4. Example of cyclic T-bar results in GoG clay. a) qt versus N cycles – b) sensitivity versus Ncycles.

factors found have been used to deduce a continuous undrained shear strength profile from all the CPTs across the site. The standard deviation of the resulting Su profiles is equal to 2.49 kPa which is remarkably close to the value calculated from the in-situ VSTs only (Figure 3). A similar approach could equally be applied with T-bar or Ball probe tests instead of CPT tests. Full flow penetrometers present the advantage of providing (in addition to the static resistance) quick access to residual strength parameters and rate of loading effects through dedicated cyclic tests. Figure 4 presents the result of a cyclic T-bar test performed at 5.5 m depth on the same Gulf of Guinea site. The sensitivity is equal to 4 after 10 cycles but continues to increase to reach 6 after 25 cycles. This relatively high value was confirmed by installation © 2011 by Taylor & Francis Group, LLC

255

analyses (Colliat et al., 2010). Puech et al (2010) recommend assessing the sensitivity in Golf Guinea clay by using cyclic T-bar resistance after 30 cycles. At this stage of the interpretation the proposed shear strength profiles are all referenced to VST data whereas preliminary design is usually based on:

where subscripts indicate mode of shearing (C: compression; E: extension; SS: direct simple shear; av: average). The final correspondence between SuVST and SuSS as well as the anisotropy factors will of course be fully known only after completion of the laboratory tests. However good estimates can be anticipated, particularly when similar investigations have been carried out

Figure 6. Mini T-bar testing with Fugro DECKSCOUT™ system in box corer sample. Figure 5. Final SuE , SuSS , and SuC values compared to SuVST .

Complex soil-pipe interaction models have been elaborated based on in-situ tests of large pipeline sections or on small scale laboratory tests. Determining the required parameters and calibrating the model for each particular situation remains a very challenging exercise. It is now agreed that soil-pipe interaction laws in very soft soils should be better obtained directly on the sea-bottom.

previously in the same area. In the Gulf of Guinea, the factor between SuVST and SuSS is of the order of 0.75–0.8, correlating well with the rate of loading effect generated in these clays by the different modes of shearing. The final profiles of SuE , SuSS and SuC values are given in Figure 5. It must be highlighted that, for this case study, the final value of the DSS profile was only 5% different from the one provided two weeks after site work completion. In conclusion, the proposed site investigation procedure makes it possible to provide the engineering team shortly after completion of field work with i) general trends of undrained shear strength across the whole site with good estimate of the natural variability, ii) accurate and well calibrated undrained shear strength profiles at each selected location, iii) reliable intact and remoulded design shear strength values for preliminary suction caisson design.

3 3.1

3.2

Piston gravity corers have been extensively used for pipeline soil investigations in deep water. The analysis of the top decimetres of samples however often discloses partial loss or remoulding of the material which cannot be used for highly accurate measurements at very low stresses. Until recently, the box corer was the best option for recovering fully intact seabed samples. Geotechnical engineers are however confronted with a major issue once the sample, with typical dimensions of 0.5×0.5×0.5 m, is recovered on deck. The extremely soft nature of the material prevents any sub-sampling without disturbing the clay. Undrained shear strengths cannot therefore be measured with usual laboratory devices. The most effective solution consists of testing the samples directly onboard with in-situ type equipments (Low & Randolph, 2008). Fugro developed a mini T-bar system (called DECKSCOUT™) mounted on a specially designed frame to test the box core sample on deck (Figure 6) immediately after recovery (Puech et al., 2010). A similar set up can also be used with laboratory vane equipment.

SEABED PIPELINES/FLOWLINES Required parameters

Recent experiments confirm that in very soft soils, the longitudinal and lateral restraints of pipelines/ flowlines laid on the sea bottom cannot be modelled by simple friction coefficients but are highly dependent on the penetration of the pipeline into the sediment. This penetration is not only driven by the submerged weight of the pipeline but also by installation conditions (vertical overstressing, lateral sweeping). An accurate prediction of the pipeline response under cycles of pressure and thermal expansion forces requires both: – a very accurate measurement of the undrained shear strength over the first decimetres (accuracy of the order of or better than 1 kPa), – a precise soil-pipe interaction model. © 2011 by Taylor & Francis Group, LLC

Existing methods

3.3

New developments

Two complementary tools were recently introduced to the market by Fugro to address the challenging issue of pipe-soil interaction assessment. The SMARTPIPE® tool is aimed at measuring directly in-situ pipe-soil interactions. The equipment has been described in detail by Hill and Wintgens

256

Figure 7. SMARTPIPE® result example.

(2009). It is equipped with an instrumented pipe section and a 2 m stroke standard T-bar. Tests are performed at selected locations. These locations should be representative of the soils encountered over the entire flowline network. The first survey was conducted from a prototype tool in May 2008 and disclosed very positive outcomes, particularly for the vertical and axial behaviour. Some necessary changes were highlighted. Re-design of the launching system and of the tool itself were conducted and a second survey was successfully performed by the end of 2009. A preliminary example of the results acquired during the last campaign is presented in Figure 7. The SMARTPIPE® will necessarily be used at a limited number of discrete locations to obtain detailed information on the in-situ soil-pipe behaviour. Accurate information on surface soil properties will be required at a much larger number of locations regularly distributed all along the pipeline routes to assess that the chosen test locations are representative of the whole route. It is important that soil conditions are investigated with the same quality standard at the testing locations and all along the flowline network to extrapolate results with confidence. The SMARTSURF™ seabed module has been designed in response to this challenging issue. The frame and launching system are identical to the SMARTPIPE® . The SMARTSURF™ seabed module (Figure 8) is equipped with: – a 3 m stroke standard T-bar or CPT, – a 1 m stroke mini T-bar (L = 75 mm; D = 12 mm) – a pushed stationary piston sampler specifically designed to allow 2 m stroke cores of soft undisturbed sediment to be recovered. The module can be operated to a maximum water depth of 3500 m. During its development, special attention has been given to the control of the module with regard to settlement in the seabed and to the identification of the water / soil interface. The module is equipped with a seabed detector sensor and with a camera which allow a perfect control of the © 2011 by Taylor & Francis Group, LLC

Figure 8. SMARTSURF™ module.

soil interface during the penetration of the in-situ and sampling tools. Another challenge for the development of the module was the piston push sampling system. An innovative active core catcher and pressure monitored piston were implemented. The corer is smoothly lowered close to the seabed and a pressure gauge below the piston can detect the presence of mud, which triggers the penetration of the corer.The pressure is then monitored during the penetration in order to prevent possible depression and resultant remoulding of the top of the sample. The first survey with the SMARTSURFTM module has been successfully conducted in 2009. Intact and remoulded undrained shear strength profiles along the pipeline routes are obtained through Nk factors applied to qn (CPT) or qT (T-bar). T-bar is normally preferred because of i) its better accuracy in extremely soft materials ii) the quasi insensitivity of Nt to the overburden pressure and iii) the capacity to perform cyclic tests. However over the very first decimetres of penetration results are perturbed by surface effects. This can be mitigated by using mini T-bars provided appropriate testing and interpretation procedures are applied (Puech et al., 2010).

3.4

Proposed investigation strategy

The SMARTPIPE® and SMARTSURF™ can be operated with the same Launch and Recovery System (LARS), allowing safe handling from any vessel of opportunity fitted with a convenient A-frame. Specific design features have been introduced to eliminate any risk of soil remoulding at impact and to guarantee a proper assessment of the seabed level. These features include i) large deployable bearing plates, ii)sensors for mudline determination, and iii) increased clearance between tool base and mudline).

257

For soil-pipe interaction assessment, the combined use of the newly developed SMARTPIPE® and SMARTSURF™ modules is certainly the best way to obtain fast and highly reliable data for direct use in engineering procedures. In both cases close coordination with the client during the preparation phase and flexibility during operations are keys to success. Laboratory testing will remain indispensable to confirm and refine the preliminary set of parameters obtained with the proposed methodology and to complement it with specific parameters (e.g. anisotropy, cyclic behaviour, thixotropy,).

The SMARTPIPE®and the SMARTSURF™ can also be deployed from a specialist drilling vessel. The proposed strategy to obtain fast and reliable design parameters for soil-pipe interaction assessment is entirely based on in-situ testing: – first the SMARTSURF™ module is used to investigate all preselected locations (number depending on network extension and type of seabed structures). At each location T-bar and mini T-bar testing should be performed. Cyclic tests with standard Tbar should be performed below 1 m and at about 0.3 and 0.7 m with mini T-bar. Samples can also be retrieved for subsequent lab testing. Once these data are available, site homogeneity can be assessed and number and final location of in-situ soil-pipe interaction measurements can be chosen. – the SMARTPIPE® module is then operated at the small number of selected locations. In addition to all vertical, axial and lateral instrumented pipe tests, T-bar testing should be performed to allow correlations with corresponding SMARTSURF™ data and confident extrapolation of soil-pipe interaction laws to the entire network. 4

CONCLUSION

Geotechnical contractors are subject to a pressing demand of the industry to compress the time required to produce engineering parameters following a site investigation. This paper explains how it is possible under certain circumstances to deliver a preliminary but reliable set of geotechnical design parameters shortly (typically two weeks) after completion of field activities, i.e. without waiting for laboratory testing results. For caisson type foundations, an efficient procedure consists of i) assessing the site variability based on CPT data, ii) performing “clusters” of in-situ tests at selected locations (including CPT, T-bar and VST), iii) correlating data and building a undrained shear strength model with reference to VST data, iv) propagate the model to other locations where only CPT are available. Moving from VST to DSS data can be confidently made using past regional experience and/or measurement of rate of loading effects from cyclic T-bar testing.

© 2011 by Taylor & Francis Group, LLC

REFERENCES Borel, D., Puech, A., Dendani, H. & de Ruijter, M. 2002. High quality sampling for deepwater engineering: the Stacor®experience. Ultra deep engineering and technology (UDET), Brest, France. Colliat, J.L., Dendani, H., Puech, A. & Nauroy, J.F. 2010., Geotechnical properties of Gulf of Guinea deepwater sediments, Proc. of the 2nd Intern. Symposium on Frontiers in Offshore Geotechnics, Univ. of Western Australia, Perth, 8–10 November, 2010. Hill, A.,J. & Wintgens J-F. 2009. In-situ measurement of pipesoil interaction in deep water – Results of a successful offshore campaign. Proc. of the SUT Annual Conference, Perth, Western Australia, 2009. Kolk H.J. & Wegerif J. 2005. Offshore site investigations: New frontiers.Proc. of the 1st Intern. Symposium on Frontiers in Offshore Geotechnics, Univ. of Western Australia, Perth, 19–21 September, 2005. Low, H.E. & Randolph, M.F. 2008. Characterization of near seabed surface sediment. Proc. of Offshore Technology Conference, Houston, Texas, USA, 2008. Lunne, T., Berre, T.V. & Strandvik, S. 1998. Sample disturbance effects in deep water soil investigations. Proc. Int. Conference on New Frontiers in Offshore Site Investigation and Foundation Behaviour, SUT, London. Puech, A., Orozco-Calderon, M. & Foray, P. 2010. Mini Tbar testing at shallow penetration, Proc. of the 2nd Intern. Symposium on Frontiers in Offshore Geotechnics, Univ. of Western Australia, Perth, 8–10 November, 2010.

258

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Enhancement of the ball penetrometer test with pore pressure measurements N. Boylan & M.F. Randolph Centre for Offshore Foundation Systems (COFS), University of Western Australia

H.E. Low Benthic Geotech Pty Ltd. (formerly of COFS)

ABSTRACT: The spherical ball penetrometer is a full flow penetrometer which was developed to overcome some of the shortcomings of the traditional cone penetrometer, particularly in soft offshore sediments. The addition of pore pressure measurements to this test will allow for the development of soil classification charts, dissipation tests to obtain consolidation parameters and further insight into the soil degradation process during cyclic penetrometer tests. To date, a number of ball penetrometers with the facility for pore pressure measurement, or piezoballs, have been developed. However, the various piezoballs have used different pore pressure filter locations and experience of pore pressure measurements is still very limited. Research is required to develop relationships with geotechnical parameters utilizing the measurement of excess pore pressure in this test. This paper describes recent research conducted at COFS using a piezoball which measures the pore pressure at two locations (tip and mid-height positions). The magnitude of excess pore pressure generation with filter location is discussed using the results of field tests conducted in a soft silty clay deposit and comparisons are made to the piezocone test.

1

INTRODUCTION

Full flow penetrometers were developed initially from the miniature T-bar used in the centrifuge to improve the measured resistance over cone penetrometers (Stewart & Randolph 1991) and was later scaled up for use in offshore site investigations (Randolph et al. 1998). The ball penetrometer was then developed to reduce the chance of the load cell being subjected to bending moments induced from non-symmetric resistances along the T-bar (Watson et al. 1998). These devices have several advantages over the standard cone penetrometer (Randolph 2004): – improved accuracy in soft soils due to the larger projected area, which results in improved resolution and reduced sensitivity to any load cell drift and temperature effects; – minimal correction for overburden and pore pressure effects; – theoretical solutions available for deducing the shear strength from the penetration resistance; – remoulded shear strength assessable from cyclic penetration and extraction of the full flow penetrometer. A recent development, particularly for the ball penetrometer, has involved fitting pore pressure sensor(s) to obtain parameters in addition to the penetration resistance, thus enhancing the capability of full flow penetrometers for estimating geotechnical parameters © 2011 by Taylor & Francis Group, LLC

259

other than undrained shear strength. Kelleher and Randolph (2005) and Peuchen et al. (2005) showed the excellent potential of the ball and T-bar penetrometer with pore pressure measurement for assessing soil stratigraphy. In their tests, Kelleher and Randolph (2005) measured pore pressure at the mid-height of the ball while Peuchen et al. (2005) measured the pore pressure along the axis of the T-bar (with one sensor at the centre and one at the edge) and at the tip of the ball. In the characterisation of peaty soil, Boylan and Long (2006) also showed that the pore pressure data measured at a location of one third the ball diameter from the tip of the ball appeared to be useful in identifying the relative decomposition within a peat deposit. Low et al. (2007) and DeJong et al. (2008) showed that, as for the piezocone, piezoball dissipation tests (with pore pressure measurement at the mid-height of the ball) may be used to estimate the consolidation parameters of a soil. However, experience of pore pressure measurement during ball penetration tests is still very limited and further research is required to develop soil behavioural classification charts, interpret consolidation parameters from dissipation tests and develop relationships with geotechnical parameters other than shear strength. This paper presents a piezoball that measures the pore pressure simultaneously at both the equator and the tip of the penetrometer. Results of testing with this piezoball are examined with particular emphasis on the trends of the excess pore water pressure.

Burswood research site (Low et al, in press) which has been used as a soft clay research site by researchers at UWA since 1989. This site is similar to the Burswood site and is underlain by approximately 20 m of soft clay, deposited in an estuarine environment. The specific testing location is underlain by 1.7 m of fill material which overlies silty clay to a depth of 20.7 m. Shells are frequent in the profile, particularly above 5 m and reduce with further depth. Laminations are particularly evident below 14.7 m. The soil is of high plasticity, with a natural moisture content (wn ) decreasing from 114% at 4 m to 63% at 17.3 m. The unit weight (γ) increases from 13.6 kN/m3 to 15.6 kN/m3 over the same depth interval. The water table is high and sits approximately at the surface.

Figure 1. UWA piezoball penetrometer.

2

UWA PIEZOBALL PENETROMETER

3.2 Testing procedures

For this research, a piezoball penetrometer was developed in-house at The University of Western Australia (UWA) (See Fig. 1). The ball has a diameter of 60 mm and is connected to a reduced shaft section of 20 mm diameter which results in shaft to penetrometer area ratio (As /Ap ) of 0.11. The reduced shaft section is 185 mm long before tapering out to the standard rod diameter of 35.7 mm. This reduced shaft length is to minimize the migration of the excess pore pressures generated by the taper to the pore pressure field around the ball. On the ball, the pore pressure is measured simultaneously at both the tip (utip ) and the mid-height (um ), or equator of the ball. At the tip position there is a single pore pressure filter while at the mid-height there are four filter position located at 90◦ to each other. At each filter position, the pore pressure is measured by miniature total stress transducers (Kyowa PS-10KD) which are located in a recess close to the surface of the ball. The recess allows for a small fluid filled cavity covered by a porous high-density polyethylene (HDPE) filter which is flush with the surface of the ball. The locating of the sensor close to the surface of the ball is to minimize problems saturating the ball and ensures fast response times at the pore pressure filter locations. The penetration and extraction resistance of the ball is measured by a load cell located directly behind the ball within the reduced shaft section. The load cell is a strain gauged section, fitted with a double full bridge of strain gauges and includes temperature compensation. The ambient temperature is also measured by a thermistor located alongside the load cell. The inclination of the ball during testing is monitored by an inclinometer within the standard diameter shaft section above the ball.

3 3.1

To aid penetration through the fill material, each test location was predrilled to a depth of 2 m before testing. Testing was carried out using the piezoball described in Section 2 and a 10 cm2 piezocone. Prior to each piezocone test, the cone was saturated by attaching an inverted funnel reservoir to the penetrometer and filling with glycerine. Ensuring the pore pressure cavity was free of air bubbles, the filter element was then put in place, followed by the cone head and a latex membrane was placed over the pore pressure filter. For the piezoball, the ball was saturated by immersing the ball in a container of glycerine. A syringe was used to ensure no air bubbles remained in the fluid cavities and the filters were then inserted. A latex membrane was placed over the ball to maintain saturation prior to each test commencing. At the beginning of each test, the penetrometer was lowered into the borehole and allowed to equilibrate in the ground water. When the sensor readings reached equilibrium, zero readings were taken and the test commenced. Penetration and extraction were conducted at the standard rate of 20 mm/sec throughout. During each piezoball test, at least one cyclic penetrometer test was carried out to evaluate the symmetry of penetration and extraction resistance and allow the data to be re-zeroed if necessary. At the end of each test, zero readings of all sensors were again taken for checking against zero readings taken at the beginning.

SITE & TESTING PROCEDURES

4 4.1

RESULTS Data interpretation

For the piezocone tests, the net resistance (qnet ) was calculated by correcting the measured cone resistance (qc ) for unequal pore pressure effects and overburden resistance using (Lunne et al. 1997):

Site

Testing was carried out at a site on the northern bank of the Swan River close to the centre of Perth, Western Australia. This site is approximately 3 km from the © 2011 by Taylor & Francis Group, LLC

260

where α = unequal area ratio of the cone (0.87 in this case) and u2 = excess pore pressure measured at the shoulder position behind the cone. The pore pressure

Figure 2. Typical piezocone result (a) Net cone resistance (b) Excess pore pressure (c) Pore pressure parameter.

parameter from piezocone tests (Bq ) was calculated from:

where u0 = ambient pore pressure. For the piezoball tests, the net ball resistance (qball ) was calculated by correcting the measured resistance (qm ) using the following expression (Chung & Randolph 2004):

where α = unequal area ratio of the piezoball (0.85 in this case). The pore pressure parameter from piezoball tests (Bball) was calculated for the pore pressure measured at the mid-height (um) and the pore pressure measured at the tip (utip) using the expressions shown in Equations 4 and 5 respectively:

For the mid-height pore pressure, um represents the average pore pressure measured by the four pore pressure sensors. 4.2 Monotonic piezocone tests Figure 2 shows the result of a typical piezocone test conducted at the testing location, in terms of the net cone resistance (qnet ), excess pore pressure (u) and the pore pressure parameter (Bq ). The net resistance initially reduces with depth above 5 m, before © 2011 by Taylor & Francis Group, LLC

increasing with depth at an average rate of rate 20–30 kPa/m. Excess pore pressures follow a similar trend to qnet , with values reducing above 6 m before increasing at a rate of 15 to 22 kPa/m. Bq values are relatively uniform and generally lie between 0.4 to 0.5. 4.3

Monotonic piezoball tests

Figure 3 shows the result of a typical piezoball test conducted at the test location, shown in terms of the net ball resistance (qball ), the excess pore pressures at the mid-height and tip (um & utip ) and the pore pressure parameter at these positions (Bball-m & Bball-tip ). The net ball resistance initially reduces above 6 m before increasing with depth at an average rate of 15 kPa/m. During penetration, the excess pore pressure at the mid-height position (um ) is relatively uniform above 12 m and is typically less than 50 kPa. Below 12 m, a slight increase with depth is evident. The corresponding pore pressure parameter (Bball-m ) ranges from about 0 to 0.2. At the tip position, the excess pore pressure (utip ) follows a similar pattern to the net ball resistance (qball ) in that it is reducing above 6 m, before increasing with depth at a rate of 20–30 kPa/m. The corresponding pore pressure parameter (Bball-tip ) ranges from 0.8 to 1.1. Comparing the mid-height and tip of the ball, the pore pressure parameter at the tip is on average 4 to 11 times higher than the mid-height position. The difference between the two filter locations reflects the nature of the stress changes in the soil at each location. Beneath the tip, the soil is undergoing large compressive stresses while at the mid-height the soil is being subjected to shear stresses. In a saturated soil, compression will yield positive pore pressures while an increase in shear stresses may induce either positive or negative pore pressures depending on the dilatancy properties of the soil and the mobilized shear stress level. In fine grained soils, contractive behaviour is expected due to shear, resulting in positive excess pore pressures. Compared to the piezocone, Bball-m is

261

Figure 3. Typical piezoball result (a) Net ball resistance (b) Excess pore pressures (c) Pore pressure parameters.

Figure 4. Cyclic piezoball test (a) Net ball resistance (b) Pore pressure parameter at mid-height (c) Pore pressure parameter at tip.

3 to 5 times lower than the Bq measured at the u2 position, while Bball-tip is ∼2 times higher than Bq . 4.4 Cyclic piezoball test Figure 4 shows the results of a cyclic piezoball test conducted between 16.4 and 16.8 m in terms of net ball resistance (qnet ) and excess pore pressures (um & utip ). To aid discussion about the changes in soil resistance and pore pressures during cycling, Figure 5 illustrates the values of net ball resistance and excess pore pressure midway (16.6 m) in the cyclic test region. Examination of the qnet during cycling shows that between 10 to 12 cycles of penetration and extraction were required until a relatively stable resistance was measured. After 12 cycles, the degradation in subsequent cycles is less than 1 kPa and the zero value has not been adjusted in this case. The excess pore pressure at the mid-height position (um ) reaches a stable pattern after approximately 4 cycles and during further cycles there is a constant gap between the penetration and extraction pore pressure of up to 12 kPa. The overall pattern is similar to that observed by DeJong © 2011 by Taylor & Francis Group, LLC

262

et al. (2008) for pore pressures measured at the same position. It was suggested that the differences in the penetration and extraction pore pressures that remain after the initial cycles is due to the presence of the shaft which alters the flow mechanism between penetration and extraction. Another explanation is a slight asymmetry of the excess pore pressure field due to the movement of the penetrometer. The gradient of the pore pressure during penetration and extraction has been shown to be due to the changes in mean stress which occur due to the presence of the shaft (Zhou & Randolph, in press). At the tip position, the excess pore pressure (utip ) similarly reaches a stable pattern after approximately 4 cycles. During the penetration half of a cycle, a large positive excess pore pressure develops and increases slightly with depth. During the extraction part of a cycle, the unloading of the soil beneath the ball results in a large suction developing which remains throughout the extraction half cycle. It is clear that the excess pore pressures reach a stable pattern before the penetration resistance which may reflect the nature of the remoulding taking place at different stages of the

latter decaying to a constant pattern after relatively few cycles. Further research and analysis is required to fully understand the mechanisms taking place during cycling, the development of excess pore pressures and the relationship of these to the remoulding process. ACKNOWLEDGEMENTS This work forms part of the activities of the Centre for Offshore Foundation System at UWA, which was established under the Australian Research Council’s Special Research Centre scheme and is now supported by the State Government of Western Australia through the Centre of Excellence in Science and Innovation program. The research in this paper forms part of the CSIRO Flagship Collaboration Cluster on Subsea Pipelines. The various supports are gratefully acknowledged. The authors are also grateful to the East Perth Redevelopment Authority (EPRA) and WorleyParsons for providing access to the site. Senior TechnicianAlex Duff is acknowledged for carrying out the testing on site. REFERENCES

Figure 5. Values mid cycle (a) Net ball resistance (b) Excess pore pressures.

cyclic penetrometer test. DeJong et al. (2008) hypothesized that the general degradation of the bulk soil structure may be complete within the first few cycles while in subsequent cycles the degradation is due to the remoulding of soil clusters, which may take place without further contractive or dilative volumetric tendencies which generate the excess pore pressure.

5

CONCLUSIONS

A piezoball which measures the excess pore pressures simultaneously at both the mid-height and tip of the ball during testing has been presented. Measurement of the excess pore pressures during ball penetrometer tests can lead to the development of soil classification charts, dissipation testing to assess consolidation properties and a better understanding of the soil degradation process during cyclic penetrometer tests. Results from field tests in a soft silty clay site in Perth, Western Australia have shown the magnitude of pore pressures measured at each location and the relationship of these, to measurements made at the shoulder position of the piezocone. Examination of the pore pressures during cyclic penetrometer tests reveals different levels of decay between the penetrometer resistance and the excess pore pressures, with the © 2011 by Taylor & Francis Group, LLC

263

Boylan, N. & Long, M. 2006. Characterisation of peat using full flow penetrometers. Proc. of the 4th Int. Conf. on Soft Soil Engineering. Vancouver, Canada: 403–414, Taylor and Francis. Chung, S. F. & Randolph, M. F. 2004. Penetration resistances in soft clay for different shaped penetrometers. Proc. of 2nd Int. Conf. on Geotechnical and Geophysical Site Characterization, ISC’2. Porto, 1: 671–677, Millpress. DeJong, J. T., Yafrate, N. J. & Randolph, M. F. 2008. Use of pore pressure measurements in a ball full-flow penetrometer. Proc. of 3rd Int. Conf. on Geotechnical and Geophysical Site Characterization – ISC’3. Taipei: 1269–1275. Kelleher, P. J. & Randolph, M. F. 2005. Seabed geotechnical characterisation with a ball penetrometer deployed from the Portable Remotely Operated Drill. Proc. Int. Symp. on Frontiers in Offshore Geotechnics (ISFOG). Perth: 365– 371. Low, H. E., Landon, M. M., Randolph, M. F. & DeGroot, D. J. 2009. Geotechnical Characterisation and Engineering Properties of Burswood Clay. Submitted to Géotechnique for publication. Low, H. E., Randolph, M. F. & Kelleher, P. J. 2007. Comparison of pore pressure generation and disipation from cone and ball penetrometers. Proc. of 6th Int. Conf., Society for Underwater Technology, Offshore Site Investigation and Geotechnics (SUT-OSIG). London: 547–556. Lunne, T., Robertson, P. K. & Powell, J. J. M. 1997. Cone Penetration Testing in Geotechnical Practice, Blackie Academic and Professional, London. Peuchen, J., Adrichem, J. & Hefer, P. A. 2005. Practice notes on push-in penetrometers for offshore geotechnical investigation. Proc. Int. Symp. on Frontiers in Offshore Geotechnics (ISFOG). Perth, Australia: 973–979, Taylor and Francis, London. Randolph, M. F. 2004. Characterisation of soft sediments for offshore applications. Proc. of 2nd Int. Conf. on Geotechnical and Geophysical Site Characterization, ISC’2. Porto, 1: 209–232, Millpress.

Randolph, M. F., Hefer, P. A., Geise, J. M. & Watson, P. G. 1998. Improved seabed strength profiling using T-bar penetrometer. Proc. Int. Conf. Offshore Site Investigation and Foundation Behaviour — “New Frontiers”, Society for Underwater Technology, London. 221–235. Stewart, D. P. & Randolph, M. F. 1991. A new site investigation tool for the centrifuge. Proc. Int. Conf. on Centrifuge Modelling, Centrifuge ’91. Boulder, Colarado: 531–538.

© 2011 by Taylor & Francis Group, LLC

Watson, P. G., Newson, T. A. & Randolph, M. F. 1998. Strength profiling in soft offshore soils. Proc. 1st Int. Conf. On Site Characterisation – ISC ’98. Atlanta, 2: 1389–1394. Zhou, H. & Randolph, M. F. 2009. Effect of shaft on resistance of a ball penetrometer. Submitted to Géotechnique for publication.

264

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Laboratory free falling penetrometer test into clay S.H. Chow & D.W. Airey School of Civil Engineering, University of Sydney, NSW 2006, Australia

ABSTRACT: A series of laboratory tests has been conducted to investigate the resistance of a model penetrometer free falling into normally consolidated clay. The penetrometer with different masses and end diameters has been dropped into normally consolidated kaolin with a range of undrained shear strengths at various impact velocities. The penetration depth and dynamic penetration resistance have been measured for each penetrometer drop. The effect of penetrometer mass, end diameter, soil undrained shear strength and impact velocity on penetration depth and dynamic penetration resistance have been examined. The test results show that an increase in dynamic penetration resistance occurs with increasing penetrometer mass, soil undrained shear strength and impact velocity, and with decreasing penetrometer diameter. Penetration depth was found to increase with increasing penetrometer mass, impact velocity, and decreasing diameter and soil strength. A power-law relationship has been established between these variables, enabling prediction of undrained shear strength from the measured penetration depth.

1

INTRODUCTION

Offshore in-situ site investigation is often costly due to site accessibility issues. In some situations free falling penetrometers can offer a solution. These penetrometers (retrievable or expendable) can be deployed from small vessels and are simply allowed to freefall into the seabed, with data transmitted via attached cables or wirelessly. Numerous free falling penetrometer systems have been developed in the past for nuclear waste disposal, mine sweeping and offshore site investigation purposes (Dayal & Allen 1973; Beard 1985; Denness et al. 1981; CYR 1990; Stoll et al. 2007; Stark et al. 2009). A maximum penetration depth of 9 m and impact velocities up to 27 m/s in water depths up to 5000 m have been reported by Beard (1985) using his expendable acoustic doppler penetrometer system. However, accurate interpretation of test data is hard to achieve due to the lack of an appropriate theory to predict the soil response under rapid penetration, and limited experimental data. Rate effects have been widely reported in undrained tests on clay with the soil resistance increasing with strain rate. It is often assumed that the strength increases 10% per log cycle of strain rate, however, experimental data show rates of increase varying from 3% to 150% per log cycle in laboratory and model penetrator tests conducted using varying strain rates and velocities (Dayal & Allen 1975; Biscontin & Pestana 2001; Diaz-Rodriguez & Martinez-Vasquez 2005). The large variation in the rate effect from different tests and soil types suggests some dependency on the geometry of the penetrating object (shape, size and mass) and on the clay properties (density, plasticity, strength and stress history). Attempts to examine the

© 2011 by Taylor & Francis Group, LLC

265

effects of these factors during dynamic penetration into seabed sediments and clays have been carried out using laboratory model studies (Dayal & Allen 1975; Hurst & Murdoch 1991; Allan et al. 2007; Stoll et al. 2007). These studies have shown that higher rate effects are observed for the shaft resistance than for the end bearing resistance (Dayal & Allen 1975); that there is no systematic influence of penetrometer tip geometry (Hurst & Murdoch 1991) for tip shapes that are square, 45◦ pyramids or horizontal cylinders; and that there is an increasing tip resistance with decreasing tip diameter for circular flat ended penetrometers (Allan et al. 2007). Field penetrometer studies (e.g. Denness et al. 1981; Stark et al. 2009) have tended to focus on the functionality of the various penetrometers, and have provided limited data on rate effects. This brief review indicates that further study of rate effects, and impact velocity on penetration resistance is required to improve confidence in the interpretation of free falling penetrometer tests. In this paper experiments that span a wide range of impact velocities, and investigate the influence of penetrometer geometry, including mass, in well controlled laboratory tests are presented.

2 TEST MATERIALS 2.1 Kaolin clay and clay-bed preparation Commercially available kaolin clay, Q145, which has the properties shown in Table 1, was used in this study. The kaolin clay-beds were prepared by one-dimensional compression in cylindrical confining

Table 1.

Properties of kaolin Q145.

Specific gravity Plastic limit Liquid limit Average coefficient of consolidation, cv Compression Index, Cc Friction angle, φ


2.64 27% 44% 9.05 m2 /yr 0.208 24◦

Figure 2. (a) Side elevation of model test setup; (b) A completed free falling penetrometer drop into a pot of kaolin.

Figure 1. Model free falling penetrometer.

tanks. Kaolin powder was mixed at a moisture content of 1.1 times the liquid limit and packed into the cylindrical tanks. Filter papers were provided at the top and bottom to facilitate two-way drainage, which was also assisted by a sand drainage blanket at the base of the tank. Hydraulic pressure was used to apply a uniform stress via a membrane to the top of the clay. Settlement was measured by a displacement transducer connected to a data logging system. Two cylindrical tanks with dimensions of 305 mm diameter × 370 mm high and 310 mm diameter × 335 mm high were used in the study to increase productivity. The kaolin specimens were all normally consolidated and, depending on dimensions and final stress level, required from 3 to 6 weeks to reach equilibrium with negligible change in settlement. 2.2

Free falling penetrometer

Figure 1 shows the model free falling penetrometer. The model penetrometer consists of a thin steel shaft with exchangeable tips of 5 mm thick brass circular plate in various diameters (10, 20, 30, 40 and 60 mm). Two different shaft geometries were used, with diameter × length of (5 mm × 197 mm) and (6.5 mm × 430 mm) respectively. The longer penetrometer was introduced at a later phase of testing to enable deeper penetration. Brass weights in multiples of 50 g and 100 g could be attached to the shaft as required. To improve the aerodynamic stability a fixture with four plastic fins was placed on top of the shaft. A magnetic fixture was also added on the top of the shaft to provide an automatic release mechanism from an overhead mount. 2.3 T-bar penetrometer T-bar penetrometers (Stewart & Randolph 1994) have been used in this study to determine the soil undrained © 2011 by Taylor & Francis Group, LLC

shear strength. They consist of a thin steel shaft onto which is screwed a cylindrical end of diameter, d and length, L. Two T-bar sizes were employed in the study, with d × L of 8 mm × 40 mm and 12.5 mm × 50 mm respectively. Although there is a slight difference in the aspect ratio (L/d = 5 and 4 respectively), this should not affect the estimated undrained strength (Chung et al. 2006). The T-bar was driven with velocity, v between 11 and 20 mm/s into the kaolin bed, to provide a normalised velocity, V (=vd/Cv ) greater than 10, as required to ensure undrained conditions (House et al. 2001).

3

MODEL TEST SETUP AND PROCEDURE

Figure 2a shows the laboratory free falling penetrometer test setup. The model penetrometer is initially secured to an overhead mount, which can be raised by a cable and winch system. The penetrometer has been released at different heights, up to 7.6 m, allowing freefalling into the kaolin specimens at impact velocities up to 12.20 m/s. For fall heights of up to 1.2 m the penetrometer has been instrumented with an accelerometer, mounted at the upper end of the shaft, which has been connected by thin wires to a data acquisition system. For greater fall heights the accelerometer could not be used and high speed photography coupled with PIV (particle image velocimetry) image analysis has been used to detect the displacement of the penetrometer and hence its velocity and acceleration (Chow et al. 2010). Several drops were conducted in each clay-bed, the number depending on the tip diameter, with a minimum 5 diameter spacing between adjacent drops, as shown in Figure 2b. The penetration depth was measured using a rule (up to 1 mm accuracy) at the end of each drop. Upon completion of all test drops, the undrained shear strength of the clay was measured

266

Figure 3. Typical penetrometer drop output.

using the T-bar penetrometer at two or three locations, and moisture content in the clay was measured. Figure 4. Undrained shear strength profile of kaolin.

4

RESULTS AND DISCUSSION

with diameter, d and length, L. Figure 4 shows a typical shear strength profile inferred from the T-bar. Because of the method of specimen preparation the strength is expected to be reasonably constant, and an average shear strength value has been computed for subsequent analysis.

A total of 22 pots of kaolin have been prepared and 109 penetrometer drops conducted. The following parameters have been investigated: • Penetrometer tip diameter, d (10, 20, 30, 40 and

60 mm) • Penetrometer mass, m (79 to 737 g) • Undrained shear strength, su (2.5 to 29.3 kPa) • Impact velocity, v (0.011 to 12.2 m/s)

4.2

Figure 3 shows an acceleration record from a typical penetrometer drop. The velocity and displacement were obtained by single and double integration of the acceleration with time. Upon release, the acceleration climbs quickly to the free-fall value of 9.81 m/s2 until impact with the clay when a sharp deceleration occurs. It may be noted that the penetrometer rebounds slightly after first coming to rest owing to stored elastic energy. Similar acceleration traces have been reported in free-fall penetrometer tests by Dayal & Allen (1973). Analysis of the test results has involved determination of the undrained shear strength from the T-bar test data, computation of the expected static penetration resistance (qs ), and estimation of the dynamic penetration resistance (qd ) and penetration depth (H) from the acceleration, time records. The association of the four independent variables (d, m, su and v) with qd and H has also been examined using statistical analysis and dimensional analysis. 4.1

Static penetration resistance

The static penetration resistance (qs ) has been estimated using the general bearing capacity formula of Vesic (1975) as shown in Equation 2. The formula has the advantage of allowing for penetration to diameter ratio. when z < d,

when z > d,

where d = penetrometer diameter, z = penetration depth, su = undrained shear strength, γ
= effective unit weight. In some pots slow pseudo-static tests were performed. These gave static resistances within 10% of the estimated values. 4.3

Undrained shear strength of clay

Dynamic penetration resistance

The undrained shear strength is obtained using Equation (1):

The dynamic penetration resistance (qd ) has been obtained by using Equation (3):

where F = load on theT-bar penetrometer; N = bearing factor, assumed to be 10.5 following Stewart & Randolph (1994); A = L × d the projected area of the T-bar

where F = soil resistance = mg – ma – FD ; m = penetrometer mass; g = 9.81 m/s2 ; a = acceleration measured from accelerometer; FD = hydrodynamic drag, ignored in this study; A = penetrometer tip area.

© 2011 by Taylor & Francis Group, LLC

267

Figure 6. Influence of undrained shear strength and impact velocity on qd . (d = 20 mm, m = 333 g). Table 2. Association between dynamic penetration resistance and the four independent variables. Figure 5. Dynamic penetration resistance.

Standardised independent variables

It is not possible to separate the friction and end bearing resistances as the penetration resistance is calculated from the acceleration, time record. However, as the tip area is generally much larger than the thin shaft, it is reasonable to assume that the contribution of the shaft resistance is negligible. The dynamic penetration resistance has been computed from impact until at rest. The results of two tests showing the resistance as a function of penetration are presented in Figure 5. Two identical penetrations into the same clay specimen are shown to indicate the repeatability of the data. It can be seen for these tests that the resistance and final penetration are essentially identical. In other repeat tests differences in penetration and resistance were always less than 10%. A feature of the responses in Figure 5 is the practically uniform dynamic resistance from impact until just before the final penetration depth is reached, and this pattern is evident in almost all the tests. As the penetration resistance is uniform with depth, an average value of dynamic penetration resistance has been computed for each drop for use in the further analysis below. The apparent oscillation in qd immediately after impact is believed to occur because the acceleration, from which qd is calculated, is affected by a stress wave, created by the impact, passing up and down the penetrometer shaft. The effect of impact velocity on the dynamic resistance is shown in Figure 6 for three undrained strengths with a fixed tip diameter and mass. A general trend of increasing resistance with velocity is evident. To further investigate the relationship between qd and the four influencing factors (d, m, su and v) multiple linear regression analysis using SPSS has been performed with output as shown in Table 2. Prior to the analysis, all variables were standardised by subtracting the mean and dividing by the standard deviation to provide a common scale for comparison. To confirm the association between qd and the four independent variables, the statistical significance was checked using p-values. At the 95% significance level, it is observed that all four variables produced p-values of less than 0.05, indicating strong association between © 2011 by Taylor & Francis Group, LLC

Undrained shear strength, su Impact velocity, v Diameter, d Mass, m

Beta coefficients

Significance, p

0.703

<0.00001

0.342 −0.250 0.147

<0.00001 <0.00001 <0.00001

qd and the four variables. To examine which independent variables had a greater effect on qd , standardised beta coefficients were computed. It can be seen that undrained shear strength affects qd the most, followed by impact velocity, diameter and penetrometer mass in descending order. The positive beta coefficients indicate an increasing trend in dynamic penetration resistance with increasing undrained shear strength, impact velocity and mass. On the contrary, a decrease in diameter would increase the qd . To reduce the complexity and better explore the relationship between these variables, dimensional analysis using Buckingham’s Pi theorem was conducted. Two dimensionless terms were formed, consisting of qd /su and su d3 /mv2 and their relationship is illustrated in Figure 7. The resistance might also be expected to depend on the rigidity index, G/su but as only normally consolidated clay has been tested this should be constant. A power-law equation can be fitted to the data in Figure 7 as given by Equation 4, however this has a less than satisfactory least-square value of R2 = 0.784.

268

In an attempt to examine the rate effect the ratio of dynamic to static resistance, qd /qs is plotted in Figure 8 for values of the dimensionless group su d3 /mv2 . Points are identified for various strain rate ratios, K = (v/d)/(v/d)ref , where (v/d) is the ratio of impact velocity to diameter of the free falling penetrometer and the reference ratio (v/d)ref is taken from the T-bar

Figure 9. Power-law relationship between H/d and su d3 /mv2 .

Figure 7. Power-law relationship between qd /su and su d3 /mv2 .

at the 95% significance level. It can also be seen from the beta coefficients that the penetration depth is most affected by impact velocity, followed by diameter, undrained shear strength and penetrometer mass. As expected, penetration depth is shown to increase with impact velocity and mass and decrease with diameter and undrained shear strength. The relationship between the dimensionless groups H/d and su d3 /mv2 is illustrated in Figure 9. There is good correlation between these groups (R2 = 0.959) which can be described by the power-law relationship given by Equation 5. Figure 8. Effects of loading rate on normalised dynamic resistance. Table 3. Association between penetration depth and the four independent variables.

5

Standardised independent variables

Beta coefficients

Significance, p

Impact velocity, v Diameter, d Undrained shear strength, su Mass, m

0.851 −0.768 −0.650

<0.00001 <0.00001 <0.00001

0.609

<0.00001

tests. A wide scatter can be observed with the rate factor ranging from 1.1 to 4.2 and no obvious trend. The reasons for the wide scatter are not well understood, and are the subject of further work currently underway. 4.4 Penetration depth The penetration depth was estimated from the accelerometer data, and measured directly by a ruler. The difference between the two estimations was always small and typically less than 5%. Similar to qd , the relationship between penetration depth, H and the four influencing factors (d, m, su and v) has been examined by multiple regression analysis with the output shown in Table 3. As before, strong association between penetration depth and the four variables was observed with p-values of less than 0.05 © 2011 by Taylor & Francis Group, LLC

DISCUSSION

The dimensionless group su d3 /mv2 can be interpreted as the ratio of the work done during penetration to the initial kinetic energy. It is thus not surprising that this correlates with the final penetration depth. It is however less obvious why the normalised dynamic resistance (qd /su ) should be related to this dimensionless quantity. It has been shown that the dynamic resistance increases with impact velocity, but it does not appear to be related to velocity as during penetration the velocity is reducing while the resistance remains approximately constant. If the difference between the dynamic and static resistances is considered to arise from rate effects a wide range of rate factor is implied with no detectable relation with the parameters studied. These observations indicate that factors other than those considered may be important. For example, a strain softening effect has been proposed by Einav and Randolph (2005) to explain the behaviour of cylindrical and spherical penetrometers. The data set considered has covered an order of magnitude difference in the four factors investigated and further extension may provide greater insight into the relevant dimensionless groups. For instance, it would be desirable to conduct more tests with deeper penetration beyond H/d = 12. This can be achieved by having higher fall heights or smaller penetrometer diameters.

269

To provide a better understanding of the problem, large displacement finite element dynamic analyses modelling rapid penetration of penetrometer into soil using arbitrary Lagrangian-Eulerian (ALE) method with operator-spilt technique are currently ongoing (Nazem et al. 2009). Preliminary results have indicated good matching between the numerical analysis and laboratory data. Further study is required before the application of these results in practice where the undrained strength varies with depth, the soil may not be normally consolidated and may have greater sensitivity and layers of different soils are present. 6

CONCLUSIONS

A series of laboratory model scale free falling penetrometer tests have been conducted on normally consolidated clay. Tests with a range of penetrometer geometries and fall heights have been conducted. The effects of penetrometer mass, end diameter, soil undrained shear strength and impact velocity on penetration depth and dynamic penetration resistance have been examined using statistical and dimensional analysis. The test results suggest that an increase in dynamic penetration resistance occurs with increasing undrained shear strength, impact velocity and penetrometer mass, but decreasing penetrometer diameter. The results have also indicated a unique relationship that can be used to predict undrained strength from the measured penetration. The data show that the dynamic resistance is clearly greater than the static resistance, by a factor from 1.1 to 4.2, however, the causes of the apparent rate effect could not be discerned from this study. REFERENCES Allan, R.,Airey, D. & Carter, J. 2007. Undrained strength from free falling penetrometer tests. 10th ANZ geomechanics conference, Brisbane, Australian Geomechanics Society. Beard, R. M. 1985. Expendable doppler penetrometer for deep ocean sediment measurements. Strength Testing of Marine Sediments: Laboratory and In-situ Measurements, ASTM STP 883. American Society for Testing and Materials: Philadelphia, 101–124. Biscontin, G. & Pestana, J. M. 2001. Influence of peripheral velocity on vane shear strength of an artificial clay. Geotechnical Testing Journal 24(4): 423–429.

© 2011 by Taylor & Francis Group, LLC

Chung, S. F., Randolph, M. F. & Schneider, J. A. 2006. Effect of penetration rate on penetrometer resistance in clay. Journal of Geotechnical and Geoenvironmental Engineering 132(9): 1188–1196. Chow, S. H., Nazhat, Y. & Airey, D. W. 2010. Applications of high speed photography in dynamic tests. 7th International Conference on Physical Modelling, Switzerland. In press. CYR, R. 1990. Sea Bed Sampling with an expendable acoustic penetrometer system. In D. A. Ardus & M. A. Champ (eds), Ocean Resources 2:45–56. Kluwer Academic: Netherlands. Dayal, U. & Allen, J. H. 1973. Instrumented cone penetrometer. Canadian Geotechnical Journal 10: 397–409. Dayal, U. & Allen, J. H. 1975.The effect of penetration rate on the strength of remolded clay and sand samples. Canadian Geotechnical Journal 12: 336–348. Denness, B., Berry, A., Darwell, J. & Nakamura, T. 1981. Dynamic seabed penetration. IEEE OCEANS 13: 662– 667. Diaz-Rodriguez, J. A. & Martinez-Vasquez, J. J. 2005. Strain rate behaviour of Mexico city soils. ISCMGE-2005, Osaka, Japan, 333–336. Einav, I. & Randolph, M. F. 2005. Combining upper bound and strain path methods for evaluating penetration resistance. International Journal for Numerical Methods in Engineering 63(14): 1991–2016. House, A. R., Oliveira, J. R. M. S. & Randolph, M. F. 2001. Evaluating the coefficient of consolidation using penetration tests. International Journal of Physical Modelling in Geotechnics 1(3): 17–25. Hurst, R. B. & Murdoch, S. 1991. Measurement of sediment shear strength for mine impact burial predictions. In Report of Meeting of the Technical Cooperation Program Panel GTP-13, Nov 1991. Nazem, M., Carter, J. P. & Airey, D. W. 2009. Arbitrary Lagrangian-Eulerian method for dynamic analysis of geotechnical problems. Computers and Geotechnics 36(4): 549–557. Stark, N., Hanff, H. & Kopf, A. 2009. Nimrod: a tool for rapid geotechnical characterization of surface sediments. Sea Technology 50(4): 10–14. Stewart, D. P. & Randolph, M. F. 1994. T-Bar penetration testing in soft clay. Journal of Geotechnical Engineering 120(12): 2230–2235. Stoll, D., Yue-Feng, S. & Bitte, I. 2007. Seafloor properties from penetrometer tests. IEEE Journal of Oceanic Engineering 32(1): 57–63. Vesic, A. S. 1975. Bearing capacity of shallow foundations, Handbook of Foundation Engineering, Winkerton & Fang (Eds), von Nostrand: New York.

270

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Offshore sediment overpressures: Overview of mechanisms, measurement and modeling B. Dugan Rice University, Houston, Texas, USA

T.C. Sheahan & J.M. Thibault Northeastern University, Boston, Massachusetts, USA

T.G. Evans BP Exploration, Sunbury-on-Thames, Middlesex, UK

ABSTRACT: Offshore sediment pore fluid overpressures (fluid pressure in excess of hydrostatic) contribute to potential mechanical instability of these sediments, and during offshore drilling explorations, present potential hazards to equipment and operational crews. Overpressure is likely to be found where low permeability layers have inhibited pore fluid escape. Rapid sedimentation (or disequilibrium compaction), tectonic loading and lateral transfer increase the overpressure. Aquathermal expansion, hydrocarbon generation, mineral diagenesis, organic maturation, and fluid charging increase the pore fluid volume. Dissolution and cementation can create low permeability boundaries and alter porosity. Particularly acting in combination, these mechanisms have the potential to cause near-lithostatic overpressures and potentially unstable sediments. By understanding the causal processes and the rates of overpressure genesis, potentially unstable zones can be detected and modeled more efficiently. Better measurements and models will lead to more accurate analysis of potentially unstable offshore slopes, and enable safer and more cost effective drilling practices. 1

INTRODUCTION

pore fluid overpressures. The paper is the result of a US National Science Foundation-sponsored workshop held in Oslo, Norway in March, 2009 to provide a forum for presentations and discussions about these subseafloor zones (Sheahan & DeGroot 2009). Workshop attendees identified a need for a comprehensive document detailing the current state of knowledge on the genesis, measurement and practical implications of overpressure. This paper is a brief overview of these aspects of the problem.

1.1 Importance of the problem In offshore soft sediments, where strengths are typically very low, and deposits are either still consolidating (due to high sedimentation rates) or normally consolidated, accurate measurement of the pore pressure profile is essential (Bruce & Bowers 2002). Excess in situ pore pressures (or overpressures) are believed to have been a major contributing factor to the massive Storegga submarine landslide (e.g., Solheim et al. 2005), and many other offshore regions of the world have excess pore water pressure caused by rapid sedimentation rates, mud volcano activity, and other mechanisms (e.g., Gulf of Mexico, Caspian Sea, offshore West Africa). Overpressured zones can cause major technical difficulties during site exploration investigations (e.g., Caspian Sea) and even more so during installation of offshore infrastructure, and can place the long-term stability of such structures at risk. There are many scientific, engineering and technical aspects of this issue that require a coordinated research effort. 1.2 Scope of this paper This paper provides a literature review and state of current knowledge on the genesis of offshore sediment © 2011 by Taylor & Francis Group, LLC

2 2.1

GENESIS OF OVERPRESSURE ZONES Fundamentals of overpressure behavior

The mechanisms causing overpressures and their impacts on subsequent sediment behavior are best understood using Terzaghi’s effective stress relationship:

where σ
= effective vertical stress, σ = total vertical stress = depth, u = total pore fluid pressure. Total vertical stress is defined by the integrated bulk density (ρb ) times gravity (g). Total pore fluid pressure is controlled by hydrostatic pressure from the integrated fluid density (ρw ) time gravity and the overpressure

271

Figure 1. Conceptual depiction of hydrostatic pressure, overpressure, effective stress, and overburden stress plotted against depth (after Bruce & Bowers 2002).

(u*). The total stress is referred to in the geosciences literature as the lithostatic pressure or overburden, which depends on porosity, which in turn depends on effective stresses (Osborne & Swarbrick 1997). Many models consider only the vertical stress because it is the stress component that can be readily determined. In reality, horizontal stresses also change (e.g., uniaxial strain conditions, faulting stresses, or stress block effects) and this can lead to changes in u∗ . However in the absence of horizontal stress conditions, a vertical stress-pressure relationship is typically employed. As shown in Fig. 1, overpressure zones result in deviations from hydrostatic pressure and lead to potentially significant losses in σ
. The uppermost point at which the pore pressure profile deviates from hydrostatic is referred to as the “top of overpressure.” There have been documented cases where overpressures push the u profile to near overburden stress (e.g., Dugan & Flemings 2000). 2.2

Natural mechanisms and regional overpressure

Osborne & Swarbrick (1997) divided the natural mechanisms that generate overpressure into 3 categories. First, overpressures can result from an increase in the total stress that can be caused by either rapid sedimentation (or what geoscientists refer to as disequilibrium compaction) or tectonic compression, which is typically horizontal and may be initiated by fault movement, or mud or salt diapirism. This applied total stress coupled with a mechanism preventing the escape of pore fluid, either due to the rapid rate of total stress application compared to the coefficient of consolidation or a boundary layer inhibiting fluid flow, produces overpressure. Disequilibrium compaction is thought to be the primary mechanism of overpressure in a number of basins worldwide, including the Gulf of Mexico, Caspian Sea, and North Sea (Osborne & Swarbrick 1997). This mechanism can create overpressure beginning near the seafloor and extending to depths of many kilometers. Tectonic loading can cause high overpressures because of increased horizontal stresses (Yassir

© 2011 by Taylor & Francis Group, LLC

& Addis 2001, Luo et al. 2007). Similar to rapid sediment deposition, tectonic loading yields the highest overpressures in settings dominated by fine-grained sediment. In many cases, only the vertical stress can be reliably estimated so many assessment techniques relate overpressure to the applied vertical stress as a first-order approach. Where available, the full stress field can be used to better constrain stress-induced overpressure. The second category for overpressure generation is an increase in fluid volume within the fixed pore space.As noted by Ward (1995), this can generate overpressures that may fracture the sediment. Aquathermal expansion results from temperature increases. The volume expansion is considered to be relatively small at shallow depths; in deeper sediments (>1 km) where temperatures exceed 40◦ C aquathermal effects can be more pronounced. As noted by Osborne & Swarbrick (1997), if fresh water is heated from 54.4 to 93.3◦ C, a volume increase of 1.65% results, so it is unlikely that aquathermal expansion alone would generate considerable overpressures. Diagenesis, or mineral transformation, can release large volumes of water. Two such transformations are montmorillonite (or smectite) to illite, and gypsum to anhydrite. The smectite-to-illite reaction can release bound water at high σ
and temperatures less than 200◦ C, with the first pulse of water loss occurring at less than 60◦ C (ColtonBradley 1987). However, Osborne & Swarbrick (1997) note that volume changes are less than 4%, and could not unilaterally cause significant overpressure without perfect sealing. On the other hand, the gypsum-toanhydrite transformation can result in the loss of 39% of bound water by volume at moderate temperatures (40–60◦ C). It is therefore identified as an important mechanism for overpressure generation where temperatures are high enough and gypsum is present. Fluid volume changes could also be caused by hydrocarbon generation due to organic maturation and migration from source rocks into adjacent sediments (Osborne & Swarbrick 1997, Luo & Vasseur 1996). Meissner (1978) has reported a volume expansion of 25% due to this process. Hydrocarbon generation also requires moderate to high temperatures and thus is more important at greater depths. The last category for overpressure generation covered by Osborne & Swarbrick (1997) is flow movement due to different mechanisms, such as in semiisolated permeable pathways. Head differences can drive flow toward the lowest part of the aquifer, resulting in pressures well in excess of hydrostatic at the downdip end of the aquifer. This is sometimes referred to as artesian conditions or artesian pressure. However, such a mechanism requires continuous, sealed pressure “cells” not common in the offshore environment. Osborne & Swarbrick (1997), noting overpressures in the North Sea of 41 MPa at 5 km depth, proposed other flow-based mechanisms. One is due to buoyancy of hydrocarbons (less dense than water) that rise to the top of deep aquifers, try to expand under reduced hydrostatic pressures, and thus cause increased pressures.

272

Figure 2. Flow focusing model after Yardley & Swarbrick (2000) Pressure in the sand equilibrates (i.e., PA = PB ) and is higher (thick solid line) relative to a 1D scenario (dashed line) at location A; the opposite is observed at location B.

Large differences in salt concentrations in the water could also lead to osmotic processes that move water across formations acting as semi-permeable membranes. Finally, as described by Dugan & Flemings (2000) and Yardley & Swarbrick (2000), and summarized by Bowers (2002), lateral transfer of fluids from deep overpressured zones through differentially buried aquifers can enhance overpressure in the upper reaches of those deposits where lithostatic stress is low. This results from pressure equilibration in permeable aquifers giving the aquifer a constant overpressure, reflecting the average load of the low permeability overburden, whereas the low permeability overburden has nearly a 1-D burial pressure (Flemings et al., 2002; Gibson, 1958). The resulting pressure field (Fig. 2) has overpressure in the sand exceeding that in the clay (PA ) where overburden is thin, and pressure in the sand less than in the clay where overburden is thick (PB ). This phenomenon is called flow focusing and is completely and quantitatively developed by Lupa & Flemings (2002) and Flemings et al. (2002). Flow boundaries, which prevent pressure relief, can be formed by faults, by low permeability layers, and by cementation mechanisms. Osborne & Swarbrick (1997) describe how evaporate-rich sediments could create a seal. Another chemically based means for such boundaries is gas hydrate formation. Grauls (2001) tied gas occurrence in a petroleum system to gas hydrates acting as a seal; overpressure was further enhanced by free gas accumulation beneath the gas hydrate (as previously noted regarding buoyancy effects). Flemings et al. (2003) also showed the presence of overpressure in a gas hydrate system at Blake Ridge, offshore southeast USA. A different means of flow boundary formation is capillary sealing. Osborne & Swarbrick (1997) speculated on this phenomenon, noting that mudrock with hydrocarbons present in its pores could generate such a boundary. Cathles (2001) developed a theory on the formation of capillary seals as an effective flow © 2011 by Taylor & Francis Group, LLC

273

barrier; high capillary entry pressure prevents fluids from migrating upward and facilitates the accumulation of hydrocarbons. Buoyancy effects can then lead to significant overpressure. Capillary seal efficiency is largely dependent on grain size. As overpressure increases due to buoyancy of the hydrocarbon column, flow eventually will breach the capillary seal. In coarse-grained systems, porous flow occurs when the seal is breached. In fine-grained systems, the overpressure may approach a critical value before exceeding the capillary entry pressure of the seal, which may drive hydrofracture or failure of the overburden (e.g., Xinong et al. 1999). Another potential source of fluid overpressures is the dissociation of gas hydrates arising from lowered sea level and/or an increase in temperature. The excess fluid pressures arising from hydrate dissociation has been cited as a contributory cause of submarine slides (Mienert et al. 2005); as gas is produced from hydrate dissociation and gas expands, pressures increase and effective stresses decrease, potentially initiating slope failure. While the accumulation of gas beneath seals can create overpressure, its presence can also hinder measurement. 2.3

Induced, localized overpressure

Localized excess pore fluid pressures can also result from offshore activities such as hydrocarbon exploration and production. For example, in well drilling the use of high fluid pressures to maintain hole stability can cause hydraulic fracture and swelling of the surrounding sediments (Schroeder et al. 2008). Overpressures may also develop during routine hydrocarbon production operations such as the injection of water and gas to enhance recovery and the reinjection of produced water or drilling cuttings. Heat from production wells drilled through hydrate-bearing sediments may also result in hydrate dissociation and localized formation overpressures. These induced overpressures are local in scale, but can be important factors for developing safe and efficient resource production scenarios. 3 3.1

MEASUREMENT/DETECTION OF OVERPRESSURE ZONES Penetrometers

Penetrometers have been used for rapid (<1 day) measurement of overpressure in shallow, scientific ocean drilling holes for gas hydrate systems (Dugan et al. 2003) and for shallow water flow systems (Flemings et al. 2008). These tools allow quick assessment of in situ pressure for scientific studies, geotechnical assessment and infrastructure design.These penetrometers rely on modeling the decay of pressure pulses generated during insertion to estimate in situ pressures (Long et al. 2007, Flemings et al. 2008). Data in soft, shallow clays have shown overpressures beginning within 10 m of the seafloor and increasing to 1 MPa at depths as shallow as 200 m below seafloor in

the Gulf of Mexico (Flemings et al. 2008). In a review of penetrometer data from Ormen Lange, Strout & Tjelta (2005) point out that heterogeneous sediments, overconsolidation, and uncertainty in pressure extrapolation techniques can lead to highly variable and potentially unreliable pressure estimates. In the presence of gas and water, it is still unknown what the penetrometer pressure reflects, thus making evaluation of the gas or water pressure problematic. 3.2

Long-term observatories

The Ocean Drilling Program and Integrated Ocean Drilling Program (ODP/IODP) have been involved in a series of long-term observatories, CORKs (Circulation Obviation Retrofit Kits), which consist of isolated pressure sensors in sealed boreholes (see review by Becker & Davis 2005). Most CORK studies have been used to understand the origin of overpressure due to tectonic loading, how overpressures drive flow along decollement (or fault) zones, and how overpressure may weaken faults. In addition, CORK experiments are often used to interpret formationscale permeability and compressibility based on well tests and pressure transients such as tidal loading or earthquakes. Screaton et al. (2000) used a dual CORK experiment to interpret the permeability of the Barbados accretionary complex. These permeability data have been used in numerical studies to understand the distribution of overpressure and flow pathways. Integrated pressure measurements and modeling studies have also been used to evaluate thermal and chemical anomalies in accretionary settings. In the Nankai accretionary complex, Sawyer et al. (2008) used tidal loading and multiple level pressure sensors in CORKs to estimate hydraulic diffusivity as well as to evaluate the reliability of CORK data as influenced by equipment stiffness. Offshore Norway, piezometers in boreholes have been shown as the most reliable means to measure in situ pore pressures (Strout & Tjelta 2005). While providing reliable data and adaptability in different settings for geohazard assessment, Strout & Tjelta (2005) noted that piezometers can have operational, installation, and longevity problems. This warrants future research and development on installation and use of piezometers. 3.3

Geophysical methods

The use of shear (S) and compression (P) wave seismic velocities can be used to detect and evaluate overpressure zones, but this requires a suitable velocity to enable pore pressure prediction (Sayers et al. 2002). Offshore Nile Delta, Badri et al. (2000) show that velocity-pressure transforms successfully delineate different pore pressure regimes in an active hydrocarbon province. Even with calibrated data and detailed elastic models (e.g., Lee 2003), Huffman (2002) states that serious challenges remain for geophysical methods to be effective for pressure © 2011 by Taylor & Francis Group, LLC

274

prediction; such challenges include finding appropriate velocity-pressure transforms and making reliable S-wave measurements in deepwater environments. In general, as sediments consolidate, porosity decreases and P- and S-wave velocities increase because of increases in modulus. Thus a normal velocity trend increases with depth. If overpressures develop during burial, consolidation is prevented which creates a deviation in the normal velocity trend. This may manifest as a constant velocity with depth, a decrease in velocity with depth, or a change in the ratio between P-wave and S-wave velocity. Using the deviation from the normal trend to determine overpressure magnitude is difficult, as the trend depends on many factors including burial depth, porosity, lithology, anisotropy, fluid type, and stress history (Dutta 2002, Prasad 2002, Zimmer et al. 2002). Velocitypressure transforms are best suited for analysis of primary overpressure mechanisms where the in situ effective stress is the maximum effective stress. In settings with secondary pressurization (e.g., hydrocarbon generation), velocity-pressure transforms are problematic because porosity is not recovered during unloading. Overpressure in accretionary complexes has also been interpreted from seismic velocity data; Tobin & Saffer (2009) used seismic velocity data from geophysical surveys and boreholes to interpret the pressure field at Nankai subduction zone; from velocitypressure transforms they explained the weakness of faults and extended it to interpret the extent of the seismic and aseismic zones in the region. 4

MODELING OVERPRESSURES

Numerous modeling studies have coupled sedimentation, consolidation, and fluid flow to simulate the origin and impact of offshore overpressure. Much of this work builds on one-dimensional consolidation during loading (Gibson, 1958). Early models (Bethke et al. 1988, Harrison & Summa 1991) modeled fluid flow, sediment loading, and aquathermal expansion to show regional overpressure development across large basins (100s km) over geologic timescales (106 yrs). More recently, models have shown the importance in permeability architecture (Fig. 2) on local and regional overpressure especially in the shallow subsurface (0–1000 m) where aquathermal expansion and chemical reactions are minimal pressure sources. Dugan & Flemings (2000) showed that asymmetric loading of highly permeable sand layers by low permeability clay layers can lead to high overpressure and near-zero effective stresses even where sedimentation rates are low; this results from high pressure generation where the sand is deeply buried (location B, Fig. 2) and lower pressure generation where the sand is shallowly buried (location A, Fig. 2). However, pressures equilibrate in the sand resulting in an increase in pressure in the sand at location A and a decrease in pressure at location B, relative to 1D loading. The controls on this phenomenon, termed flow focusing, are derived

for various sand geometries in 1D, 2D, and 3D by Flemings et al. (2002). Yardley & Swarbrick (2000) have shown similar relations for overpressure due to dipping sand bodies in the central North Sea. This process of flow focusing has been used to explain geohazards including shallow water flow sands and slope failure on low angle sediments (e.g., Dugan & Stigall 2010; Stigall & Dugan 2010) as well as to explain sea floor venting driven by overpressure in deep reservoirs with significant structural relief, i.e., high vertical difference between sand base and sand crest (Seldon & Flemings 2005). Modeling studies have also been completed for accretionary complexes to understand overpressure generation and distribution. While many of these models include overpressure from sediment loading and fluid production, they also include porosity loss from tectonic compression. In a conceptual model, Westbrook and Smith (1983) proposed that tectonic loading generated high overpressures within and underneath accretionary settings; this drove fluids laterally toward the toe of the prism and created mud volcanoes in front of the Barbados accretionary prism. More advanced models have simulated porosity loss from tectonic compression taking advantage of the selfsimilar growth patterns of critically tapered accretionary complexes (e.g., Bekins & Dreiss 1992). Saffer (2003) and the references therein summarize much of the numerical modeling that has been done on pressure evolution in the accretionary complexes of Nankai, Costa Rica, and Barbados. Saffer (2003) presents a sensitivity study of the key parameters that affect overpressure generation including permeability, sediment geometry, and convergence rate, and then extends the models to show how overpressure affects the critical taper angle (stable geometry of sediment wedge with continuous deformation) of systems worldwide. Other models have looked at overpressure transients and cyclic loading as a trigger mechanism for earthquakes (Byerlee 1993). 5

SUMMARY AND CONCLUSIONS

Overpressure in offshore sedimentary environments is important for understanding migration pathways of fluids (water, hydrocarbons, etc.) as well as for understanding the mechanical strength and potential hazards in different basins. Hazards associated with high overpressure include large-scale slope failures, shallow water flow sands, and well instability. Recent advancements in theoretical and numerical models have helped constrain how different pressurization mechanisms and spatial scales are critical to understanding regional and local overpressure systems. These models are now being tested through increased efforts for real-time in situ pressure evaluation and long-term monitoring. The observations test the models and provide new constraints on bulk physical properties (e.g., permeability, lithology) that allow the models to better represent the formation. Remote imaging of overpressure (e.g., geophysics) is still a growing field with © 2011 by Taylor & Francis Group, LLC

many uncertainties; but continued studies and data sets should help strengthen our remote imaging of overpressure compartments. REFERENCES Badri, M.A., Sayers, C.M., Awad, R., & Graziano, A. 2000. A feasibility study for pore-pressure prediction using seismic velocities in the offshore Nile Delta, Egypt. The Leading Edge 19:1103–1108. Becker, K., & Davis, E. 2005. A review of CORK designs and operations during the Ocean Drilling Program. Proceedings of the IODP: doi:10.2204/iodp.proc.301.104.2005. Bekins, B.A., & Dreiss, S.J. 1992. A simplified analysis of parameters controlling dewatering in accretionary prisms. Earth and Planetary Science Letters 109: 275–287. Bethke, C., Harrison, W., Upson, C., & Altaner, S. 1988. Supercomputer analysis of sedimentary basins. Science 239(4837): 261–267. Bowers, G. 2002. Detecting high overpressure: The Leading Edge 21: 174–177. Bruce, B., & Bowers, G. 2002. Pore pressure terminology. The Leading Edge 21: 170–173. Byerlee, J. 1993. Model for episodic flow of high-pressure water in fault zones before earthquakes. Geology 21: 303–306. Cathles, L.M. 2001. Capillary seals as a cause of pressure compartmentation in sedimentary basins. In: Petroleum Systems of Deep-Water Basins: Global and Gulf of Mexico Experience: Proceedings of the Gulf Coast Section Society of Economic Paleontologists and Mineralogists Foundation, 21st Annual Bob F. Perkins Research Conference 561–572. Colton-Bradley, V.A.C. 1987. Role of pressure in smectite dehydration – effects on geopressure and smectite-to-illite transition. AAGP Bulletin 71: 1414–1427. Dugan, B., & Flemings, P. 2000. Overpressure and fluid flow in the New Jersey continental slope: Implications for slope failure and cold seeps. Science 289(5477): 288–291. Dugan, B., Flemings, P., Rack, F., Bohrmann, G., Trehu, A., & Schroeder, D. 2003. Measuring pore pressure in marine sediments with penetrometers: Comparison of the piezoprobe and DVTP-P Tools in ODP Leg 204. Energy, Simulation-Training, Ocean Engineering and Instrumentation Research Papers of the Link Foundation Fellows, 3; 179–197. Dugan, B. & Stigall, J. 2010. Origin of Overpressure and Slope Failure in the Ursa Region, Northern Gulf of Mexico. In: Mosher, D.C., Shipp, C., Moscardelli, L., Chaytor, J., Baxter, C., Lee, H. and Urgeles, R. (eds), Submarine Mass Movements and Their Consequences IV; Advances in Natural and Technological Hazards Research 28, Springer, The Netherlands, 167–178. Dutta, N.C. 2002. Deepwater geohazard prediction using prestack inversion of large offset P-wave data and rock model. The Leading Edge 21: 193–198. Flemings, P., Stump, B., Finkbeiner, T., & Zoback, M. 2002. Flow focusing in overpressured sandstones: theory, observations, and applications. American Journal of Science 302(10): 827–855. Flemings, P., Liu, X., & Winters, W. 2003. Critical pressure and multiphase flow in Blake Ridge gas hydrates. Geology 31(12): 1057–1060. Flemings, P., Jackson, J., Jackson, K., Long, H., Dugan, B., Germaine, J., John, C., Behrmann, J., & Sawyer, D. 2008. Pore pressure penetrometers document high overpressure near the seafloor where multiple submarine

275

landslides have occurred on the continental slope, offshore Louisiana, Gulf of Mexico. Earth and Planetary Science Letters 269(3–4): 309–324. Gibson, R. 1958. The progress of consolidation in a clay layer increasing in thickness with time. Geotechnique 8: 171–182. Grauls, D. 2001. Gas hydrates: importance and applications in petroleum exploration. Marine and Petroleum Geology 18: 519–523. Harrison, W., & Summa, L. 1991. Paleohydrology of the Gulf of Mexico Basin. American Journal of Science 291: 109–176. Huffman, A.R. 2002. The future of pore-pressure prediction using geophysical methods. The Leading Edge 21: 199–205. Lee, M.W. 2003. Elastic properties of overpressured and unconsolidated sediments. U.S. Geological Survey Bulletin 2214: 10 pp. Long, H., Flemings, P., & Germaine, J. 2007. Interpreting in situ pressure and hydraulic properties with borehole penetrometers in ocean drilling: DVTPP and Piezoprobe deployments at southern Hydrate Ridge, offshore Oregon. Journal of Geophysical Research 112: B04101–B04118. Luo, X., & Vasseur, G. 1996. Geopressuring mechanism of organic matter cracking: numerical modeling. AAPG Bulletin 80(6): 856–874. Luo, X., Wang, Z., Zhang, L., Yang, W., & Liu, L. 2007. Overpressure generation and evolution in a compressional tectonic setting, the southern margin of Junggar Basin, northwestern China. AAPG Bulletin 91(8): 1123–1139. Lupa, J., & Flemings, P. 2002. Pressure and trap integrity in the deepwater Gulf of Mexico. The Leading Edge 21: 184–187. Meissner, F.F. 1978 Petroleum geology of the Bakken Formation, Williston basin, north Dakota and Montana. In D. Rehrig (ed.). Williston Basin Symposium Guidebook: Montana Geological Society, 207–227. Mienert, J., Vanneste, M., Bûnz, S., Andrreassen, K., Haflidason, H., & Sejrup, H.P. 2005. Ocean warming and gas hydrate stability on the mid-Norwegian margin at the Storegga Slide. Marine and Petroleum Geology 22: 233–244. Osborne, M.J., & Swarbrick, R.E. 1997. Mechanisms for generating overpressure in sedimentary basins: a reevaluation. AAPG Bulletin 81(6): 1023–1041. Prasad, M. 2002. Acoustic measurements in unconsolidated sands at low effective pressure and overpressure detection. Geophysics 67(2): 405–412. Saffer, D. 2003. Pore pressure development and progressive dewatering in underthrust sediments at the Costa Rican subduction margin: comparison with northern Barbados and Nankai. Journal of Geophysical Research 108(B5): EPM9-1 -16. Sawyer, A., Flemings, P., Elsworth, D., & Kinoshita, M. 2008. Response of submarine hydrologic monitoring instruments to formation pressure changes: theory and application to Nankai advanced CORKs. Journal of Geophysical Research 113(B1): B01102-1-16.

© 2011 by Taylor & Francis Group, LLC

Sayers, C.M., Woodward, M.J., & Bartman, R.C. 2002. Seismic pore-pressure prediction using reflection tomography and 4-C seismic data. The Leading Edge 21: 188–192. Schroeder, F.C., Jardine, R.J., Kovacevic, N., & Potts, D.M. 2008. Assessing well drilling disturbance effects on offshore foundation piles in clay. ASCE Jnl. of Geotechnical and Geoenvironmental Eng.’g 134: 1261–1271. Screaton, E., Carson, B., Davis, E., & Becker, K. 2000. Permeability of a decollement zone: results from a two-well experiment in the Barbados accretionary complex. Journal of Geophysical Research 105(B9): 21403–21410. Seldon, B. & Flemings, P.B. 2005. Reservoir pressure and seafloor venting: predicting trap integrity in a deepwater turbidite reservoir. AAPG Bulletin 89(2): 193–209. Sheahan, T.C. & DeGroot, D.J. 2009. “Seabed Sediment Pore Pressure: Genesis, Measurement and Implications for Design/Analysis,” Final Rpt., Intl. Workshop for US Natl. Sci. Fdn. Project, www.offshoregeohazards.org, 114 pp. Solheim, A., Bryn, P., Berg, K., & Mienert, J., Eds. 2005. Ormen Lange – an integrated study for the safe development of a deep-water gas field within the Storegga Slide Complex, NE Atlantic continental margin: Marine and Petroleum Geology 22(1-2): 1–318. Stigall, J. & Dugan, B. 2010. Overpressure and Earthquake Initiated Slope Failure in the Ursa Region, Northern Gulf of Mexico, Journal of Geophysical Research 115(B04101): doi:10.1029/2009JB006848. Strout, J.M. & Tjelta, T.I. 2005. In situ pore pressures: What is their significance and how can they be reliably measured? Marine and Petroleum Geology 22: 275–285. Tobin, H., & Saffer, D. 2009. Elevated fluid pressure and extreme mechanical weakness of a plate boundary thrust, Nankai Trough subduction zone. Geology 37(8): 679–682. Ward, C. 1995. Evidence for sediment unloading caused by fluid expansion overpressure-generating mechanisms: Workshop on Rock Stresses in the North Sea, Norwegian Technical Institute, Trondheim, 13–14 February: 7 pp. Westbrook, G., & Smith, M. 1983. Long decollements and mud volcanoes: evidence from the Barbados Ridge Complex for the role of high pore-fluid pressure in the development of an accretionary complex. Geology 11: 279–283. Xinong, X., Sitian, L., Weiliang, D., & Qiming, Z. 1999. Overpressure development and hydrofracturing in the Yinggehai Basin, South China Sea. Journal of Petroleum Geology 22(4): 437–454. Yardley, G.S., & Swarbrick, R.E. 2000. Lateral transfer: a source of additional overpressure? Marine and Petroleum Geology 17: 523–537. Yassir & Addis 2001. Relationships between pore pressure and stress in different tectonic settings. AAPG Memoir 76: 79–88. Zimmer, M., Prasad, M., & Mavko, G. 2002. Pressure and porosity influences on Vp -Vs ratio in unconsolidated sands, The Leading Edge, 21, 178–183.

276

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Angolan deepwater soil conditions: GIS technology development for sediment characterization P. Enjaume & M. Hamon Angolan Deepwater Consortium (DORIS Engineering), Paris, France

K. Epalanga Angolan Deepwater Consortium (Sonangol), Luanda, Angola

B. Mackenzie Fugro GeoConsulting, Wallingford, United Kingdom

ABSTRACT: The first dedicated deepwater geotechnical soil investigations in Angola were conducted in 1998 on the Kuito and Girassol fields. Since that date, geotechnical data have been gathered in a large number of deepwater sites. The Angolan GIS database represents an unprecedented gathering of factual data in Angolan Deepwater Blocks which can be viewed, accessed, queried and analysed through a map-based user-friendly interface. It helps experts gain an improved region-wide understanding of soil conditions. In the early stage of future field developments, it helps in making a preliminary assessment of soil data trends and seabed terrain features, and in planning soil investigations. 1

INTRODUCTION

geotechnical data against a background representing high resolution bathymetry and as-built facilities in the deepwater blocks. Fugro Geoconsulting Limited was chosen to construct the database and install it on the internationally well-known ArcGIS software which was customized for this project.

1.1 Background As part of a general effort to improve oil development in deepwater offshore Angola, Angola Deepwater Consortium (ADC) was formed in 2000 by Sonangol with industrial partners Doris Engineering and Pride Foramer. ADC has performed a number of studies under the guidance of a JIP Steering Committee composed of representatives and experts of BP, CABGOC, Esso Exploration Angola and Total E&P Angola (Da Costa & al., 2009). The study reported in this paper was selected by Sonangol and the Steering Committee Participants. In order to obtain the maximum benefits, the deepwater operators in Angola (BP Angola, CABGOC, Esso Exploration Angola, Total E&P Angola) set up an expert committee to agree on how to improve understanding of seabed soil conditions and their interaction with foundation structures for deepwater Angola. Angola Deepwater Consortium (ADC) was appointed for executing the agreed task and established, with the support of this expert committee, a review of the state-of-the art in deepwater site characterisation, soil-structure interaction and a review of anchoring and foundation systems. The expert committee worked also with ADC to determine how the factual data collected through all these project-specific geotechnical campaigns could be combined in a single database. The task thereafter included the development of a Geographic Information System (GIS) containing information on © 2011 by Taylor & Francis Group, LLC

1.2

Challenge

The primary objective was to compile all available factual soils information for the deepwater Angolan license blocks in order to have a regional overview of soil conditions. The database covers the Angolan licence blocks 14 to 18 and 31 to 34. The ArcGIS map interface shows a wider area so that the soils data can be viewed in the context of the Angolan shelf, including the Congo Canyon and shoreline (Figure 1). The geotechnical database contains detailed geotechnical data for over 40 soil investigation projects across the deepwater regions offshore Angola. It also shows the deepwater facilities layout of the four companies. The data are believed to be all that exist from dedicated geotechnical soil investigations performed in these deepwater areas offshore Angola. 2 2.1

MATERIAL AND METHODS GIS capabilities

The power of the GIS is that it attaches the spatial intelligence to the geotechnical data. In this

277

Figure 2. Geotechnical tables in the database.

2.3 Figure 1. Location of blocks studied.

way the data can be searched, selected and analyzed spatially. Data from geotechnical soil investigations has spatial attributes: the information can be associated with a point in space. Such data is therefore good candidate for integration into a GIS database. Data from various sources can be integrated such as bathymetry, existing geotechnical data, existing facilities, geological features. A great benefit of using GIS is data presentation. Layouts, graphs and tables can be created for use in reports and presentations in varying page sizes and formats. All of the tabular features that are viewed in the GIS User Interface can be exported in spreadsheets, for example. The user can select the layer to be used for interrogation. The results can be shown on the map using the highlight option. Engineers can also zoom to specific parts of an alignment and see the associated cross-section area, perform spatial and nonspatial queries, and print any required reports and maps. 2.2

Material

The overall database comprises an ArcGIS map user interface (MXD file), alongside a set of folders containing geodatabase tables, thematic map layers, and hyperlink files. The database does not require the use of any proprietary software other than ArcView (it was built on version 9.2) nor does it need any special expertise other than basic use of ArcView. The database can be updated, i.e. data from future soil investigations or field developments can be added to the existing database. © 2011 by Taylor & Francis Group, LLC

System architecture

At the heart of the Angolan deepwater geotechnical database there is a relational database of linked tables containing the following three layers of information (Figure 2): Projects – a project is a named set of information, grouped by geotechnical soil investigation campaign, e.g. client, year, vessel, etc. The coverage area of each project is idealised by a polygon on the map. Individual Locations – giving location information for each individual borehole, seabed sample or insitu test, giving key data associated with that location e.g. penetration, water depth, borehole/ testing method, hyperlinks to CPT files, etc. Laboratory Results – giving detailed laboratory test results – extending to advanced test types – and geotechnical descriptions versus depth for every borehole. In addition to the map layers depicting soil investigation locations, the database contains also the map layers listed below: Cultural & Licensing Data – contains fabrication yard and land base locations (complete with hyperlinked images); city and village locations; Angolan license block outlines. Facilities Data – contains layout data on offshore facilities provided by the four operating companies. The offshore facilities layouts are viewable in three ways: by status (as built, proposed, unknown), by type (subsea lines, subsea equipment, surface facilities), and by operator (BP, CABGOC, Esso, Total and Gas Gathering, for this purpose also termed an “operator”). The attributes have been standardised across all the facilities datasets and added to the database structure. Bathymetry Data and Land Topography – where available, high resolution bathymetry for Blocks 14, 15, 17, 18, 31 and 32 and high resolution land imagery were introduced to complement a basic region-wide low resolution bathymetry and land topography.

278

Geotechnical Cross Sections – there are links to pre-created image files depicting bathymetric cross sections with appropriate geotechnical data plotted on the image. Engineering Geological Desk Top Study output – bathymetry, slope and interpreted terrain unit maps have been placed in the database from a separate substudy which was part of this work and was performed by Fugro GeoConsulting Limited. Throughout the progress of the database development, customizations have been made to the standard ArcView functionality. Specifically, the database customizations are as follows: Custom Toolbar – developed specifically for use with the Angola Deepwater Geotechnical Database, this toolbar contains both standard ArcView tools as well as custom tools developed for the database. This toolbar is designed to make the database more accessible to users with little or no previous GIS experience, by providing useful tools for frequently performed operations. GIS Glossary – accessed from the custom toolbar, this window provides access to an extensive glossary of terms and abbreviations used throughout the database, making the database more accessible to users unfamiliar with the English acronyms or without specific technical knowledge of this subject. In-Situ Graph Tool – this tool allows the user to make comparative profiles of in-situ test data against depth, for a selection of in-situ test locations on the map. These graphs are created on-the-fly, and can therefore be used to make comparisons on any userdefined selection of data. Parameters plotted include measured and derived Cone Penetration Test results (cone resistance, sleeve friction, pore pressure, Net cone resistance, Total cone resistance, pore pressure ratio, friction ratio); T-bar and mini T-bar resistance. Automated Data Input Tool – the database can be updated when new data become available. There are several steps involved in adding geotechnical data, so to assist in this process key step has been automated by the development of this tool. Templates of Excel tables have been provided to operators to further facilitate integration of new data. 2.4 Geotechnical data Currently the geotechnical database contains detailed geotechnical data for over 40 soil investigation projects across the deepwater regions offshore Angola. Location of the geotechnical data is spotted against a background representing high resolution bathymetry and as-built facilities. The data are believed to be all that exist from dedicated geotechnical soil investigations performed in these deepwater areas offshore Angola. The numbers of tests (in-situ, sampling and laboratory) currently available in the GIS database are presented in Table 1. Most in-situ CPT data are also available under electronic tabulated format, accessed via a hyperlink from the map. © 2011 by Taylor & Francis Group, LLC

Table 1.

Number of tests available in the database.

Designation

Number

In situ testing (i.e. CPT, FVT and T-bar) Core samples Laboratory testing

>400 >600 >40 000

Few deep penetration borings have been performed, the majority of the available geotechnical data come from two types of investigation: • Samples from long seabed piston samplers to

approximately 20 m, • In-situ testing including: Cone Penetration Test-

ing, Field Vane Tests and T-bar tests, typically to a maximum penetration of 40 m below seabed. 2.5

Data integration

Throughout the soil investigation history of offshore Angola a range of co-ordinate and projection systems has been employed. But in order to present the data on a single map based interface, a consistent single geodetic system is required. The WGS 84 Transverse Mercator 12◦ East was selected to ensure non-negative co-ordinates for the western-most locations in the study region. Coordinate data from every project has been transformed into this chosen system (i.e. the data are fully geo-referenced) using ArcGIS basic functions.

3 APPLICATION 3.1

GIS application for analysis

As an overall capability statement, the GIS geotechnical database provides a map-based viewing environment depicting project and individual borehole/ test locations against a high resolution bathymetric and facilities layout background. It allows all underlying soils and facilities data to be queried and viewed in tabular and graphical format using the existing and wide-reaching ArcView relational database functionality. Potential applications of the GIS geotechnical database are basically endless. Using the GIS database, individual users can, for example:

279

• Establish the soil investigation history and spatial

coverage of existing soils data in the vicinity of proposed new development areas/ pipeline route corridors. • Access geotechnical data, to the level of detailed, advanced laboratory test results, on a region-wide or site-specific basis, for subsequent manipulation. • Perform attribute based searches, e.g. “show all locations where measured undrained shear strength in the top 0.5 m is more than 10 kPa”. • Perform correlations between soil parameters and other potentially influential controls such as water

depth or sampling type, hence investigating and identifying regional trends. • Using custom-written application code, generate composite plots of measured and derived CPT parameters for any user-defined selection of locations – whether within a project or region-wide. 3.2

Desktop study

Early in a deepwater project development, a geological desktop study is often performed. The purpose of this study is to create a model of the geological conditions and considerations facing a project.The main objective is to place the project in the context of its surroundings early in the design process in order to identify key parameters which could have potential impact on field architecture and facilities layout. A successful preliminary investigation may result in significant cost savings in design, fabrication and installation. The developed GIS is a very useful tool in this activity. It helps in carrying out this preliminary geotechnical analysis. 4

EXAMPLES OF APPLICATION

Installations in deepwater Angola have shown specific behaviour of the sediments when compared to those encountered on the continental shelf and in other deepwater sites in the world (e.g. Gulf of Mexico) (Puech & et al., 2005) Understanding of the mechanical properties of the soils is necessary for the design of required installations e.g. anchoring systems, pipelines etc. A global review of the geotechnical properties of the soils investigated in deepwater offshore Angola has been performed by ADC primarily in order to demonstrate capabilities of the GIS database, then to highlight the unprecedented amount of available data for deepwater Angola and tentatively confirm some trends of soil conditions. 4.1

Overview of soil properties

A first step of this preliminary analysis consisted in “picturing” all available soil parameters for all deepwater blocks together. This includes: • Index parameters: e.g. Atterberg limits, porosity,

initial void ratio • Deformation parameters: e.g. compression and swelling index, coefficient of consolidation • Strength parameters: e.g. undrained shear strength, cone resistance. Given the large amount of data available, results typically show a quite large dispersion of values and it was often difficult to derive trends for these amounts of raw data so that the analysis generally reduced to give the order of magnitude of soil parameters with its dispersion. Still, given the large amount of available data, results have been presented for three ranges of © 2011 by Taylor & Francis Group, LLC

280

Figure 3. Plasticity Index versus penetration depth for the first 20 m (2360 data points).

penetration depth which correspond to different types of foundations; they also correspond to three different sampling methods: • 0–2 m: Shallow foundations and pipeline analysis

(box coring and shallow coring), • 0–20 m: Semi-deep foundations, suction caissons,

vertical loaded anchors (giant corers like Jumbo Piston Core or STACOR), • 0–200 m: Deep foundations, driven piles (drilled soil boring) An example of the plasticity index distribution over the first 20 m is presented in Figure 3. 4.2

Correlation

Performing correlations between physical and mechanical soils properties, or between soil properties and other external influences, can efficiently contribute to the understanding of the soil behaviour. A second step of this analysis consisted in establishing correlations for Angolan deepwater sediments and their potential match with already existing correlations in the literature. Eventually more than 20 correlations have been studied, some of which are presented below. 4.2.1

Undrained shear strength versus preconsolidation pressure Mesri (1975) showed that the mobilized undrained shear strength, su , for stability analysis can be

Figure 4. Undrained shear strength versus preconsolidation pressure.

Figure 5. STACOR penetration vs water depth.

expressed, as a function the preconsolidation pressure, p
c :

Figure 4 below shows results in good agreement with such correlation. 4.2.2 STACOR penetration versus water depth Figure 5 has been plotted to confirm the trend highlighted by Borel and al. (2002) showing that the penetrations are larger in deeper waters. 4.2.3 Net cone resistance versus water depth Figure 6 has also been plotted to confirm the trend highlight by Borel and al. (2002) showing that the qnet gradients were lower for deepest sites. Most of the PCPTs have a net cone resistance gradient which quasi linearly increases with depth. However in some cases large variations of qnet could be observed especially due to geological features (pockmarks, diapirs). Only those qnet gradients which have a coefficient of determination (R2 ) as least of 0.9 have been plotted. 5

CURRENT PROJECT STATUS AND FUTURE PERSPECTIVES

This project is one of a number which have been selected by the group of offshore operators for © 2011 by Taylor & Francis Group, LLC

Figure 6. 1/qnet gradient versus water depth.

execution by ADC over the last decade. The goal of this study was to produce a complete collection of the soils data collected in the Angolan deepwater licence block areas and to present it in a format which allows a more complete view of the enlarged data set. The data set structure and the GIS system which was customized and equipped with user-friendly tools were developed by a close cooperation between the experts from the companies, the ADC study team and Fugro. This effort has produced a successful system to hold, manage and interrogate the hard-won soils data from this active offshore hydrocarbon area. The database will be used to assist in future deepwater field development in Angola. It will also be maintained and have new soils data added to it as that becomes available. The next phase of work is to transfer the database to be maintained and operated in Angola. This work is planned to be carried out by ADC using personnel from Sonangol and from Agostinho Neto University in Luanda. The expert oversight will continue to be provided by BP, CABGOC, Esso Exploration Angola and Total E&P Angola with Sonangol, they have now been joined in this role by Eni and Petrobras. Another objective set in ADC projects is to arrange for the maximum of education benefit and technology transfer to be achieved using the study activity and results. By involving Agostinho Neto University, a leading Angolan educational institution, in the follow up of the work performed by ADC and in the updating and expanding of the database, the study group aims to make strides to achieve this objective. The Participants are discussing the options of sharing the information contained in the databases with other Angolan institutions. Currently, the database contains 40 soil investigation projects and represents around 1000 information points (in-situ testing and borehole) and more than 40,000 individual measurements. The GIS database has a format that lends itself to house other data. The study database includes topographic data on the region, data on towns and fabrication facilities used by the oil industry, field facilities installed in deepwater etc. These data sets

281

can be extended to other zones and new data sets can be considered for inclusion.

ACKNOWLEDGMENTS The creation of this database is the result of exceptional cooperation between the Angolan deepwater operators: BP, CABGOC, Esso Angola and Total E&P Angola. The companies are being joined by Eni and Petrobras in the next objective, to operate and maintain this database in Angola to provide a comprehensive, updatable and evolving repository of hard-won soils data relevant to future deepwater field developments offshore Angola.

© 2011 by Taylor & Francis Group, LLC

282

REFERENCES Da Costa, R., Teixeira, J., Des Déserts, L., Harris, R.J., Burban, P.Y. 2009. The Intentions and Achievements of the Angola Deepwater Consortium. Offshore Technology Conference, Houston, Texas, OTC 19940. Borel, D., Puech, A., Dendani, H., de Ruijter, M. 2002. High quality sampling for deepwater engineering: the STACOR® experience. Ultra Deep Engineering and Technology (UDET), Brest, France. Mesri, G. 1975. New Design Procedure for Stability of Soft Clays. Discussion, Journal of the Geotechnical Engineering Division, ASCE, 101. Puech, A., Colliat, J.L., Nauroy, J.L. and Meunier, J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments. Proc. Intern. Symp. On Frontiers in Offshore Geotechnics, Perth, Australia.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Strength measurement in very soft upper seabed sediments P. Kelleher & H.E. Low Benthic Geotech Pty Ltd, Sydney, New South Wales, Australia

C. Jones University of Massachusetts, Amherst, Massachusetts, USA

T. Lunne & S. Strandvik Norwegian Geotechnical Institute, Oslo, Norway

T.I. Tjelta StatoilHydro Petroleum AS Stavanger, Norway

ABSTRACT: Accurate upper seabed strength profiling is critical for pipeline and riser design. In a recent deepwater investigation offshore Norway, a series of miniature T-bar, ball and vane shear tests were undertaken in box cores recovered from a very soft clay seabed. Seabed strength profiling was also undertaken using field scale piezocones and piezoballs deployed using a seabed founded, remotely operated drill rig. Via comparison with miniature vane data, it was shown that the miniature full-flow penetrometers can be used to estimate the continuous strength profile of box cores. In addition, it was shown that, with proper landing control of a seabed rig, a similarly accurate strength profile can be obtained in situ using field scale penetrometers. This is important because box coring is generally limited to the upper 0.5 m of seabed (which may be insufficient for pipeline design), whilst field scale penetrometers can penetrate many metres below the seabed.

1

INTRODUCTION

Precise characterization of the upper seabed is critical to the geotechnical design of a pipeline or riser structure. A key input to predicting structural performance both during initial installation as well as during operation, is the estimated bearing response of the seabed during pipeline penetration. Traditional seabed investigation equipment has been found to have significant limitations, particularly in deepwater environments where very soft, sensitive soils (with typical undrained strengths of 0 to 5 kPa) are encountered. The measured results may be affected due to: • Prior disturbance of the upper seabed soils associ-

ated with initial touchdown of seabed templates; • Effects of artificial surcharge (associated with the

application of vertical stress to the seabed in the vicinity of the intended testing area e.g. Lunne et al. 1997); • Limited sensitivity of heave compensation systems used with surface deployed drilling systems; • Poor borehole depth estimation (arising due to inadequate control, and imprecise monitoring of the position of seabed support frames, relative to the seabed). © 2011 by Taylor & Francis Group, LLC

283

A demonstration of the inability of vessel deployed systems to accurately define upper, very soft seabeds is provided in Pennington & Kelleher (2007). In response to the need to better characterize the strength profile of near seabed sediments, Randolph et al. (2007) recommended the strength of upper seabed sediments could be better characterized via miniature penetrometer testing of high quality box core samples. This paper presents a case study of a deepwater site investigation recently undertaken offshore Norway. Extensive field scale, in situ penetrometer testing and piston sampling was undertaken utilizing PROD (the Portable Remotely Operated Drill) (Offshore Engineer, 2010). The scope of work additionally included miniature penetrometer and vane shear testing of box core samples recovered from the seabed. The ability of miniature full flow penetrometers to accurately define the strength profile of box core samples is evaluated, via comparison with miniature vane test data. The penetrometer data acquired using the field scale penetrometers is additionally evaluated, in terms of their capacity to accurately emulate the results observed from within the box cores.

Figure 2. (a) PROD initial landing on leg footings; (b) Lowered and leveled for seabed operations. Figure 1. LUVA site location (taken from www.statoil. com).

2

PROJECT BACKGROUND

Benthic Geotech, developers and operators of the PROD, were contracted by StatoilHydro AS Norway to undertake a preliminary seabed site investigation at the proposed LUVA development. The LUVA development is located in the Norwegian North Sea (Fig. 1) and the water depth at the site is approximately 1,500 m. The seabed investigation was undertaken to thoroughly characterise the seabed geology at a number of proposed facility locations at LUVA, as well as along a proposed pipeline route. Seabed piston sampling, in situ piezocone/piezoball penetrometer testing and installation of piezometers were undertaken to depths of 102 m below seabed. In addition to in situ sampling and testing, box cores were recovered at the investigation locations. The box cores were subjected to miniature full-flow penetrometer (T-bar and ball), and vane shear testing onboard the vessel. The miniature instruments were deployed using a box core tester (mounted directly on the box). The box core tester was designed and built by University of Massachusetts, Amherst and was mobilised and operated by representatives from NGI (the Norwegian Geotechnical Institute).

3

achieved. Further details of the PROD footings system is described in Kelleher et al. (2008). PROD seabed operations are remotely controlled from the surface. An appropriate in situ testing tool (piezocone or piezoball) or sampling barrel (piston or rotary core) is selected from PROD’s magazines. Via the progressive makeup of drill string, the selected tool is progressively lowered to the bottom of the borehole and penetrated until refusal. At the conclusion of penetration, the drill string and the tool are recovered to the PROD magazines.

PROD EQUIPMENT

PROD is a seabed founded, remotely operated drill rig. PROD is launched from a vessel, lowered and landed on the seabed on 3 landing ‘legs’. Large soft soil footings are fitted to each leg – these are located at large radius from the centre of the rig, where sampling and testing tools are deployed (Fig. 2). Following the formation of a catenary in the control umbilical, the machine is lowered via hydraulic actuators until the central testing area of the machine is precisely level with the seabed. Via the use of displacement transducers fitted to the drill head, and monitoring of the height of the machine relative to the seabed, borehole depth accuracies of up to 25 mm (but typically better than 10 mm) are regularly © 2011 by Taylor & Francis Group, LLC

4

BOX CORING

The box corer deployed on this project was a USNL type box corer, manufactured by Oktopus, Germany. This device recovers samples of 0.5 m × 0.5 m in plan, and up to 0.5 m below the seabed. Box core samples were recovered generally within 30 m to 85 m from the PROD penetrometer tests. The device was armed on the vessel deck and then lowered with the vessel crane to the seabed. A trip release mechanism (activated at the interception of the box corer’s footing with the seabed) released the master link from the top of the box corer, enabling the box corer to plunge into the seabed. On recovery a blade was rotated beneath the underside of the box corer – thereafter the corer was recovered to deck. Once on deck, the box cores were transferred to a laboratory. The box core tester was fitted to the top of the box and miniature penetrometer and vane shear tests were performed within the soil sample. 5

PENETROMETER TESTING EQUIPMENT

5.1

Miniature full-flow penetrometers and vanes

A series of miniature full-flow penetrometers and vane shear tests were performed on each box core sample. The miniature penetrometers comprised:

284

• a T-bar, which was a cylindrical penetrometer of

7.9 mm diameter, and 49.3 mm long; • a ball penetrometer of 22.2 mm diameter;

• a cone of 6.35 mm diameter, with a tip apex angle of

on the cone tip was not directly measured. These corrections are estimated to be less than 4 kPa in terms of full-flow penetrometer resistance for the data sets presented herein. The applied corrections for push shaft friction (as a component of total mobilised penetrometer force) were observed to increase linearly, from zero at 0 mm depth, to about 20% at 450 mm depth. Motorized miniature vane shear tests were performed at test depths of 150 mm, 300 mm and 400 mm. For each test completed, the vane was pushed to target depth, held stationary for 1 minute, then rotated at a rate of between 6◦ to 12◦ per minute.

60◦ (used to estimate the contribution of push shaft friction to measured penetration resistance).

The diameter of the push shaft used to insert the penetrometers was 6.35 mm. The force mobilised on the tools was measured using a load cell fitted to the upper end of the push shaft. The motorized miniature vane shear tests were undertaken using a Wykeham Farrance vane shear apparatus. Two sizes of vane blade were used: a large vane of diameter 19.05 mm and height 38.1 mm; and a small vane of diameter 12.7 mm and height 25.4 mm. The torque applied to rotate the vane was measured using a calibrated helical spring. An anti-friction sleeve was used to reduce torque losses between the push shaft and the soil. 5.2 PROD piezocone and piezoball The PROD piezocone used to acquire field data had a projected area of nominally 1,000 mm2 and a calibrated unequal area ratio (α, Lunne et al. 1997) of 0.69. Pore pressure was measured at the shoulder of the cone (u2 position). The PROD piezoball is a 60 mm diameter, hardened steel ball, equipped with a pore pressure measurement capability at the midpoint of the ball. The ball is attached to a 20 mm diameter, 200 mm length push shaft. The push shaft is connected to a modified piezocone penetrometer of nominally 36 mm diameter. Piezoball penetration resistance was measured with the cone tip load cell. In order to eliminate load mobilized on the push shaft due to soil friction, the push shaft is isolated from the soil via the inclusion of an anti friction sleeve. Pore pressure is measured at the mid-height of the ball via a circumferential, porous filter ring.

6.2

PROD piezocone testing (CPTu) was undertaken in accordance with ASTM D5778-00 (2000). Reference readings before the start and the end of a test were taken with the probe tip located approximately 700 mm above the seabed. Tool penetration was then undertaken at a nominal rate of 20 mm/s. PROD piezoball testing (BPT) was undertaken in a similar manner to the CPTu. The piezoball was penetrated at a nominal rate of 20 mm/sec. No cyclic BPT testing was undertaken in the upper 0.5 m of seabed during this project. 7 ASSESSMENT OF PENETRATION RESISTANCES To enable a valid comparison, the miniature and field scale penetrometer data must firstly be corrected for the effects of unequal pore pressure and overburden pressure. Measured cone tip resistance was corrected to corrected cone tip resistance, qt , as follows (Lunne et al. 1997):

6 TESTING PROCEDURES

where u2 was the measured pore pressure at the cone shoulder, and α is the net area ratio. The net cone penetration resistance is then calculated as:

6.1 Box core tests All miniature penetrometer tests were performed at a displacement rate of 2 mm/s. Prior to cyclic testing, a full cycle of penetration and extraction was completed. Cyclic tests (of amplitude 50 mm) were then completed at test depths of 150 mm, 300 mm and 400 mm below the soil surface. As the force on the miniature T-bar and ball was measured above the soil surface, an estimate of the friction mobilized on the push shaft of these instruments was required. This component was estimated by inserting the miniature cone at the same locations as the full-flow penetrometers. The force data from the miniature cone was then subtracted from the full flow penetrometer data to correct for shaft friction. It is noted that limited non closure of the soil against the shaft behind the full flow penetrometers was observed over the upper (nominally) 50 mm of soil sample. Further, the force component associated with end bearing © 2011 by Taylor & Francis Group, LLC

Field scale penetrometer tests

where σv0 is the estimated total overburden stress assuming a bulk unit weight of 16 kN/m3 for the site. In a similar manner, T-bar and ball resistances were corrected for unequal pore pressure and overburden pressure effects using the expression proposed by Chung & Randolph (2004):

where qT-bar and qball are the net penetration resistances for the T-bar and ball penetrometer, respectively; qm is the measured penetration resistance; u0 is the hydrostatic water pressure; As is the cross-sectional area of the connection shaft; Ap is the projected area of the penetrometer in a plane normal to the shaft; and α is

285

Table 1. Normalised velocity (V) and ratio of penetration rate to penetrometer diameter (v/d) for penetrometer tests. Penetrometer test

V (=vde /cv(NC) )∗

v/d

Miniature T-bar test Miniature ball test PROD CPTu PROD BPT

1405 1400 22517 37843

0.25 0.09 0.56 0.33

Note ∗ cv(NC) was assumed as 1 m2 /year (NGI 1997)

the net area ratio (equal to 1.0 for all T-bar and ball penetrometers used in this project). The net remolded T-bar and ball penetration resistances are denoted as qT-bar,rem and qball,rem , respectively, in this paper. (All penetration resistances reported from hereon are quoted as net penetration resistances, unless otherwise noted). The effects of drainage and strain rates experienced by soils adjacent to a penetrometer can be important when comparing the penetration resistances measured by penetrometers of different sizes, penetrated at different rates. The effects of the soil drainage condition may be assessed using the normalized velocity, V (=vde /cv(NC) ), where v is the penetration rate; de is the diameter of a circle of equivalent projected area to that of the penetrometer; cv(NC) is the coefficient of consolidation in a normally consolidated stress state (Chung et al. 2006, Low et al. 2008). The effects of the strain rate imparted to soils around a penetrometer may be assessed using the ratio of penetration rate to diameter (v/d) (Einav & Randolph 2005, Randolph & Andersen 2008). To ensure undrained conditions prevail during testing in clay soil, Low et al. (2008) showed that V should exceed 300. V and v/d values for each penetrometer test in this project are summarized in Table 1. As indicated, the V values for all instruments well exceed a magnitude of 300. This implies that all penetrometer testing in this study was undertaken in undrained soil conditions. It may also be noted that the difference between the v/d ratio for the miniature ball penetrometer, and that for the PROD piezoball, is less than half an order of magnitude. Based on the data presented by Low et al. (2008), the differences in the measured penetration resistance between the field scale and miniature devices is estimated to be less than 5%.

8

responses increasing with depth. At the completion of remolding of the soil, however, the remolded resistances (qball,rem and qT-bar,rem , Fig. 4) more closely approximate each other. The observed differences between qball and qT-bar during installation are inconsistent with alternate data published in the public domain (Low et al. in press, Low & Randolph, in press). Low & Randolph (in press) performed a series of miniature full-flow penetrometer tests at a displacement rate of 4 mm/s in very soft clays, using miniature penetrometers of similar geometry. Their experimental data indicated that the measured values of qball should be practically identical to qT-bar during both the initial and post-remoulding penetration phases. It is suspected (though it cannot be conclusively proven) that the observed differences could be attributed to bending effects imparted to the box core tester’s load cell, as the instruments were thrust into natural soils.

9

PROD field scale penetrometer data (qnet and qball ) is also presented on Figure 3. The PROD data is found to generally plot close to the values measured using the miniature ball and T-bar penetrometers. It is further observed that despite the differences in magnitude, the depths at which significant changes in penetration resistance occur are almost exactly coincident. The presented data indicates that similar, if not better quality penetrometer data can be acquired insitu using field scale penetrometers, provided these tools are precisely deployed at the seabed, and the soils beneath the testing zone is not disturbed prior to the investigation.

PENETRATION RESISTANCES IN BOX CORE SAMPLES

Penetration resistances measured using the miniature penetrometers are compared in Figure 3. Despite the similarity in the characteristic penetration resistance profiles measured using the T-bar and the ball, it may be noted that the qball values (BC Ball) are generally lower than the qT-bar values (BC T-bar) at the same depth. The qball data plots at approximately 50% to 80% of qT-bar , with the difference between the © 2011 by Taylor & Francis Group, LLC

COMPARISON OF MINIATURE AND FIELD SCALE PENETROMETER RESULTS

286

10

COMPARISON BETWEEN PENETRATION RESISTANCE AND UNDRAINED SHEAR STRENGTH

In Figure 3, the undrained shear strengths measured from the motorized miniature vane shear tests (su,vane ) are plotted at a different scale to the penetration resistance curves, at a ratio of 1:10. It may be noted that all the measured net penetration resistance profiles reflect the variation of strength data with the depth reasonably well. In Figure 3, it may be noted that the measured su,vane values are generally plot close to the net penetration resistances measured by both miniature and field scale penetrometers. This implies that a bearing factor of 10 may be used for the first order estimation of su,vane from the measured net penetration resistances at this site. It is noted that a lower bearing factor should be adopted for sample depths of less than 0.15 m, where the effects of shallow penetration mechanisms apply (Randolph et al. 2007).

Figure 3. Measured penetration resistance profiles and undrained shear strength.

The above observations suggest that penetrometers deployed using seabed drill rigs can acquire shear strength profiles that are as reliable as those obtained from miniature penetrometer and vane shear tests on © 2011 by Taylor & Francis Group, LLC

287

box core samples. This capability is contingent on careful landing and levelling of the rig, precision depth control, and the application of appropriate bearing factors to the acquired data.

REFERENCES

Figure 4. Comparison between qT-bar,rem and qball,rem .

11

CONCLUSIONS

This paper compared two approaches to the strength profiling of very soft, near seabed deepwater sediments. Via comparison with miniature vane shear test data, it is demonstrated that miniature full-flow penetrometer testing can be used to estimate the continuous strength profile of upper seabed sediments. It is further demonstrated that equivalently accurate strength profiling is possible using field scale, in situ testing instruments, provided such tools are precisely deployed from a seabed founded drill rig, and no disturbance of the seabed occurs prior to the investigation. For many pipeline and riser installations, penetrations in excess of 0.5 m are predicted. However, the approach of characterizing near seabed soils via testing of box cores is generally limited to the upper 0.5 m of seabed. In contrast, field scale penetrometers, precisely deployed from appropriately founded seabed drill rigs, are capable of penetrating to many metres below the seabed, sufficient to enable estimation of a complete pipeline or riser bearing response within the seabed. ACKNOWLEDGEMENT The authors would like to express their thanks to Adriane G. Boscardin and Professor Don J. DeGroot of University of Massachusetts Amherst for designing, fabricating and proof testing the box core test equipment, the development of which was supported in part by the US National Science Foundation.

© 2011 by Taylor & Francis Group, LLC

ASTM-D5778-00. (2000). Standard test method for performing electronic friction cone and piezocone penetration testing of soils. ASTM International, West Conshohocken, PA, www.astm.org. Chung, S.F. & Randolph, M.F. 2004. Penetration resistance in soft clay for different shaped penetrometers. In A. Viana da Fonseca & P.W. Mayne (eds.), Proc., of 2nd Int. Conf. on Geotechnical and Geophysical Site Characterisation, Porto, 19–22 September 2004: Vol. 1, 671–677. Rotterdam: Millpress. Chung, S.F., Randolph, M.F. & Schneider, J.A. 2006. Effect of penetration rate on penetrometer resistance in clay. J. of Geotechnical and GeoEnvironmental Engineering, ASCE, 132(9), 1188–1196. Einav, I. & Randolph, M.F. 2005. Combining upper bound and strain path methods for evaluating penetration resistance. Int. J. of Numerical Methods in Engineering, 63(14), 1991–2016. Norwegian Geotechnical Institute. 1997. Ad Notam Soil Investigation 1997. NGI Report 972527-1. Kelleher, P., Samsuri, N. & Carter, J. 2008. Footing design for temporarily founded seabed drilling systems, Proc., of Offshore Technology Conference, Houston, Paper OTC 19686. Low, H.E. & Randolph, M.F. 2010. Strength measurement for near seabed surface soft soil using manually operated miniature full-flow penetrometer. J. of Geotechnical and Geoenvironmental Engineering, ASCE. (in press) Low, H.E., Randolph, M.F., DeJong, J.T. & Yafrate, N.J. 2008. Variable rate full-flow penetration tests in intact and remoulded soil. In A-B. Huang & P.W. Mayne (eds.), Proc., of 3rd International Conference on Geotechnical and Geophysical Site Characterization, Taipei, 1–4 April 2008. 1087–1092. London: Taylor & Francis Group. Low, H.E., Randolph, M.F., Lunne, T., Andersen, K.H. & Sjursen, M.A. 2010. Effect of soil characteristics on relative values of piezocone, T-bar and ball penetration resistances. Géotechnique (in press). Lunne, T., Robertson, P.K. & Powell, J.J.M. 1997. Cone penetration testing in geotechnical practice. London: Blackie Academic and Professional. Offshore Engineer. 2010. PROD probes Statoil seabed soils. Offshore Engineer, February 2010: 42–44. Pennington, D. & Kelleher, P. 2007. PROD delivers an accurate site investigation at Maari. Proc., of the 6th Int. Conf. on Offshore Site Investigation and Geotechnics Conference: Confronting New Challenges and Sharing Knowledge: 81–88. London: SUT. Randolph, M.F. & Andersen, K.H. 2006. Numerical analysis of T-bar penetration in soft clay. Int. J. of Geomechanics, 6(6): 411–420. Randolph, M.F., Low, H.E. & Zhou, H. 2007. Keynote lecture: In situ testing for design of pipeline and anchoring systems. Proc., of the 6th Int. Conf. on Offshore Site Investigation and Geotechnics Conference: Confronting New Challenges and Sharing Knowledge: 251–262. London: SUT.

288

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

CPT in polar snow – preliminary observations A.B. McCallum Scott Polar Research Institute, Cambridge, UK

A. Barwise & R. Santos Gardline Geosciences, Great Yarmouth, UK

ABSTRACT: Innovative cone penetration testing was conducted on the Brunt Ice Shelf, near Halley Base, Antarctica over the Austral Summer 2009/2010. CPT soundings were obtained from both prepared and virgin sites, and preliminary interpretation was conducted in conjunction with gravimetric density profiles. Snow stratigraphy and snow-layer mechanical properties are examined, and layer horizons tracked spatially via ground penetrating radar. The development of cone penetration testing in snow and ice may prove beneficial in the future development of polar infrastructure.

1

INTRODUCTION

Investigation of snow pack mechanical properties has long been of interest to researchers, both as a means of assessing avalanche potential, and for estimating the load bearing capacity of snow pavements. Polar snow is a sedimentary geomaterial consisting of ice and air; free water is seldom found as temperatures rarely reach 0 deg C. It is typically laid down seasonally, and undergoes metamorphism under overburden pressure and in situ temperature and vapour gradients. Ice grains increase in size with depth due to either temperature-gradient driven mass transport or vapour pressure differences (Colbeck, 1983) and compaction and sintering result in densification with depth. Snow strength is positively correlated with snow density, although variations in both snow microstructure (bonded or unbonded) and testing rate result in variation about this relationship. There are three primary methods for assessing snow strength: surface load tests, sample testing and probing, but only probing provides a time and cost effective in situ means of assessment. The use of existing geotechnical cone penetration test (CPT) equipment provides suitable robustness and capacity for conducting CPT in hard polar snow packs. The work described herein was conducted with the desire to investigate the assessment of the strength and surface bearing capacity of polar snow via rapid in situ means. 2

EQUIPMENT AND PROCEDURES

With the assistance of the British Antarctic Survey, access was provided to Halley V Research Station on the Brunt Ice Shelf, Antarctica during the summer season 2009/10 (Fig. 1). © 2011 by Taylor & Francis Group, LLC

289

Figure 1. Test location (adapted from http://www.smitha. demon.co.uk/zfids/index.htm).

An innovative CPT rig was constructed that mounts directly onto the standard three-point hitch of an agricultural tractor (Fig. 2); such vehicles are currently employed at Halley for load-hauling duties. A set of 10 kN ‘basement’ rams with a stroke of 50 cm was connected to a proprietary steel A-frame and mounted within a fabricated steel box. The standard threepoint hitch cannot counter substantial upward reaction forces, hence an additional heavy steel bar was used to connect the base of the A-frame with the upper tractorlink as necessary. This link could be easily removed to allow lifting of the ‘box’ for conduct of additional tests, and additional screw anchors can be used should

at a site were required, the rigid link was released, the box raised, and the tractor moved forward for immediate conduct of the next test. Tests were separated by a minimum of 100 mm (approximately 2½ cone diameters (L.J. Gibson, 2008, personal communication)) to minimise interference whilst still seeking spatial consistency. 3

Figure 2. CPT ‘box’ mounted on BAS tractor.

additional reaction force be necessary. This arrangement allowed rapid testing over large areas and could be mounted on whichever tractor was available. Hydraulic drive was provided using down-rated flow of 10 l/min from the tractor, adjustable via a valve on the hydraulic spool block at the rear of the tractor. Manipulation of this valve enabled penetration rates from approximately 0.2 mm/sec to 60 mm/sec to be attained. 12 volt electrical power obtained from a socket within the tractor cab provided power for the Golog logging equipment, laptop and the wire-draw depth encoding equipment. Cones used were GeoPoint 35.7 mm ‘scientific’ compression cones, modified to measure a maximum tip resistance of 20 MPa, and A. P. van den Berg’s standard “GOnsite!” software was used (A. P. van den Berg, 2002). Preliminary temperature calibration of the cones had previously been conducted in the cold rooms of the Scott Polar Research Institute (SPRI) to verify their zero-shift and linearity over the expected operational temperature range. These results suggested that although the cone temperature compensation system proved adequate at temperatures greater than 8 deg C, below this temperature the zero-shift was inversely proportional to temperature, although linearity of the applied load (N) vs output (mV) relationship was maintained. The operational consequence of this testing was that the cones were stored outside at ambient temperatures so that minimal temperature change was experienced prior to a test. A typical test involved driving the tractor to the test location before lowering the box to the snow surface where it would then settle under its self-weight of approximately one tonne. The level of the box could be adjusted to some extent by extending or retracting the top link to the tractor. After connecting the hydraulic and electrical connections, the cone was lowered to a level just above the snow surface and initial zero values recorded; the test was then conducted at the desired rate. If the rigid link was used, this was checked for tightness every 50 cm of penetration in an attempt to reduce excessive vibration or lifting of the box. Once the desired depth was reached the rods and cone were retracted and zero levels again recorded once the cone was above the surface. If additional tests © 2011 by Taylor & Francis Group, LLC

PRELIMINARY OBSERVATIONS

Almost one hundred cone penetration tests, typically to 5 m depth, were conducted at various locations within the vicinity of the Halley V Station over the period 21 Jan to 22 Feb 2010. Tests were conducted as weather and operational commitments allowed, and were designed to investigate numerous factors including: 1. 2. 3. 4. 5. 6. 7. 8. 9. 10. 11.

Local snow stratigraphy, Relationship with snow density, Relationship with other snow strength indices, Comparison with horizons determined via Ground Penetrating Radar (GPR), Effect of penetration rate, Effect of confining pressure, Existence of a compacted zone ahead of the cone, Detection of (air) pore pressure, Impact of compaction on measured resistance, ‘Sensing’ ahead of the cone, and Relationship with surface bearing capacity.

Preliminary observations on a number of these matters are described below: 3.1

Stratigraphy and density

The variation of tip resistance with snow density is evident from initial analysis. Figure 3 shows a comparison between cone tip resistance obtained at a rate of 25 mm/sec with densities obtained gravimetrically from an adjacent snow pit. Although detailed interpretation of the resistance signal has not yet been undertaken, tip resistance appears to provide both a measure of snow stratigraphy and snow density, although the coarseness of the density measurement is apparent. The relationships of both sleeve friction and friction ratio with density are yet to be examined. 3.2

Comparison with GPR

The use of GPR to define horizons in snow is becoming increasingly popular (e.g. Marshall et al., 2007). Large variations in layer density and composition, such as occurs between a weak, loose layer and an ice layer can be identified on a GPR linescan image due to variations in material dielectric properties. Initial comparison of cone tip resistance with layers identified via 400 MHz GPR (Figure 4) suggests a correlation between the two methods. In this image darker areas correspond to areas of low reflection, or limited dielectric contrast, whereas lighter areas suggest a stronger

290

Figure 3. Depth profile of tip resistance and snow density.

reflection or a higher dielectric contrast. This initial comparison highlights the promise that GPR offers as a means of rapidly determining the extent of specific layers or horizons within the snow pack over large spatial areas between point penetration tests.

Figure 4. Depth profile of tip resistance superimposed over 400 MHz GPR trace.

3.3 Effect of penetration rate Like its primary constituent ice, polar snow (a matrix of ice) is a rate sensitive material and variation in resistance was expected between tests at different rates. Figure 5 shows mean tip resistance versus penetration rate, and suggests behaviour consistent with that of ice within the brittle regime, where brittle behaviour dominates creep processes. The solid line is a power-law fit with R2 = 0.97. It is noted that actual strain rates at the cone tip may be substantially greater than the penetration rates presented. 3.4 Confinement pressure Although triaxial tests on low-density snow suggest that confinement pressure does not affect the stress/strain relationship (Scapozza and Bartelt, 2003) many other geomaterials including ice do display a dependency on confinement (Renshaw and Schulson, © 2011 by Taylor & Francis Group, LLC

291

Figure 5. Tip resistance vs penetration rate.

2001). Preliminary analysis of limited tests at the Halley V site suggests a dependency of cone tip resistance on confining pressure (or vertical effective stress) in high-density polar snow. 3.5

Formation of a ‘compacted zone’ and ‘sensing’ ahead of the cone

Previous studies using small diameter penetrometers in alpine snow have suggested the formation of a compacted zone of broken snow moving ahead of the cone tip (Floyer and Jamieson, 2006; Kinosita, 1965; Johnson, 2003), the size of this zone varying with snow type and cone shape. No compacted zone was observed to form ahead of the standard 60◦ cone in the testing conducted at Halley V. Lunne et al. (1997) and others (e.g. Houlsby et al., 1988) note the tendency of a cone to ‘sense’ an upcoming layer some 3–20 cone diameters ahead of the layer, depending on material stiffness. Tests were conducted to investigate this phenomenon, however initial analysis suggests that limited sensing occurs ahead of a cone in polar snow. This is likely due to the high modulus of the material (E = ∼1 × 106 kPa), enhanced due to grain bonding, and the brittle nature of snow, when subjected to high rates of strain. 3.6

investigations. In his review of ice bearing capacity and construction for resource extraction, Masterson (2009) argues that cone penetration testing must be accepted as a standard in situ strength testing technique for assessing the strength of ice. As the demand for substantial infrastructure in the polar areas of the world increases, the application of this technique to dense snow packs also appears timely. REFERENCES

Similarities with other geomaterials

Initial inspection of the tip resistance and friction sleeve values obtained in polar snow suggest behaviour similar to that observed in calcareous sands where precipitated bonds have formed between particles (Beringen et al., 1982). These highly compressible sands tend to have low cone resistance and high friction ratios. Cursory application of the resistance vs strength relationship proposed by Beringen et al. (1982) suggests a similar relationship may hold in polar snow. 4

CONCLUSION

This paper provides some preliminary insight into using cone penetration testing to investigate both the physical and mechanical properties of polar snow. Initial analysis suggests that factors such as snow stratigraphy, density, strength and ultimately surface bearing capacity may be derived from such

© 2011 by Taylor & Francis Group, LLC

292

Beringen, F.L., Kolk, H.J., & Windle, D., 1982. Cone Penetration and Laboratory Testing in Marine Calcareous Sediments. In: Geotechnical Properties, Behavior, and Performance of Calcareous Soils, ASTM Special Technical Publication 777, pp. 179–209. Colbeck, S.C., 1983. Theory of metamorphism of dry snow. Journal of Geophysical Research, 88(C9), pp. 5475–5482. Floyer, A.J. & Jamieson, J.B., 2006. Empirical analysis of snow deformation below penetrometer tips. Proceedings of the International Snow Science Workshop, Telluride, Colorado, pp. 555–561. Houlsby, G.T., Evans, K.M. & Sweeney, M., 1988. End Bearing Capacity of Model Piles in Layered Calcareous Soils. Proceedings of the International Conference on Calcareous Sediments, Perth, Vol. 1, pp. 209–214. Johnson, J.B., 2003. A Statistical Micromechanical Theory of Cone Penetration in Granular Materials. ERDC/CRREL TR-03-3, Cold Regions Research and Engineering Laboratory, Hanover. Kinosita, S., 1965. Intrusion of a Rigid Cone into snow. Low Temperature Science, Series A, 23. (In Japanese with English summary). Lunne, T., Robertson, P.K. & Powell, J.J.M., 1997. Cone Penetration Testing in Geotechnical Practice. London, Blackie Academic & Professional. Marshall, H.-P., Schneebeli, M. & Koh, G., 2007. Snow stratigraphy measurements with high-frequency FMCW radar: Comparison with snow micro-penetrometer. Cold Regions Science and Technology, Vol. 47(1–2), pp. 108. Masterson, D.M., 2009. State of the art of ice bearing capacity and ice construction. Cold Regions Science and Technology, Vol. 58(3), pp. 99. Renshaw, C.E. & Schulson, E.M., 2001. Universal behaviour in compressive failure of brittle materials, Nature, 412, pp. 897–900. Schulson, E.M., 1990. The brittle compressive fracture of ice. Acta Metall. Mater, Vol. 38(10), pp. 1963–1976. Scapozza, C. & Bartelt, P., 2003. Triaxial tests on snow at low strain rate. Part II. Constitutive behaviour. Journal Of Glaciology, Vol. 49(164), pp. 91–101. van den Berg, A.P., 2002. Gonsite Software v1.2. A.P. v.d. Berg Machinefabriek B.V. Neerenveen, Netherlands.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Parametric study of a free-falling penetrometer in clay-like soils M. Nazem & J.P. Carter Centre for Geotechnical and Materials Modelling, The University of Newcastle, NSW, Australia

ABSTRACT: A free-falling penetrometer (FFP) is a useful apparatus for providing information on the mechanical properties of soil layers, especially where the site in question is inaccessible, which is normally the situation for many seabed, river and swamp deposits. FFP tests can provide useful data such as the total depth and time of penetration and the deceleration characteristics of the penetrometer, provided it is suitably instrumented. The main quantities of interest measured in FFP tests, viz., the time and depth of penetration, depend on the mechanical properties of the soil, such as its shear strength and rigidity index, as well as those parameters associated exclusively with the penetrometer, such as its geometry and the energy with which it first impacts the soil. Potentially, these penetration data can be used to deduce strength parameters for the soil in situ, either through empirical correlations or else by application of a rigorous numerical solution for the ideal penetration problem. A robust finite element analysis procedure was developed at the University of Newcastle, Australia, especially for the analysis of geotechnical problems, including problems that involve soil penetration. This method of analysis is based upon the Arbitrary Lagrangian Eulerian (ALE) finite element technique and takes into account material nonlinearity and rate dependence, large deformations, changing boundary conditions, and time-dependent loading, including an allowance for dynamic forces when necessary. The present study aims to provide some numerical results describing the effect of the mechanical properties of the soil and the energy of the penetrometer on the penetration response. This goal has been achieved by performing a wide range of FFP analyses using the proposed ALE finite element method.

1

INTRODUCTION

A free-falling penetrometer (FFP) is sometimes used instead of the more popular cone penetration test (CPT) at inaccessible sites such as ocean beds, lakebeds, wetlands and rivers. This apparatus can be used to explore valuable information on soil properties which may be essential in the design of foundations for infrastructure such as bridges, offshore platforms, pipelines and dams. A FFP is usually released from a boat or it can be deployed from a helicopter. Frequently, onboard devices are able to measure the total depth and time of penetration and the deceleration characteristic of the penetrometer. Such data can be used to deduce fundamental strength parameters for the soil in situ provided a good correlation between these parameters and the penetrometer travel information is known. The idealised penetration problem is considered to be one of the most complicated initial value problems in geomechanics, because of its dynamic nature and the fact that it involves non-linear and possibly time and rate-dependent materials response, moving boundaries and generally large deformations. Clearly, an analytical solution to the ideal problem of FFP penetration does not exist in the literature, but in addition, ordinary finite element methods consistently fail to provide a valid numerical solution to this problem. © 2011 by Taylor & Francis Group, LLC

Among many others, Aubeny and Dunlap (2003), Lee and Elsworth (2004) and Einav et al. (2004) have presented numerical models for objects penetrating into soil. But the main shortcomings of these models include the use of simplified constitutive laws for the soil, or they may ignore the large deformations and the associated rate effects. Realistic and accurate finite element analysis of a FFP requires special strategies and techniques due to the highly nonlinear characteristics of the problem. Recently, Carter et al. (2010) presented a robust and validated method, based on the Arbitrary LagrangianEulerian (ALE) finite element technique, which can accurately analyse the FFP problem. This strategy is briefly introduced here. The effect of several parameters on the penetration response, such as those governing the strain rate effects on the strength of clay soils and the strength increase with depth normally observed in the seabed, is addressed and some relevant numerical results are reported.

2

PROBLEM DEFINITION

As depicted in Figure 1, a free-falling penetrometer, with mass m and diameter d, impacts the ground surface at initial velocity vo and penetrates into the soil

293

3 3.1

Figure 1. Problem definition.

layer. The total depth and time of penetration are represented by p and tp , respectively, and correspond to the stage when the penetrometer comes to rest. Although a conical tipped, cylindrical penetrometer is shown in Figure 1, the method presented here can be used for other types of objects such as spherical or annular penetrometers. The mechanical properties of soil are represented by its undrained Young’s modulus, Eu , Poisson’s ratio, vu , and undrained shear strength, su . In order to represent incompressible behaviour, vu is assumed to be 0.49 (as an approximation of 0.5). Note that the shear strength of the soil may increase linearly with depth as well as logarithmically with strain rate. The specific forms of these variations are described later. Given the values of vo , m, d, and the mechanical properties of the soil, the ALE finite element method should predict the penetration response and in particular it should provide predictions of p and tp . A numerical model capable of accurately analysing this problem must include the following features: • the ability to model accurately the large deforma-

tions induced in the soil due to penetration of the object; • be able to consider the inertia forces developed in the soil due to the impact of the FFP; • involve a constitutive model for the soil capable of considering its incompressibility during the short loading period, and any possible increase in shear strength due to strain rate effects and depth increase; and • the ability to model the contact interface between the penetrometer and the soil. © 2011 by Taylor & Francis Group, LLC

294

FINITE ELEMENT ASPECTS Large deformations

Due to its many advantages, numerous attempts have been made to employ the finite element method for problems of penetration in geomechanics. In general, this method can handle both simple and complex constitutive soil models, solve problems with complicated geometries and boundary conditions and still provide reasonably accurate results. Nonetheless, mesh distortion is often the main drawback of the more common finite element solution methods, such as the Updated Lagrangian (UL) formulation, when applied to problems of deep penetration. Mesh distortion in zones with high stress/strain concentration such as around the tip of a cone will be inevitable if the finite element mesh employed to solve the problem is made fine enough to capture accurately such high stress concentrations. The finite element method offers two main strategies, h-adaptivity and r-adaptivity, to overcome mesh distortion in problems with relatively large deformations. The h-adaptive finite element method may be able to avoid mesh distortion by generating a new mesh based on sub-dividing into more elements the area where the interpolation should be improved. radaptive finite element strategies overcome the mesh distortion problem by relocating the nodal points in the finite element domain without changing the total number of nodes and elements and without changing the connectivity of the elements. The Arbitrary Lagrangian-Eulerian (ALE) method is probably the most popular r-adaptive technique. It is based on the concept of distinguishing the mesh displacements from the material displacements. This method was adopted in the present study and is briefly discussed in the following. The ALE method, which originated from the disciplines of fluid and solid mechanics, has proved to be robust and very efficient in solving a wide range of geotechnical problems (e.g., Nazem et al., 2006; Nazem et al., 2008; Nazem et al., 2009a). The ALE method eliminates mesh distortion in the Updated Lagrangian (UL) method by separating the material and mesh displacements. In a UL formulation, the spatial coordinates of all nodal points are updated according to the incremental displacements at the end of each time step. In problems with relatively large deformations, this incremental update causes mesh distortion, usually commencing with elements being twisted out of their favourable shapes, decreasing the accuracy of the analysis, and perhaps ultimately resulting in a negative Jacobian. One of the best ways to overcome this shortcoming of the UL method is to separate the material and mesh displacements, leading to the ALE method, as previously indicated. The equilibrium equation in the ALE method can be written in terms of two sets of unknowns, the mesh displacements and the material displacements, leading to the so-called coupled ALE method. Alternatively, one may write the global equations in terms of material displacements only. Such a formulation, known

as the decoupled ALE method, or the operator-split technique, facilitates two steps for the analysis: a UL step followed by an Eulerian step. In the UL step, the incremental material displacements are obtained by solving the equilibrium equation which potentially may result in a distorted mesh if the nodal coordinate are updated. Next, an Eulerian step is carried out, aiming to minimise the mesh distortion by actually refining it. Mesh refinement can be achieved by generating a new mesh for the entire domain or by moving current nodal points to new positions. The Eulerian step is particularly important if significant mesh distortion is likely to occur during the analysis. After mesh refinement, all state variables at the nodal points as well as at the Gauss points must be mapped from the distorted mesh to the new mesh. The ALE method addressed in this study is based upon the operator split technique presented by Nazem et al. (2006, 2009a).

behaviour, the soil is represented by an ideal associated Tresca material, assuming a uniform Young’s modulus. In addition, it is assumed that the undrained shear strength of the soil increases with rate of straining according to the following expression (Graham et al. 1983):

where su represents the undrained shear strength of soil, su,ref is a reference undrained shear strength measured at a reference strain rate of γ˙ ref , γ˙ denotes the shear strain rate and λ is the rate of increase of strength per log cycle of shear strain rate. The reference undrained shear strength at depth y is obtained by the following linear equation:

3.2 Discretisation in time In the first step of an ALE analysis, the displacements, velocities and accelerations are obtained by time-integrating the governing equation that results from an application of the principle of virtual displacements. In its weak form this principle states that for any virtual displacement δu, equilibrium is achieved by satisfying the following equation:

where k is the total number of bodies in contact, σ denotes the Cauchy stress tensor, and δε is the variation of strain due to virtual displacement δu. The symbols u, u˙ and u¨ represent material displacements, velocities and accelerations, respectively, while ρ and c are the material density and damping, b is the body force, q is the surface load acting on area S of volume V , δgN and δgT are the virtual normal and tangential gap displacements, and tN and tT denote the normal and tangential forces at the contact surface Sc . The implicit generalised-α method presented by Chung and Hulbert (1993) is used here to integrate with respect to time the global matrix equations obtained by linearisation of Equation (1).

where su,ref (0) is the reference undrained shear strength of soil at the ground surface (y = 0) and ks indicates the increase in strength per unit depth. Note that in a large deformation analysis, the stress-strain relations must be frame independent. This requirement, known as the principle of objectivity, can be satisfied by introducing an objective stress rate, such as the Jaumann stress rate σ˙ ∇J , into the stress-strain relations as follows:

where C ep represents the elastoplastic constitutive matrix. Among several options to integrate Equation (4), Nazem et al. (2009b) showed that it is slightly more efficient to apply the rigid body rotations implicit in this definition of stress rate while also integrating the constitutive equations. This strategy is adopted in this study.

3.4

Description of contact interface

To describe the contact at the interface between the soil and the penetrometer, constitutive equations must be provided for both the tangential and normal directions to the contact surface. Among several main strategies available in contact mechanics we use the penalty method to formulate these constitutive relations. In the penalty method the normal contact is described by:

3.3 Soil constitutive model As mentioned previously, the FFP is useful for estimating the mechanical properties of soil in remote sites such as seabeds or wetlands. The soils found at such sites are very often cohesive or clay-like, incompressible in the short term, and their shear strength often increases with strain rate as well as with depth below the surface of the deposit. To model such complex © 2011 by Taylor & Francis Group, LLC

where εN is a penalty parameter for the normal contact. The response in the tangential direction is captured by the so-called stick and slip actions. In the former no tangential relative movement occurs between the bodies whilst the latter represents a relative displacement of bodies in the tangential direction. This assumption facilitates splitting the relative tangential velocity

295

between the bodies, g˙ T , into a stick part g˙ Tst and a slip g˙ Tsl part as in the following rate form:

The stick part can be used to obtain the tangential component of traction by:

where εT is a penalty parameter for the tangential contact. Note that the assumption in (6) is analogous to the theory of plasticity in which incremental strains are divided into an elastic part and a plastic part. Following this analogy, we must provide a function to describe the slip criterion. This has been achieved by using the classical Coulomb friction law. For more details see Sheng et al. (2006, 2009). 3.5 Implementation The operator-split technique explained in the previous section has been implemented in the finite element code, SNAC, developed at the University of Newcastle, Australia. A summary of the steps performed in each increment is as follows. 1. UL step: • Assemble the global momentum equation in matrix form and find the displacements, velocities and accelerations. • Integrate the stress-strain relations to update the stresses and internal forces. • Iterate until the unbalanced forces are smaller than a prescribed tolerance. • Update the initial coordinates of all nodal points according to total displacements. 2. Eulerian step: • Refine the mesh, if necessary. • Map all state variables from the old mesh to the new mesh and update mesh coordinates. • Recompute internal forces and perform further equilibrium iterations if necessary. 4

NUMERICAL RESULSTS

4.1 Finite element model

Figure 2. Finite element mesh of free falling penetrometer.

The finite element mesh for a rigid penetrometer falling freely into an undrained soil layer in the vertical direction is shown in Figure 2. Due to axial symmetry, only one-half of the problem cross section is considered. The mesh consists of 10,252 nodes and 4,988 6-noded triangular elements each containing 6 integration points. To avoid further complexity, zero friction between the soil and the penetrometer is assumed in all cases studied here, representing a smooth penetrometer. No material damping is also assumed in the model. The right and bottom boundaries are able to absorb the energy of waves that propagate through the soil due to penetrometer impact, preventing the reflection of such © 2011 by Taylor & Francis Group, LLC

waves. Carter et al. (2010) have reported the validation of the ALE method described here by comparing the numerical predictions obtained using this approach with the results of physical experiments conducted in the laboratory. Typical time steps used in all analyses were either 2×10−5 or 10−5 sec.

4.2

Finite element results

Among the various parameters affecting the total time and depth of penetration, we limit our study

296

Figure 4. Effect of strength increase parameter ks on total penetration.

Figure 3. Effect of rate parameter λ on total penetration.

to the strength rate parameter, λ, and the strength increase parameter, ks . To represent these effects, the normalised initial kinetic energy of the FFP, i.e., 0.5 mvo2 /(πsu,ref (0)d 3 /4), is plotted versus the total (final) penetration depth normalised by the diameter of the cylindrical penetrometer. In order to be specific, it is assumed that Eu /su,ref (0) = 200 in all cases reported here. Recently, Carter et al. (2010) also addressed the effect of the rate parameter λ on the penetration response. Their study was limited to total penetration and normalised initial kinetic energy values of 5.5d and 40, respectively. As shown in Figure 3, the effect of the strength rate parameter is now demonstrated for deeper penetration values achieved by increasing the initial kinetic energy of the penetrometer up to a value of 200. Obviously, larger values of the rate parameter λ correspond to an increase in the shear strength of the undrained soil under rapid dynamic loading. As shown graphically in Figure 3, increasing the value of rate parameter decreases the total depth of penetration. Moreover, the plots on Figure 3 show that for a given value of the rate parameter λ there is almost a linear relation between the normalised initial kinetic energy and normalised penetration, for values of normalised energy up to 200. The effect of the strength increase parameter ks on total penetration is depicted in Figure 4. Note that in all these numerical results the undrained shear strength of the soil at the ground surface is assumed to be 1 kPa, representing a relatively soft surface layer of soil. Two typical values of 1.0 and 2.5 kPa/m were selected for ks and the results are compared with those obtained assuming ks = 0. Also note that in all cases the diameter of penetrometer considered in this study was 0.06 m and the rate parameter was assumed to be 0.2. According to Figure 4, in order to achieve the same total penetration larger initial kinetic energy is required for higher values of ks . For instance, the normalised initial kinetic energy values required to obtain a total penetration of 6d are respectively 136, 151, and 170 for ks = 0, 1, and 2.5 kPa/m. It is noted that the effect of the rate of strength increase with depth is likely to depend also on the scale © 2011 by Taylor & Francis Group, LLC

Figure 5. Velocity versus penetration for d = 0.04 m, m = 0.2 kg, Eu /su,ref = 100, su,ref = 5 kPa, λ = 0.1.

of the problem, i.e., the diameter of the penetrometer. This particular effect will be studied in a subsequent paper. The effect of assuming a constant value of the ratio Eu /su at all depths will also be investigated in future work. Finally, the velocity of the penetrometer versus the penetration is plotted in Figure 5 for the cases where d = 0.04 m, m = 0.2 kg, Eu /su,ref = 100, su,ref = 5 kPa, and λ = 0.1. Figure 5 shows that the velocity of the penetrometer changes approximately quadratically with the penetration and hence the velocity should reduce approximately linearly during penetration.

5

CONCLUSIONS

The analysis of a free-falling penetrometer by the Arbitrary Lagrangian-Eulerian finite element method was address in this study. The numerical results obtained by this method indicated that the total penetration depth of the penetrometer is sensitive to strain rate effects in the soils as well as its inhomogeneity.

297

ACKNOWLEDGEMENT The research in this paper was supported by a Discovery Project grant, DP0666778, funded by the Australian Research Council, which the authors gratefully acknowledge. REFERENCES Aubeny, C. & Dunlap, W. (2003). Penetration of cylindrical objects in soft mud, Oceans 2003, San Diego, 2068–2073. Carter, J.P., Nazem, M., Airey, D.W. & Chow, S.W. (2010). Dynamic analysis of free-falling penetrometers in soil deposits. Plenary paper accepted for presentation at GeoFlorida 2010, ASCE, Feb. 2010. Chung J. & Hulbert G.M. (1993). A time integration algorithm for structural dynamics with improved numerical dissipation: the generalized-α method, Journal of Applied Mechanics, 60, 371–5. Einav, I., Klar, A., O’Loughlin, C.D. & Randolph, M.F. (2004). Numerical modelling of deep penetrating anchors, Proc. 9th ANZ Conference on Geomechanics, 2, 591–597. Graham, J., Crooks, J.H.A. & Bell, A.L. (1983). Time effects on the stress–strain behaviour of natural soft clays. Géotechnique, 33, 327–340.

© 2011 by Taylor & Francis Group, LLC

Lee, D.S. & Elsworth, D. (2004). Indentation of a free falling sharp penetrometer into a poroelastic seabed, Journal of Engineering Mechanics, ASCE, 130, 2, 170–179. Nazem, M., Sheng, D. & Carter, J.P. (2006). Stress integration and mesh refinement in numerical solutions to large deformations in geomechanics. International Journal for Numerical Methods in Engineering, 65, 1002–1027. Nazem, M., Sheng, D., Carter, J.P. & Sloan, S.W. (2008).Arbitrary Lagrangian-Eulerian method for large-deformation consolidation problems in geomechanics. International Journal for Analytical and Numerical Methods in Geomechanics, 32, 1023–1050. Nazem, M., Carter, J.P. & Airey, D.W. (2009a). Arbitrary Lagrangian-Eulerian Method for dynamic analysis of Geotechnical Problems. Computers and Geotechnics, 36 (4), 549–557. Nazem, M., Sheng, D., Carter, J.P. & Sloan, S.W. (2009b). Alternative stress-integration schemes for large deformation problems of solid mechanics. Finite Elements in Analysis and Design, 45, 934–943. Sheng, D., Wriggers, P. & Sloan, S.W. (2006). Improved numerical algorithms for friction contact in pile penetration analysis. Computers and Geotechnics, 33, 341–354. Sheng, D., Nazem, M. & Carter, J.P. (2009). Some computational aspects for solving deep penetration problems in geomechanics, Computational Mechanics, 44, 549–561.

298

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

The future of deepwater site investigation: Seabed drilling technology? J.J. Osborne & A.G. Yetginer RPS Energy, UK

T. Halliday Subsea Minerals, UK

T.I. Tjelta Statoil, Norway

ABSTRACT: A deepwater site investigation was planned for Norwegian Sea Luva Gas Field in 2009. During the early stages of site investigation planning for Luva, various deepwater site investigation options were considered: surface mounted drilling, seabed based drilling, combined seabed PCPT and deep gravity/piston core option and acoustic coring. An extensive market survey was performed which in particular researched the currently available seabed drilling technology concepts, variations in their design and operation and their limitations. This Paper highlights the main issues raised during the market survey and technical evaluation, discusses the various site investigation systems available for deepwater geotechnical site investigations and identifies the relative advantages and disadvantages of each. The Paper is the first in a series of 4 which together describe the Project in detail. The Paper should be read in conjunction with the others which follow during this conference session.

1

INTRODUCTION

Statoil commissioned RPS Energy, working together with Subsea Minerals, to conduct a Deepwater Site Investigation System Market Survey and Technical Evaluation in preparation for a deepwater geotechnical site investigation programme for the Luva field offshore Northern Norway. Statoil considered that the exposed deepwater location, coupled with the anticipated ground conditions at Luva presented a significant risk to the success of geotechnical site investigation within a strict 4 week operational period. Therefore the choice of site investigation systems had to be optimised in advance of the campaign. The requirement was to identify a geotechnical sampling and testing system that would be able to repeatedly obtain high quality soil and in-situ test data. In addition to this the system was required to be able to install piezometers. In view of the recent developments in deepwater geotechnical site investigation systems the Market Survey and Technical Evaluation was commissioned to: – Compile an overview of the available “fit for purpose” deepwater geotechnical site investigation systems suited to the conditions at Luva; – Evaluate the viability of deploying each of these systems; – Critically appraise the systems’ performance in terms of actual data versus that promoted;

© 2011 by Taylor & Francis Group, LLC

299

– Estimate the risk of inadequate sample recovery or quality for each system; – Determine and assess the additional services provided by the system operators (to include company infrastructure, geotechnical engineering support services, offshore and onshore laboratory testing, data interpretation, reporting etc.); – Assess the risks associated with each system in terms of system operational reliability, and ability to consistently produce high quality samples and in-situ test data; – Perform a high level assessment of the commercial risks associated with the various systems; – Determine the attitude to risk and cost sharing where innovative systems are not yet proven in the field or do not yet have an established track record; – Summarise the perceived technical and commercial benefits of each option. The Market Survey and Technical Evaluation report on geotechnical site investigation systems presented to Statoil was of considerable detail and only the main discussion items are included in this Paper. Further information regarding the performance of the systems subsequently deployed at Luva are presented in the accompanying Papers, namely : Tjelta & Yetginer (2010), Yetginer & Tjelta (2010), and Tjelta & Strout (2010).

2

DATA SOURCES

The following data sources were called upon in order to acquire the necessary information required for the Market Survey and Technical Evaluation: – In-house knowledge (Subsea Minerals and RPS Energy); – Industry contacts and associates including oil & gas clients with relevant contracting experience; – RPS Energy consultants (Offshore QC reps/project managers on previous relevant contracts) – Publicly available data, (Contractor’s literature); – Published Journal/Conference Technical Papers; – Internet; – Results of a Technical and Commercial questionnaires issued to select system operators; – Unpublished, non-confidential data; – Contractors’ references. Unpublished information, views and opinions supplied by individuals highly experienced in the use and performance of the various systems was reported anonymously and confidentially and as such is not included in this Paper. However such reliable information from experienced and informed individuals was of extremely high value to this Project. Information was also obtained from press releases, articles in trade magazines and general information supplied through the scientific community. Additionally factory and site visits were conducted to view manufacturer’s equipment and system components in operation under trial conditions. Questionnaires were distributed to selected contractors/suppliers requesting specific information on existing and planned equipment. As the nonpublished information and questionnaire replies were supplied anonymously and confidentially it is not possible to include them within this Paper, however the intent of some comments provided may be inferred. Some contractors also supplied us with client references which were duly contacted in order to obtain additional information. An internet based review was also conducted and the available systems were subdivided into the following categories: – Conventional ship-borne drilling and sampling systems; – Free-fall and driven seabed “one shot” sampling and seabed PCPT systems; – Remotely operated seabed drilling and sampling systems.

3

SHIP MOUNTED SYSTEMS

Ship mounted systems are established throughout the geotechnical industry and their relative merits and weaknesses well reported, Randolph et al. (2005). The nature of the sediments (very soft), the water © 2011 by Taylor & Francis Group, LLC

300

depth and the requirement for undisturbed samples all led to the conclusion that a ship-borne drilling system may not deliver the quality information required within a constrained operational period. The trip times for the drill-pipe and the sampling and testing tools would be significant and require a minimum 48 hour weather window to deploy/retrieve the pipe and test to 40 m. The recovery of high quality undisturbed samples in the very soft sediments would be challenging in the presence of wave action on the drillship. 3.1 Advantages – Long track record – Wide range of sampling/testing systems available 3.2

Disadvantages

– Long pipe tripping times – Long sample recovery to deck potentially leading to reduction in sample quality Further discussion on the comparison of ship mounted systems with seabed drilling technology is presented by Yetginer & Tjelta (2010). 4

SEABED UNCASED ONE SHOT SYSTEMS

Including the NIOZ, STACOR and Calypso systems these samplers have obtained excellent cores in water depths of up to 2000 m. Laboratory testing has been carried out on samples recovered and in some cases the results fall into the “very good to excellent” category. Other more basic gravity corers are available generally with samples of reduced quality. There are disadvantages associated with each system and these range from their failure to sample some soils (missing sections), to the need for relatively benign sea conditions for launch and recovery. 5 5.1

REMOTELY OPERATED DRIVEN SEABED SYSTEMS Deep push PCPT

Deep push PCPT systems have been developed by a number of contractors and have delivered excellent results in a wide range of soil types. 5.1.1 Advantages – Continuous sub-seabed profiling. – Good depth control. – Several tool options (choice of cone geometry, spherical ball penetrometer and T-bar). – PCPT’s have been used extensively world-wide and have a good track record of high quality geotechnical data provision.

approximately 6,000 m. More recent integration with ROV technology has significantly increased their degree of sophistication and capability as an all round sampling and testing tool. Recent interest in deep sea sulphide deposits has led to a surge in interest of this type of technology as cored samples are required. Some units have been used by the scientific community to recover samples and cores but to date only one drill has a demonstrable track record in geotechnical site investigation, Pennington & Kelleher (2007). However at least 3 other contractors are developing seabed drill systems with a view to entering the offshore minerals and geo-technical site investigation markets.

5.1.2 Disadvantages – Geotechnical parameters are empirically derived. – Results are dependant on correct zeroing and offsets for the soil type and water depth. – Often different systems may provide different results. Whilst the quality of data that could be recovered was not in question the investigation at Luva required high quality soil samples as well as piezocone data. 5.2 Deep Water Sampler (DWS) The Deep Water Sampler (DWS) is believed to be the only commercially available driven deep penetration piston corer, Lunne et al. (2008). The DWS is penetrated at a controlled rate and the advanced lining, tool cutting edge and retracted sample retention systems offer the most advanced deepwater soft soil sampling system with proven repeatability and reliability. 5.2.1 Advantages – Over 95% core recovery (100% has been recorded). – System designed to cause minimum disturbance to the soil, disturbance index in the “very good to excellent” and “good to fair” categories. – DWS has a piston mechanically tethered to the seabed frame ensuring the piston stays at seabed level during sampling. – Cores up to 15 m below seafloor have been achieved to date – greater depth is possible if handling system were to be further developed. – Controlled penetration at 2 cm/sec. – Deployed using the ROSON system which is also used for seabed PCPT’s. – Good depth accuracy control. – Inside sampler diameter of 110 mm. – Real-time measurement of penetration, under pressure behind sample and the force in the chain tethering the piston. – Can operate on a seabed gradient of up to 12 degrees. – A retractable seabed bearing system has been developed which supports the tool during sampling and reduces potential handling difficulties when passing through the splash zone during deployment and recovery. 5.2.2 Disadvantages – Present sample depth limited to 10 m bml due to change in deployment system from earlier projects/testing. – The system requires significant care during deployment and recovery, and requires specific handling mechanism. 6

REMOTELY OPERATED SEABED DRILLING UNITS

6.1 Advantages – Rapid deployment to seabed. – Ability to traverse between sites without recovery to deck (subject to certain conditions). – Ability to utilise short weather windows. – Support vessel can remain on station during short periods of unfavourable weather conditions. – Highly accurate depth control. – Little or no surcharge to sediments during sampling/testing process. – No human interface during pipe handling operations. 6.2 – – – – –

Largely unproven in the North Sea theatre. Limited installed power. No real-time sample information. Sample diameter issues. Launch and recovery operational limitations.

7

MARKET SURVEY AND TECHNICAL EVALUATION

The market survey identified the level of technology available and this allowed for comparisons to be made between different types of system and between types of seabed drill. From these comparisons it was possible to make an objective assessment of the target technology for the Luva site investigation. Based on our evaluation of the available commercial systems matrices were compiled summarising the operational and environmental advantages and disadvantages of each system. The scoring adopted was based in the capability and suitability of the equipment to perform the headline operations, a score of zero in any of the categories indicating that the equipment does not fulfil the operational criteria, and as such would be discounted as a viable option. The scoring is summarised as follows:

Seabed drills have been used for over 30 years to recover samples from water depths of up to © 2011 by Taylor & Francis Group, LLC

Disadvantages

301

– 5: most suited or capable, delivers highest performance; – 4: good performance but requiring some modification or improvement;

Table 1.

Summary of technical considerations for seabed sampling and testing systems.

– 3: average performance or suitability, not an ideal record or history; – 2: below average performance; – 1: barely able to perform required function; – 0: totally unsuitable and/or unavailable and not considered further. A matrix is presented in Table 1, entitled “Summary of Technical Considerations for Seabed Sampling and Testing Systems”, which demonstrates the attempt to objectively compare all available systems in order to provide the lowest risk system for the Luva project. The reliability and quality of data provided by a seabed deployed PCPT system is clearly borne out in this table. A comparison table for six (6) of the seabed drills was similarly developed using the same scoring system for the assessment of the Seabed drilling System Technical Risk. This matrix is presented as Table 2. Whilst it is accepted that this table is not an empirical comparison, it is the authors’ opinion that the breadth of the category criteria fairly reflected the suitability of each unit for deployment at Luva. It is important to note that the comparison were based upon the data available at the time of the study for a specific project requirements and ground conditions. It is recognised that the remotely operated seabed drilling system industry is developing at an unprecedented rate with system capabilities being continually enhanced. The University of Bremen/Marum MeBo was unavailable due to other commitments, as was the DWACS of Williamson Associates. Seafloor Geoservices M80, John Gregg’s seabed drill and Benthic Geotech’s PROD2 were in advanced stages of either building or testing. PROD1 had been field proven for 10 years with its deployment largely restricted to Australasian carbonate rich soil conditions. The sampling and testing capabilities of each available unit appeared similar, with all offering coring, PCPT’s, push sampling and in most cases piston © 2011 by Taylor & Francis Group, LLC

302

sampling systems. Both PRODs offered a high quality Hydraulically Tethered Piston Sample but this had not been extensively utilised in North Sea normally consolidated soils. A preconception amongst the authors during the Survey was that wireline, as opposed to conventional rod coring systems, would be advantageous. The perceived advantages included being to effectively case the hole and reduce pipe handling times, and that these benefits would offset those from using a thinner kerf conventional rod system. Such benefits offered by wireline systems on land and in shallow marine conditions have been extensively demonstrated and direct parallels could be applied. However, if only the unit’s performance in normally consolidated cohesive soils is considered then the wireline/conventional argument becomes less obvious. It is not a major issue if the ground collapses upon sample withdrawal providing the sample depth can be reached rapidly on the next tool deployment with no disturbance of either the sample taken or the ground over the next test interval. In this instance PROD1 demonstrated a well evolved automated routine which offered the shortest tool handling times, particularly if the target depth was less than 50 m below mudline. None of the other units demonstrated this high degree of efficiency whilst employing auto routines. The authors considered that a remotely operated seabed drilling system would be able to satisfy the project requirements, and Benthic Geotech’s PROD1 was recommended as a viable alternative to a vessel mounted drill system for the Luva Project.

8

CONCLUSIONS

In conclusion remotely operated seabed drilling units are likely to feature as viable systems for deep water

Table 2.

Seabed drilling system technical risk.

geotechnical site investigation, although they are not necessarily to be considered as a panacea. As petrochemical exploration moves into deeper water in remote locations more investigations are likely to encounter soft normally consolidated soils. Environmental conditions in these remote locations are unlikely to favour vessel mounted systems and the deployment and recovery of deep water samplers may be challenging. Whilst this is also true of the seabed drilling units they will be able to perform several sample holes and an even higher number of PCPT locations. As confidence within the industry grows bespoke LARS, (Launch and Recovery Systems), cursors may be developed that will enable vessels to operate in more extreme environments. Hardware and software will continue to improve and these gains will be integrated into the coming generations of seabed drilling units. If the quality of piston samplers could be guaranteed with no loss of production or increase in manual handling issues the drills would be a very powerful alternative. Soils with a considerable silt, sand or gravel content, or those requiring significant energy to drill, will be more suited to vessel systems where the installed power capabilities, integrated mud systems and range of sampling and testing options will favour them. © 2011 by Taylor & Francis Group, LLC

Serious consideration should be given to using a seabed drill should any of the following criteria apply to a site investigation: – Medium to deep water. – Soft, normally consolidated cohesive soils. – Exposed locations with short seasonal weather windows. – Requirements for high quality samples in addition to in-situ test data. – Piezometer installation.

ACKNOWLEDGEMENTS The Authors would like to thank Statoil for commissioning this work and for their permission to publish the Paper at this conference. In particular we would like to acknowledge the considerable support that we have received both from Tor Inge Tjelta and his Associates within Statoil. The Authors would also like to thank Robert Goodden of SubSea Minerals, and Callum Duffy, Spencer Chiu Christopher Kilsby and Steve Mulley of RPS Energy, for their input into this Project.

303

REFERENCES Tjelta, T.I. & Yetginer, A.G. 2010. Luva deepwater site investigation programme and findings. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Yetginer, A.G. & Tjelta, T.I. 2010. Seabed Drilling vs. Surface Drilling – A Comparison. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Tjelta, T.I. & Strout, J. 2010. Piezometer installation in deepwater Norwegian Sea. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth.

© 2011 by Taylor & Francis Group, LLC

304

Randolph, M. Cassidy, M. Gourvenec, S. & Erbrich, C. (2005). Challenges of Offshore Geotechnical Engineering. Proc.1ST International Symposium on Frontiers in Offshore Geotechnics: ISFOG, Perth. Lunne, T. Tjelta, T. I. Walta, A. & Barwise, A. 2008. Design and Testing Out of Deepwater Seabed Sampler. Offshore Technology Conference, Houston. Pennington, D. & Kelleher, P. 2007. PROD Delivers an Accurate Site Investigation at Maari. Procs. 6th Intl. Offshore Site Investigation and Geotechnics Conference, London.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Mini T-bar testing at shallow penetration A. Puech Fugro Offshore Geotechnics, Nanterre, France

M. Orozco-Calderón & P. Foray Grenoble Institute of Technology, Laboratoire 3S-R, Grenoble, France

ABSTRACT: An extensive series of static and cyclic mini T-bar tests were conducted in a tank allowing a lateral visualisation and an implementation of imaging techniques in an attempt to infer the kinematic field in the soil around the bar. The shear strength as well as the nature of the clay (structured versus non-structured) was varied. Influence of parameters like bar roughness and rate of penetration were also analysed. Results clearly indicate that full flow mechanisms, corresponding to the closure of the trench opened by the T-bar, initiate at penetrations higher than those generally considered. Guidelines for shallow mini T-bar testing are tentatively suggested.

1

INTRODUCTION

1.1 Development of mini T-bar testing offshore The T-bar penetrometer was first introduced at the University of Western Australia (UWA) to improve the accuracy of strength profiling in centrifuge testing (Stewart & Randolph 1991). The probe consisted of a cylindrical bar (5 × 20 mm) attached at right angle to the penetrometer rods, just below a load cell. Some years later a larger T-bar (40 × 250 mm) was constructed by Fugro for in situ applications and firstly used in 380 m of water in the Timor Sea (Randolph et al. 1998). The T-bar was rapidly recognized as a tool of great potential for investigating the undrained shear strength of soft deepwater sediments because i) corrections for overburden pressure or pore pressure can be neglected, which is not the case for cone penetrometers ii) the Nt factor relating bearing resistance to undrained shear strength is determined by exact plasticity solutions, even if this represents ideal soil conditions, iii) measurements can be made more sensitive than for the cone (bearing area is larger and cell has only to measure differential pressures on both faces of the bar) and iv) cyclic T-bar testing gives an estimate of the soil sensitivity. Most of deepwater site investigations performed presently include T-bar profiling and cyclic T-bar testing. Soils encountered in deep waters generally present shear strength profiles linearly increasing with depth (gradient between 1 to 2 kPa/m), starting at very low values at seabed (typically 1 to 3 kPa). In these very soft soils assessing the longitudinal and lateral restraints of pipelines/flowlines laid on the sea bottom is a challenging issue for the industry. This requires both a very accurate measurement of the intact and remoulded undrained shear strength over the first © 2011 by Taylor & Francis Group, LLC

decimetres (accuracy of the order or better than 1 kPa) and a precise soil-pipe interaction model. Box corer sampling is a convenient way to recover typically 0.5 × 0.5 × 0.5 m samples of as intact as possible seabed sediment on board the vessel. Due to the extremely soft consistency of these soils, “in situ” testing inside the box is the only reasonable way to obtain relevant shear strength data. This can be done with a mini-vane but using a mini T-bar has the advantages of i) obtaining continuous strength profiles and ii) estimating the sensitivity by performing quick cyclic tests. A manually operated mini T-bar – called DMS – was developed at UWA for strength profiling in box cores (Randolph et al. 2007). In 2008, Fugro developed its DECKSCOUT™system shown on Figure 1a. The tool can be fitted with full flow penetrometers: T-bars or Ball probes. The mini T-bar is 12 × 75 mm. The stroke is 0.5 m. The actuator can apply constant rate of penetrations in the range 0.01 mm/s to 2 cm/s. Cyclic testing for sensitivity measurement can be automatically monitored. The Fugro SMARTPIPE®(Hill & Wintgens 2009) and SMARTSURF™systems (Borel et al. 2010, Colliat et al. 2010) are also equipped with a 1 m stroke mini T-bar to characterize the seabed sediments at shallow penetrations. 1.2

Interpreting mini T-bar data

Whereas the full flow mechanism developing around a bar at large penetrations is fairly well known and interpretation methods are established, poor information is available on the failure mechanisms taking place at very shallow penetration, i.e. within the first 0.2–0.3 m below seafloor, which are of paramount importance for soil-pipe interaction assessment.

305

Table 1.

Index properties of the natural and artificial soils.

Soil description

w %

wL %

wP %

γ
kN/m3

GoG, natural soil 150–200 170 125 3 50B/50K 110–150 163 132 3.7

Table 2.

Figure 1. a) Fugro DECKSCOUT system mounted on box corer and b) mini T-bar system installed on Visucuve tank.

Cv m2 /year 4

Dimensions of the T-bars used in the experiments.

T-bar type

Bar diam. mm

Length mm

Shaft diam. mm

Surface

Fugro T-bar S T-bar R

12 12 12.4

75 75 75

11.3 10 10

Smooth Smooth Rough

2.2 Tested soils Observations from tests performed in box cores indicated behaviour that is different to a deeply embedded T-bar, which is the typical application of this device: soil stuck to the central rod during T-bar extraction, evidence of a trench remaining open behind the bar at shallow penetration. As it is impossible to observe the real mechanisms occurring around the Tbar in field conditions, it was decided to perform full scale laboratory tests in a large tank allowing a lateral visualisation of the penetration. The main objectives of this experimental program were (i) check at which depth the full flow failure takes place in static and/or in cyclic conditions, (ii) define which kind of failure is taking place at lower penetration, (iii) verify the effects of the rate of penetration and surface roughness and (iv) define appropriate Nt factors for the shallow depths corresponding to the pipe-soil interaction zone.

2 2.1

Two different soils were tested: a reconstituted soil that allowed large quantities to be prepared and a natural Gulf of Guinea (GoG) soil recovered in a box corer from offshore Congo. The reconstituted soil was formed by mixing kaolin and bentonite, in equal proportions, with natural water. This artificial mixture 50B/50K simulates the characteristics of plasticity and undrained shear strength of soft soils of deepwater seabed (Orozco-Calderón et al. 2007). The index properties of the two soils are given in Table 1. 2.3 T-bars used Two T-bars, similar to the one used in field conditions by Fugro, were manufactured for this study. One was fabricated from inox steel (S) with a smooth surface, the other from aluminum alloy (R) with a fully rough surface. The dimensions of the models are given in Table 2. 2.4

EXPERIMENTAL PROGRAM Setup

The tests were conducted in the large tank “Visucuve” of Laboratoire 3S-R in Grenoble. It consists of a large rigid tank having internal dimensions of 2 m-length, 1 m-width and 1 m-depth. Only 0.4 m-depth was filled with the soft soil. The advantage of this tank is that two glass planes are placed laterally around the soil in order to allow a direct visualization of the evolution of the failure patterns around the T-bar during penetration. A vertical SKF electromechanical actuator supports the T-bar and the load cell and can be fixed in different positions to the frame above the tank (Figure 1b). This allows tests to be performed in the central part of the tank to obtain penetration profiles or close to the glass walls for visualization purpose. The rate of penetration of the cell can be varied from 2 mm/s to 20 mm/s and cyclic penetration tests can be performed. © 2011 by Taylor & Francis Group, LLC

Experimental program

In a first step, an extensive series of penetration tests were performed for the two soils in the central part of the tank to evaluate the influence of the penetration rate, surface roughness, undrained shear resistance and the presence of a film of water above the mud line. The artificial soil was prepared in order to get two values of undrained shear strength: 1.5 and 3 kPa. In each case, the soil mass was prepared with a constant water content w, aiming at an homogeneous soil profile. The undrained shear strength su was determined from a correlation previously established between su and the mixture water content using fall cone tests and miniature vane tests (Orozco-Calderón et al. 2007). The average value of su for the GoG soil was around 3.5 kPa. Static and cyclic penetration tests were conducted. The cyclic tests included a static penetration to 20 cm, followed by a first series of 30 extraction-penetration

306

Figure 2. Images taken during visualisation of T-bar first penetrations. Smooth surface, water table on surface, su = 3 kPa. (T-bar section corresponds to black dot at bottom of picture).

Figure 3. Extraction of T-bar. First penetration lower than closure depth. Smooth surface, water table on surface, su = 3 kPa.

sea sediments, particularly where mini T-bar data are of interest. Closure relative depths in the range 12 to 15 fall in fair agreement with an extrapolation of White et al’s (2010) predictions. During the extraction phase of the T-bar, a difference was observed according to the depth where the extraction started. If the extraction starts at a depth lower than the depth corresponding to the closure of the trench, a mass of soil is dragged in the trench above the T-bar, as shown in Figures 3a–c. On the opposite, if the extraction starts at a depth higher than the closure depth, a full flow mechanism is observed.

cycles between 10 and 20 cm. The T-bar was then penetrated to 40 cm and a second series of cycles were applied between 30 and 40 cm. In a second step, direct visualisation tests using a video camera were conducted with static and cyclic penetrations close to the lateral glass of the tank, varying the same parameters. In a third step, photos were taken during the visualisation tests at a rate of 3 or 4 per second with a professional Nikon D3 camera in order to get a continuous picture of the penetration and to allow the implementation of imaging techniques to infer the kinematic field around the bar.

3.2 3

DIRECT VISUALISATION RESULTS

3.1 Continuous penetration-extraction An example of visualisation of the continuous penetration of the T-bar S in the reconstituted soil 50B/50K with su = 3 kPa, and with a layer of water at the soil surface, is presented in Figure 2a for a penetration rate of 4.4 mm/s and a penetration of 10 diameters, and in Figure 2b for a rate of 25 mm/s and a penetration of 13 diameters. Figure 2 clearly shows that the trench remains open, independently of the rate of penetration and even in the presence of water which fills the trench. The closure of the trench was observed at z/D close to 15 (z = penetration depth). The same experiments in the soil with a lower value of su = 1.5 kPa indicate a similar trend, but with a width of the trench becoming smaller and a “closure” of the trench occurring at a lower depth, around 12 diameters. Results obtained from numerical modelling were presented by Barbosa-Cruz & Randolph (2005), Zhou & Randolph (2009), and White et al. (2010) suggesting that the transition between shallow and deep mechanism is governed by the normalised shear resistance su /γ
D, γ
being the submerged unit weight of the soil and D the T-bar diameter. su /γ
D values for our experiments are within the range 30 to 70, i.e. higher than those used in simulations, but corresponding to parameters which may be encountered in deep © 2011 by Taylor & Francis Group, LLC

Cyclic penetration tests

The visualisation showed that the cyclic movement of the T-bar induces the closure of the trench at depths slightly lower than the depths required for full closure at first penetration. Once initiated, closure becomes efficient over the total cyclic displacement amplitude. The stabilization of the curve representing the penetration-extraction resistance of the T-bar was observed after about 10 cycles, as shown on Figure 4 where cycles were applied at a standard rate of 20 mm/s between 10 cm and 20 cm. This corresponds to the number of cycles required to obtain the sensitivity St of the reconstituted soil 50B/50K. This figure highlights a major difference in the remoulding process depending on whether a water table is present or not. In presence of water the remoulding is stronger principally in the upper part of the cycles. This is attributed to an increase of water content of the clay during cycling. Such phenomena were never observed when cycling was performed with a water table but at depths where flow around has occurred during the initial penetration. It can be concluded that, in order to obtain the “true” sensitivity of the soil (i.e. at constant water content) tests have to be performed sufficiently deep within box cores (or below mudline). 3.3

Effects of penetration rate and bar roughness

A significant rate affect was observed, with an increase of about 20% in the T-bar resistance when changing

307

Figure 4. Comparison of cyclic T-bar tests with and without a layer of water above the mud line. Figure 6. Normalised values of Nt /Ntdeep versus z/D for two mini T-bar tests. Clay with su = 1.5 and 3 kPa and su /γ
D = 32 and 64, respectively.

in presence of an open trench (adapted from Yafrate et al. 2007) and ii) strain rate effects. All lab test data with 12 mm diameter bars were adjusted for a rate of penetration of 6 mm/s to match with standard bars of 40 mm penetrated at 20 mm/s. The derived Ntdeep value applicable below the closure depth was found close to 12, in agreement with theoretical predictions for rough bar surfaces (Randolph and Houlsby 1984). The evolution of the ratio Nt /Ntdeep with relative depth z/D (D = bar diameter) is represented in Figure 6. Results compare well with large deformation finite element modelling presented by White et al. (2010). Figure 5. Comparison of T-bar tests with smooth and rough surface, (a) first 4 cycles and (b) evolution with number of cycles.

the penetration rate from 1.6 mm/s to 22 mm/s. More interesting is the comparison between the results of Tbar tests performed with a smooth surface and with a rough surface. Figure 5 presents the penetration resistance obtained during the first 4 cycles (Fig. 5a) and its evolution with the number of cycles (Fig. 5b). No significant difference can be noted between the two tests, and this surprising result was confirmed by all the different series of tests performed in this campaign. This may be due to the very high plasticity of the clay.

4

Nt FACTOR

The values of the Nt factors (Nt = qt /su , qt bearing resistance) were derived only from the tests performed in the central part of the tank, in order to eliminate the possible effects of friction of the T-bar section on the glass wall. Raw measurements were corrected for i) differences in pressures under and above the bar © 2011 by Taylor & Francis Group, LLC

5 APPLICATION OF IMAGING TECHNIQUES ON A NATURAL DEEP SEA SEDIMENT 5.1

DIC method and surface preparation

The high quality digital images taken at different stages of the penetration were treated using the Digital Image Correlation (DIC) method (Hall et al. 2009) in order to obtain the incremental displacement field at various depths and put into evidence the different failure mechanisms around the T-bar. For this purpose, various problems had to be solved: (i) as the penetration rate of the T-bar is relatively fast, a camera which takes pictures at a rate of at least 3 to 4 images per second was necessary, in order to be able to compare successive pictures, (ii) the natural texture of the clay does not provide sufficient contrast to allow monitoring of the displacement of discrete points. Thus, it was necessary to prepare the soil surface with a painting whose texture was adapted to the image treatment. This was particularly difficult for the GoG natural soft clay, which has a brown dark colour and a high water content making it difficult for painting. After various attempts, a suitable texture was obtained by creating artificially

308

Figure 7. Surface texture of GoG natural clay for DIC tests.

a net of cracks on the painting (Orozco-Calderón 2009, Fig. 7). 5.2 Failure mechanisms versus depth The tests using imaging techniques were performed on the natural clay extracted from box cores taken in the Gulf of Guinea. The penetration-extraction tests put into evidence a trench remaining open up to a depth higher than for the reconstituted 50B/50K clay (z/D of about 20). This can be explained by the higher degree of structure of the natural GoG clay (De Gennaro et al. 2005, Colliat et al. 2010). The displacement fields obtained at different penetrations by the DIC method suggest a transition between a shallow foundation type failure close to the surface and a deep failure mechanism at large depth. The deep failure mechanism is illustrated on Figure 8 corresponding to a relative penetration z/D = 11.3.The image clearly shows that the mechanism is a combination of a flow mechanism and an expansion of the soil in front of the tip, which can be compared to the “cavity expansion” considered in a cone or pile penetration test. The flow mechanism is insufficient to close the gap above the bar. At larger penetrations (and up to z/D = 31.3, limit of the tests) the mechanism remains basically the same. Below about z/D = 20, the flow is sufficient to observe closure of the trench a few centimetres above the top of the bar, which means that a gap is still present in the track of the bar. This specific behaviour may be related to the high degree of structure of these superficial deep sea sediments and suggests that the theoretical interpretation of T-bar penetration may be more complex in that kind of soils than admitted for non structured clays. © 2011 by Taylor & Francis Group, LLC

309

Figure 8. Digital Image Correlation: vector displacements for z/D = 11.3, GoG natural clay.

5.3

Sensitivity of the natural clay

Cyclic tests applied to the GoG clay are presented on Figure 9. The sensitivity St is approached only after 30 cycles. This result is on line with field observations in GoG clays (Borel et al. 2010) and independent of the presence of a water table in the lab tests. It was clearly observed that a regime of full-flow takes place progressively during cycling. During the very first cycles, the full flow involves a relatively large zone (as large as about 5 times the diameter of the bar). This zone reduces with the number of cycles to reach 2 to 3 times the diameter of the bar after 10–15 cycles. This behaviour is likely to be related to the structure and sensitivity of the natural deep sea sediment.

6

CONCLUSIONS AND RECOMMENDATIONS

Mini T-bar testing is increasingly used to determine the undrained shear strength and sensitivity of very

REFERENCES

Figure 9. Cyclic tests on GoG natural clay and sensitivity measurements during penetration and extraction.

soft clays for direct application to soil-pipe interaction analyses. The experiments conducted show that the full flow mechanism takes place only after a certain penetration of the bar (i.e. below full closure of the trench). This critical depth depends on the characteristics of the clay (shear strength, structure) but not on the rate of penetration or on the roughness of the bar (at least in the very high plasticity clays tested). Guidelines for performing mini T-bar tests in situ or in box corers can be tentatively based on: – testing at a rate of penetration of 6 mm/s to minimize rate effect between standard and mini T-bar; – starting cycling tests at relative penetrations in excess of 15 to guarantee proper initiation of full flow around the bar and constant water content during the remoulding process; – limiting the amplitude of cycles to ±2.5D and performing at least 10 cycles. In highly structured clays as those encountered in the Gulf of Guinea, it is recommended to increase the relative penetration to 20 and perform at least 30 cycles to derive the sensitivity. Failure mechanisms during the first penetration phase are complex and deserve more detailed analysis. ACKNOWLEDGEMENTS The authors would like to thank Dr. Steve Hall, Lam Nguyen and François Bonnel for their assistance in visualisation and in applying the DIC method.

© 2011 by Taylor & Francis Group, LLC

Barbosa-Cruz, E.R. & Randolph, M.F. 2005. Bearing capacity and large penetration of a cylindrical object at shallow embedment. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG 2005, Perth, W.A, pp.615–621. Borel, D., Puech, A. & Po, S. 2010. Optimized site investigation strategy to obtain shear strength design parameters in deepwater soils. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG 2010, Perth, W.A. Colliat, J.L., Dendani, H., Puech, A. & Nauroy, J.F. 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences, Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG 2010, Perth, W.A. De Gennaro, V., Delage, P. & Puech, A. 2005. On the compressibility of deepwater sediments of the Gulf of Guinea, Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG 2005, Perth, W.A,pp.1063–1069. Hall, S.A., Muir Wood, D., Ibraim, E. & Viggiani, G., 2009. Localised deformation patterning in 2D granular materials revealed by digital image correlation, Granular Matter, DOI 10.1007/s10035-009-0155-1. Hill, A.J. & Wintgens, J-F. 2009. In situ measurement of soilpipe interaction in deep water – Results of a successful offshore campaign. Proc. SUT Annual Conf., Perth, W.A. Orozco-Calderón, M., Foray, P. & Nauroy, J-F. 2007. Pipe-soil dynamic stiffness in soft soils, Proc.17th Int. Offshore and Polar Eng. Conf., ISOPE 2007. Lisbon, Portugal, Paper N◦ JSC-267, Vol.2, pp.1193–1198. Orozco-Calderón, M. 2009. Etude de l’interaction cyclique sol-pipe dans les grands fonds marins, PhD Thesis, Institut Polytechnique de Grenoble, 470 pages. Randolph, M.F. & Houlsby, G.T. 1984. The limiting pressure on a circular pile loaded laterally in cohesive soil. Géotechnique, 34(4): 613–623. Randolph, M.F., Hefer, P.A., Geise, J.M. & Watson, P.G. 1998. Improved seabed strength profiling using T-bar penetrometer. Offshore Site and Foundation Behaviour’98, SUT, London, pp. 221–235. Randolph, M.F., Low, H.E., & Zhou, H. 2007. In situ testing for design of pipeline and anchoring systems. 6th Int. Conf. Offshore Site Investigation and Geotechnics, SUT, London, pp. 251–262. Stewart, D.P. & Randolph, M.F. 1991. A new site investigation tool for the centrifuge. Proc. Int.Conf. On Centrifuge Modelling – Centrifuge 91, Boulder, Colorado, pp. 531–538. White D.J., Gaudin, C., Boylan, N. & Zhou, H. 2010. Interpretation of T-bar penetrometer tests at shallow embedment and in very soft soils, Can. Geot. Journal, Vol.47, N◦ 2, pp 218–229 Yafrate, N.J., DeJong, J.T. & DeGroot, D.J. 2007. The influence of full-flow penetrometer area ratio on penetration resistance and undrained and remoulded shear strength. 6th Int. Conf. Offshore Site Investigation and Geotechnics, SUT, London pp.461–468. Zhou, H. & Randolph, M.F. 2009. Numerical investigations into cycling of full-flow penetrometers in soft clay, Geotechnique 59, N◦ 10, December 2009, pp.801–812.

310

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Piezometer installation in deepwater Norwegian Sea T.I. Tjelta Statoil, Norway

J. Strout Norwegian Geotechnical Institute, Norway

ABSTRACT: A deepwater site investigation was planned for the Norwegian Sea Luva Gas Field in 2009, as part of which five piezometers were installed to various depths below seabed in 1300 m water depth. Past experience with piezometer installation to any significant depth below seabed from a surface vessel had proved very troublesome and costly, e.g. in the Ormen Lange field (OTC 18706, 2007). For the Luva piezometers a new and more simple design was adopted. This paper will review the concept of piezometers in general and report from the successful installation of new design piezometers at the Luva field with as-installed depths ranging from 20 m to more than 100 m below seabed. 1

INTRODUCTION

water depth, as reported by Tjelta and Yetginer (2010).

In-situ pore pressure is one of the fundamental parameters required for assessing the strength and stability of soils. Past experience with piezometer installation to any significant depth below seabed from a surface vessel has proved very troublesome and costly, e.g. in the Ormen Lange field (OTC 18706, 2007). Although sophisticated equipment was available, the operational challenges related to installation posed the greatest challenges. This paper presents a new method for the installation of piezometers simplifying long term monitoring of pore pressure in deep water seabed sediments. The piezometers used represent the current state of the art technology (Strout and Tjelta 2007) which have been adapted in this campaign for installation with a seabed drilling unit as part of a long-term monitoring campaign. The basis for the piezometer is a 36 mm diameter hydraulic standpipe, originally designed for installation by a seabed CPT rig or in a traditional drilled borehole. The piezometer system was modified to accommodate installation using a seabed drill rig, leaving the drillpipe (54 mm diameter) in the ground as the standpipe, a novel solution not applied in the offshore industry before. Utilising this method pore pressures can be measured to depths of more than 100 m below mudline (bml) in water depths exceeding 1300 m. The Luva site is located downslope of the Norwegian shelf and has over the last glaciation more than 10 000 years ago received significant sediment input. North of the site there is evidence of major slide activity (Traenadjupet slide, Nyk slide) and in the west there are large diapirs rising more than 100 m above the deepwater plane. The site investigation plan included up to 5 piezometers in up to 1300 m © 2011 by Taylor & Francis Group, LLC

2

MOTIVATION AND BACKGROUND

The continental slope in the vicinity of the Luva field has experienced historical major slide events (e.g. the Storegga slide, the Nyk slide and The Traenadjupet slide). Although slopes are less than one degree, the slope is regional and runs down to 3000 m water depth. Basin sedimentation models combined with hydrogeological models indicated excess pore pressures could potentially explain large scale diapirism in the area, and potential excess pore pressure was considered the major unknown. Previous work has established it is difficult to make a reliable estimate of in-situ pore pressures, and direct measurement is the best approach to eliminate most uncertainties such estimates introduce. The Luva monitoring system was designed following the principles for designing a subsea pore pressure monitoring system described by Strout and Tjelta (2007). Due to the remoteness of the field, local soil conditions, required formation monitoring depths (below seabed) and the installation methods available, the design of the system was based on – autonomous (stand alone) system – hydraulic standpipe with sensors at seabed – penetrated into the soil from seabed or bottom of a borehole – battery and data storage sufficient for several years operation. 3

PUSH-IN PIEZOMETERS

The first prototype of these piezometers was deployed as part of the Ormen Lange subsea geohazard

311

Figure 1. Push-in piezometer instrument package.

monitoring instrumentation. The prototype was based on a standpipe installed by deadweight penetration (e.g. a gravity corer approach). The instrument package was relatively large, weighing approximately 80 kg in water. The second generation was developed in 2006 for the Troll pockmark investigation, and the third generation (current design) was developed for a BP deepwater development. The third generation design has been refined in terms of installation techniques for Luva. The instrument is based on a hydraulic standpipe pre-installed in the seabed with a removable instrument package (Figure 1) small enough to be handled by a modest size ROV. The hydraulic standpipe is fitted with a bespoke wet-mate hydraulic connector at the top, a geotechnical filter tip at depth, and the body of the standpipe can be any suitable pipe including CPT rod (original design) or as in the Luva case a 54 mm OD drill pipe. The original concept for the system is to install the hydraulic standpipes using a standard seabed CPT frame (Figure 2), where maximum penetration depth is a function of CPT frame capacity and soil conditions. Once the standpipe is in place, the instrument package is installed by ROV. Installation is a simple operation: the bottom of the instrument package has an entry cone to guide the male connector on the hydraulic standpipe into the instrument, and the hydraulic connector self-mates under the weight of the instrument package. The instrument package carries dual pore pressure sensors, power supply and data logging systems (for complete redundancy). The unit is autonomous, with power and data storage sufficient for several years of operation. Data is collected by recovering the instrument package to the vessel. The system modification for the Luva installation was to convert the standpipe to use a drill string configured for handling and installation by a seabed drill rig as a combined drilling/standpipe installation operation performed at the seabed. This modification improved © 2011 by Taylor & Francis Group, LLC

Figure 2. Hydraulic standpipe installation using seabed CPT frame to penetrate to target depth.

operational efficiency and allowed installation depths of more than 100 m below seabed to be achieved. 4

PIEZOMETER INSTALLATION USING A SEABED DRILL

A seabed drill was selected as the preferred tool for the work for the Luva geotechnical investigation, (see Osborne, et al. 2010). For the planned piezometer installations this created both a challenge and an opportunity: A challenge since the current piezometer design was based on the use of a heavy seabed CPT unit to push the piezometers into the seabed, but also an opportunity since the seabed drill allowed for much deeper installations than would otherwise have been possible. The installation method developed proved to be more efficient, more reliable and with lower risk for failures. The use of a seabed drill offers the following benefits compared to other installation techniques for offshore piezometers:

312

– local referenced drilling machine on seabed giving good depth control and no relative movements during installation works; – improved handling safety by moving pipe handling to an automated system at seabed (and not on the drill deck); – the experience at Luva was 100% success rate and very time efficient installation, with piezometer standpipes installed in less than 5 hours in 1300 m water depth deck-to-deck. A summary of the five piezometer installation times and depths is presented in Table 1 below.

Table 1.

Piezometer Installation Times and Depths at Luva.

Location ID

Final Tip Depth (m bml)

Time taken deck-to-deck (hh:mm)

Installation time on seabed (hh:mm)

L1009-PZ1 L1010-PZ2 L4005-PZ N1001-PZ V1001-PZ

21.5 85.5 19.5 102.3 24.3

(-)* (-)* 04:30 14:25 06:50

00:50 14:50 01:05 10:40 03:00

* L1009-PZ1 and L1010-PZ2 were installed in a single deployment. Total time deck-to-deck was 20 h 10 min.

Figure 4. Summary plot of all pressure data.

Residual pressures (excess pressures from installation) in the formation surrounding the filter tip of the instrument were still present when the instrument packages were installed; resulting in the capture of dissipation curves (Figure 4). Detailed examination of the pressure plot shows a slight instantaneous pressure rise at connector mating (∼1 kPa), followed shortly after by a large pressure increase due to formation pressure. The excess pressure measured was created by the installation of the hydraulic standpipe, and not due to mating of the connector (theoretically a zero volume displacement connection inducing no additional pressure). The time required for initial pressure build-up and subsequent dissipation is similar for locations 1009PZ1 and 4005-PZ1 (installed at approximately 20m bml) but for 1010-PZ2 (85.5 m bml) the initial rise is slower and subsequent dissipation is also significantly slower. Further, the magnitude of the initial pressure rise is much larger for 1010-PZ2, exceeding the range of the D/A converter resulting in clipping of the signal (flat line portion of the curve). The piezometers are designed to register excess pore pressures from 0 to 200 kPa, but much higher initial pore pressures were present indicating stiffer/harder clay generating higher initial pressures, and significantly lower permeability producing longer dissipation times. The influence of tidal response is also evident in the data (Figure 5). 1009-PZ1 and 4005-PZ1 have similar response both in magnitude and in phase, whereas 1010-PZ2 has a larger magnitude and is shifted slightly in phase (delayed variations compared to the other two instruments). The two similar instruments are installed at similar depths in similar soils, whereas 1010-PZ2 is much deeper and in a different geological formation.

Figure 3. Piezometer location V1001-PZ.

The operational planning and equipment preparation necessary for installation of the piezometers (a new operation) using the drilling, sampling and in-situ testing capabilities of the seabed drill (standard operations) were given high focus during planning stages of the campaign. Required equipment modifications included cross-over adapters to mate the specialised connectors and filter tips of the piezometer standpipe to standard dimension drill rods. Careful consideration was also given to component lengths to allow loading and operation from standard racks on the seabed drill while obtaining acceptable seabed stick-up heights after installation. The Luva piezometers were installed in mid August 2009. One of the completed installations is shown in Figure 3. The instrument logging started when the piezometer instrument package (Figure 1) left NGI workshop approximately one month prior to field installation, hence the horizontal scale in Figure 5 goes from 28 “days since instrument start”.

6 5

DATA RECOVERED AND INITIAL INTERPRETATION

A data collection campaign was performed mid December 2009 to gather data and verify operation of the Luva piezometers. Data was downloaded from three locations: 1009-PZ1, 1010-PZ2 and 4005-PZ1. © 2011 by Taylor & Francis Group, LLC

DISCUSSION OF RESULTS

The use of the seabed drill as the installation platform provided exceptional control for reaching target installation depths (cm accuracy) as well as highly efficient operations by allowing both drilling and sampling, in-situ testing as well as instrument installation in a single frame deployment to seabed. With a seabed drill

313

system, which while instantaneous on the reference side, is delayed on the measurement side due to the permeability of the surrounding material. The amount of fluid flow required is a function of internal system stiffness in the piezometer. The resultant of the measurement is a tidal response with magnitude and phase shift related to permeability and system stiffness. Tidal response observed in the measurement data will be present in piezometer data, it is therefore necessary to interpret data as average values over several tidal cycles. Further, it is necessary to collect data at a sufficient rate to avoid aliasing and misinterpretation due to the tidal variations. Experience from Luva indicates that hourly measurements interpreted as 24 h running averages is adequate for pore pressure measurements in non-dynamic systems.

Figure 5. Detail of tidal variations.

it is conceivable to configure the deployment to allow piezometer installation as an optional final completion of a geotechnical borehole. The subsea piezometer instrument packages deployed functioned perfectly. Deployment and recovery of the instruments by ROV was operationally uncomplicated. The data recovered was complete and was performed by the ROV crew without the assistance of the instrument provider. The data collected indicate that the steady state excess pore pressures present in the soil formations monitored are only moderate. At installation depths of approximately 20 m (2 instruments) the excess pore pressures where about 1–2% of hydrostatic, whereas for the deeper instrument (85.5 m) the excess pressure was about 2% of hydrostatic. The data collected also indicated a significant time for dissipation of excess pressures induced by drilling and installation of the piezometers. Although initial rates varied significantly between the installations, all installations required approximately 5–6 weeks to reach a steady pore pressure state. The form of the dissipation curve is a function of the soil permeability as well as the volume of soil influenced by the installation. The practical implications of the measurements are that offshore dissipation testing in low permeability soils will always require estimates of final in-situ pressures as the time needed for full dissipation will never be practically achievable from a drilling vessel. As the form of the dissipation curve can vary even for similar dissipation times, such estimates will always carry some uncertainty in the estimated in-situ pressure. The only possible means to obtain direct measurement of final in-situ pressures is hence through a long term monitoring program using piezometers. The tidal response artefact is a system response related to the permeability of the soil surrounding the filter tip. The system utilizes differential pressure sensors, where the reference side of the sensor has an uninhibited hydraulic connection to the sea. The measurement side is connected to the filter (and surrounding soil formation). Measurement of pressure by the sensor requires a miniscule volumetric flow into the © 2011 by Taylor & Francis Group, LLC

7

CONCLUSIONS

Strout and Tjelta (2007) have earlier indicated that the use of piezometers is currently the only reliable method for obtaining pore pressure data in low permeability subsea sediments. The Luva measurements confirm that long dissipation times are necessary in low permeability soils, in these deployments up to 5–6 weeks were required. Sophisticated equipment is available, however the challenge has been to develop installation methods which are robust, effective and reliable. The installation approach developed for the Luva site investigation utilised the capabilities of an advanced seabed drilling platform. The seabed drilling platform capabilities included the combination of drilling and direct push operations in a single deployment. This approach proved to be highly successful, with piezometers installed in water depths of 1300 m with total operations time as low as 4–5 hours per installation and all attempts succeeded. ACKNOWLEDGEMENTS The piezometer equipment deployed at Luva were provided to Statoil from BP and we are grateful for the cooperation. REFERENCES Strout, J. & Tjelta, T.I. 2007. Excess pore pressure measurement and monitoring for offshore instability problems OTC paper 18706, Offshore Technology Conference, Houston Texas, May 2007. Osborne, J.J., Halliday, T., Yetginer, A.G. & Tjelta, T.I. 2010. The future of deepwater site investigation: seabed drilling technology ?. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Tjelta, T.I. & Yetginer, A.G. 2010. Luva deepwater site investigation programme and findings. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Yetginer, A.G. & Tjelta, T.I. 2010. Seabed Drilling vs Surface Drilling – A Comparison. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth.

314

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Luva deepwater site investigation programme and findings T.I. Tjelta Statoil, Norway

A.G. Yetginer RPS Energy, UK

ABSTRACT: A deepwater site investigation was planned for Norwegian Sea Luva Gas Field in 2009. The detailed field programme was aimed at resolving all early requirements for regional and local seabed data for both geohazards and foundation studies. Seabed surveys to produce a good terrain model and sub bottom profiling to investigate stratigraphy of shallow soils were operationally combined with a geotechnical equipment package. This package consisted of the PROD seabed based drilling rig and a box corer. During this campaign the PROD rig was also for the first time utilised to install piezometers to more than 100 m depth below seabed. The paper reviews the operational planning, the field work and results from offshore and onshore lab testing for the Luva site investigation programme including considerations on sample quality, correction for sample disturbance and strength at low temperatures (seabed temperature at Luva is below freezing). 1

INTRODUCTION

A deepwater site investigation was planned for Norwegian Sea Luva Gas Field in 2009, Figure 1. The area is remote from shore and from existing field installations and is in most senses characterized as a virgin site at approximately 1,300 m water depth. The detailed field programme was aimed at resolving all early requirements for regional and local seabed data for both geohazards and foundation studies. Seabed surveys to produce a good terrain model and to compare with bathymetry obtained from 3D data sets for both local development area and long export pipeline routes, and sub bottom profiling were operationally combined with a geotechnical equipment package. Field development studies identified a subsea development with a taut moored floater as the most likely solution and included the use of steel catenary risers. Consequently reliable soils information was required both for the moorings, subsea structures and the shallow seabed.

The paper reviews the operational planning, the fieldwork, results from the laboratory testing and briefly looks at special issues like sample quality, correction for sample disturbance and strength at low temperatures. The work was part of a technology qualification programme of seabed drilling technology in deepwater soft clays and hence comparison with surface drilling in terms of operational efficiency and sample quality was also included as part of the qualification programme. 2

2.1 Aim and objectives The main aim and objective of the Luva site investigation was to develop a thorough understanding of the local geology and ground conditions at Luva to assist with the field development. The geotechnical information was primarily to be used for the design of template foundations and the anchors. Further, the Luva licence also wanted to see if seabed drilling was an option for future soil investigations at this site. 2.2

Figure 1. Luva Location Map.

© 2011 by Taylor & Francis Group, LLC

SITE INVESTIGATION PLANNING PROCESS

Initial studies

During the initial planning stages of the Luva site investigation a regional geological desk study and an extensive market study was carried out looking into the various options required and available for deepwater site investigation. The type of seabed surveys and geotechnical services, their areal extent and the depth below seafloor to be investigated were all results from these initial studies.

315

Remotely operated seabed based drilling technology was recommended as the preferred method for conducting the soil investigation at Luva. However, no experience existed within the Luva licence partners on this technology, and it was also considered unproven in the soft clays expected at the site. It was therefore not trivial to include seabed based drilling as the main option for geotechnical work and a technology qualification process was considered necessary to convince Statoil and partners that the present and future technical risk was acceptable. Of the available seabed drilling units Benthic Geotech’s PROD 1 was identified as the lowest risk option due to its demonstrable track record and wide suite of high quality sampling and testing equipment. Further details of this decision making process is presented by Osborne et al. (2010). The geology study focused on available data for the seafloor morphology, geological history, geotechnical properties, slope stability, potential for excess pore pressure and the presence of gas hydrates. Several 3D seismic cubes and 2D high resolution seismic profiles were available to the project together with geotechnical boreholes. 2.3

3 3.1

Geological conditions at Luva

Geotechnical site investigation scope of work

The site investigation scope of work at Luva consisted of the following to satisfy the requirement for both regional geology investigations and foundation studies: – Continuous sampling to 40m below mudline – Continuous cone penetrometer testing (CPT) to 40 m below mudline – Continuous ball penetrometer testing (BPT) to 20 m below mudline, with three cyclic test stages at approximately 5 m, 10 m and 15 m depth – Piezometer installations at several locations and depths from 20 m to 100 m depth below mudline – and boxcore sampling. © 2011 by Taylor & Francis Group, LLC

General

The geotechnical data acquisition using the PROD system was part of a combined geophysical and geotechnical survey programme which proved to be a good decision for this campaign. Although some technical problems were experienced with both equipment suites almost no downtime were logged for the vessel, and the campaign could be completed within scheduled time and budget with all objectives achieved. Soil samples from Luva are soft and potentially more fragile due to the large pressure relief from 1,300 m water depth. Additionally, the selection of PROD1 with its smaller sample diameter (44 mm) raised some concern with respect to sample quality, which became a focus area for the Luva site investigation.

20 km west of Luva there is a large field of diapirs that partly penetrate to the seafloor and rise more than 100 m above. The slope to the east of Luva is covered by furrows indicating downslope flow, believed to be formed during the last glacial period. There is no evidence of past sliding at the site, but evidence of mass movements is found 27 km north of Luva, i.e. the headwall of the Nyk slide, which is part of a massive slide complex including the Trænadjupet slide, believed to be 16000 and 4500 years old respectively. Although this is not recent, it was sufficiently close in time and distance to be further evaluated for active geological processes and together with the diapirs it triggered the need to check for potential in-situ excess pore water pressure at Luva, in the diapir area and at the Nyk slide headwall. Geotechnical data for foundation purposes were needed down to 40m below seafloor. Detailed seabed surveys including swath bathymetry and sub bottom profiler data were also required. 2.4

FIELDWORK AND OFFSHORE OBSERVATIONS

316

3.2

PROD1 operational performance

Following some initial problems with the umbilical, these were mitigated and PROD1 performance exceeded all expectations with more than 700 m of sampling, PCPT, Ball Penetrometer Testing and Piezometer installation conducted within 10 days, distributed over 8 locations 3–30 km apart, in 1300 m water depth. For several locations multiple boreholes were performed and the deepest hole was terminated at 102 m below seafloor. Further details of the PROD performance and a comparison with surface drilling techniques are presented by Yetginer & Tjelta (2010), which concludes that on campaigns where the ground conditions and the local geology are predicted to be relatively uniform and normally consolidated, the seabed drilling technology is likely to provide a more efficient alternative to surface drilling at almost any water depth. 3.3

Drilling, sampling and in-situ testing

Drilling in the soft Luva soils was not a challenge for the PROD1. The borehole was cased for every 1 or 2 rod lengths, and piston sampling carried out to the full depths of 40 m for majority of the boreholes at very good production rates. The operational mode for PROD1 is such that continuous sampling in one borehole and continuous CPT in a parallel borehole is almost equally efficient as composite boreholes with intermittent sampling and CPTs, and provides much more information. Further details of PROD drilling and sampling methodology is presented by Kelleher & Hull (2008). During the offshore programme it was considered that the PROD piston sampler might compromise sample quality because of the following three reasons: – small diameter, – a relatively robust sample core cutter – and the use of a core catcher. The geometry of the sampler head was not considered to be in accordance with recent work as described

– undisturbed and remoulded motorised laboratory vane shear test at the end where the torvane or pocket penetrometer test was carried out – undisturbed and remoulded fall cone tests – unit weight and moisture content determination 4.2

Onshore laboratory testing – NGI

The lab testing programme for the Luva samples onshore consisted of a suite of index tests, consolidation tests and triaxial tests together with geochemistry and geology tests as detailed below: – Index and classification testing – CRS oedometer testing – CAU triaxial compression and extension tests, including SHANSEP tests – Geochemistry including gas content and type – Age dating of samples, and – Special testing such as reconsolidated remoulded direct simple shear tests, thixotropy and x-ray tomography.

Figure 2. Sample cracking at borehole location 4002.

by Lunne et al (2008). Hence slight modifications were proposed to the sampler head offshore, attempting to sharpen the edge of the cutting shoe, removing inner clearance ratio and choosing soft core catchers. The effect of these modifications on sample quality was evaluated and is discussed in Section 5.1.1 below.

Only the following tests and topics will be discussed in this paper: triaxial and CRS oedometer tests with focus on void ratio changes during consolidation as a measure of sample quality (Lunne et al, 1998) and geochemistry tests to explain sample cracking.

3.4 Gassy soils and sample cracking At one of the locations investigated, sample cracking was observed when samples were recovered to deck, indicating presence of gas, (Figure 2). This raised a question about the “true” in-situ properties of the soil and how the “correct” undrained shear strength of the soil could be measured / predicted. The inevitable damage caused by the sample cracking led to extensive work on sample disturbance correction, which is briefly described in Section 5.2 below. 3.5

Piezometer installations

4.3

Offshore installation of piezometers in deep water has always been a real challenge and is not frequently performed. The benefit of knowing in-situ pore pressures as opposed to assuming and “guessing” from geology or estimating them based on CPT or piezo-probe dissipation testing is significant and may warrant the use of piezometer installations. The cooperation between NGI and Benthic produced a very efficient solution for offshore piezometer installation at Luva and the PROD proved an ideal tool for this. Never before have these installation been conducted so efficiently with no loss of equipment, see Tjelta et al (2010). Results from these installations have proved invaluable in removing uncertainty in geological processes, in-situ effective stresses and soil conditions at the site. 4 4.1

LABORATORY TESTING Offshore laboratory testing – Benthic

The following suite of tests was performed on the Luva samples recovered offshore with the PROD1 unit: – torvane or pocket penetrometer test at the bottom end of each sub sample © 2011 by Taylor & Francis Group, LLC

Sample preparation

At the start of the detailed laboratory testing programme there was some uncertainty regarding the best way to prepare the specimens for testing. Two options were considered which were recommended by NGI and Benthic as the preferred method of sample preparation:

317

– Extrude full length sample, i.e. extrude 100cm and cut the required length of specimen for testing – recommended by NGI – Cut the required length of sample inside the polycarbonate liner and extrude specimen for testing, i.e. extrude 10–15 cm – recommended by Benthic and UWA The first one of these sample preparation methods presented the advantage that the full length sample could be inspected prior to selection of the test specimen depth. This is also the standard way in which samples from North Sea site investigations are processed. The second method of sample preparation presented the advantage that more samples could be kept and preserved in liners and the sample is potentially subjected to less disturbance due to the reduction in the length over which the sample is extruded. No comparative testing had previously been performed to investigate the effect of sample preparation on the sample disturbance and hence the decision was made to test a limited number of samples whereby the specimens were prepared based on the two methods described above. The results of these tests indicated that cutting the liner and then extruding the samples potentially

Figure 3. Troll pockmark triaxial test results.

introduced less disturbance to the test specimens and hence the decision was made to perform the remainder of the laboratory testing programme based on this sample preparation methodology. 4.4 Tests for evaluation of temperature effects Seabed temperature at Luva is −1.5◦ C. It is known that temperature has an effect on undrained shear strength of soils and to better understand this effect a number of tests were aimed at undrained shear strength testing at low temperatures. This testing further aimed at investigating if temperature cycles from sampling at seabed until testing in onshore lab had an effect. Consequently some samples were kept at low temperatures throughout while other samples were allowed to reach 20◦ C before taken back to low temperatures and tested. The results of these tests are summarised in Section 5.3 below. 4.5 Tests on onshore block samples to confirm “theoretical undisturbed behaviour” It is considered that high quality undisturbed clay samples may exhibit close to elastic behaviour prior to failure when sheared in a triaxial compression test, ref Berre et al. (2007). This observation is known from testing of high quality block samples onshore, but has only to a limited extent been observed from offshore samples, except in cases where documented high quality samples have been collected (Figure 3). To enable a comparison between onshore work described by Berre et al. (2007) and offshore results from Luva, a discussion of “theoretical undisturbed behaviour” of in-situ soft clay took place within the Luva project team. Following these discussions, a limited number of tests were performed on onshore block samples from the Onsøy research test site south of Oslo to confirm the method described by Berre et al. (2007) for correction of soil sample disturbance was relevant for Luva soils, see further discussion in Section 5 below. © 2011 by Taylor & Francis Group, LLC

Figure 4. Sample quality comparison – Surface drilling vs seabed drilling.

5

SAMPLE QUALITY

5.1 Assessment of sample disturbance 5.1.1 Effect of modifications to cutting shoe design The sampling shoe for the PROD hydraulic tethered piston corer is more robust with a larger thickness and a larger thickness/diameter ratio than the standard 3” thin wall sample tube used in downhole drilling mode by most site investigation contactors. To check if this had an effect on sample quality, the sample shoe was modified and a limited number of samples taken with these modified shoes. In addition it was possible to make a comparison with test results on thin walled 3” piston samples obtained by Statoil using vessel based drilling at the Luva site in 1997. Figure 4 shows e/eo for CAU and CRS tests from the 1997 Ad Notam site investigation using standard downhole tools and the 2009 Luva site investigation using the standard PROD sampler. There is a lot of scatter in the data, but according to the authors the results indicate a trend for better quality using the thin walled piston tube samples. To enable a theoretical comparison between the standard PROD shoe, modified shoe and a standard 3” piston tube FEM modelling was used to look at the strains developing in the sample during penetration. The conclusion lends itself towards an influence from the thicker shoe of PROD1 and it may be possible

318

Figure 6. Luva sample correction.

this produces reduced undrained shear strength values compared to what an ideal undisturbed sample would give. If samples are disturbed and the theoretical behaviour of the soil at the actual conditions fulfil the criteria described by Berre et al. (2007) there is a potential to correct for some of the damage this disturbance has on strength and stiffness. This correction has been attempted for the Luva samples and a plot summarising the procedure is shown in Figure 6. The methodology described by Berre et al (2007) indicates that disturbance from sampling can be corrected for, however a number of reservations are made and ideally one should use the methodology only to see the potential higher shear strength that may be achieved by improved sampling. But for Luva blocksampling is not achievable, nor is it possible to perform high quality sampling by any method if gas and/or pressure relief is causing sample disturbance. However, from a comprehensive comparison of geology, soil index parameters and stress history it was considered relevant to apply the correction procedure for the Luva soil samples which are disturbed from both the presence of gas and the piston sampling.

Figure 5. Luva Sulphate test results.

to achieve a higher sample quality with a modified, sharpened cutting shoe. Also the effect of a core catcher should be further evaluated for soft clays. In conclusion, though, it is seen that both sampling methods used at Luva so far have produced samples of variable quality and sample behaviour is clearly showing signs of disturbance. 5.1.2 Effect of gas coming out of solution Following the observation of cracking in samples below 17 m depth at one of the sites, borehole location 4002, efforts were made to investigate if this was due to shallow gas and whether this was an anomaly relevant for this location only. Detailed investigation of shallow seismic, geochemistry testing and inspection of all in-situ tests from all locations revealed the following conclusion: gas is present in a minute amount dissolved in pore water at all locations, but the flux from depth towards mudline at borehole location 4002 is higher. This is shown by the sulphate test results (Figure 5) indicating all methane above 17 m at this location is consumed by bacterial activity, whilst at the other three sites the bacterial activity consuming methane reaches down to approximately 40 m below mudline. Consequently it is concluded that samples taken from below 40 m at the other locations would have experienced similar cracking as observed in borehole 4002 had the boreholes been extended below this depth. 5.2 Correction for sample disturbance Testing of Luva samples and investigation into sample disturbance causes and effects, as described above, has shown that sampling alone results in disturbance, and combined with minute amounts of gas coming out of solution when samples are recovered to deck, © 2011 by Taylor & Francis Group, LLC

5.3 Temperature effects The temperature tests confirmed that the undrained strength for soft clay samples tested at low temperatures (2◦ C) are higher than for samples tested at room temperature (20◦ C). No significant effect was seen between samples kept at low temperatures throughout, compared with those samples which were allowed to reach room temperature before being tested at low temperature, indicating the most important contribution to increased strength comes from temperature at the time of testing, where lower temperatures give higher strength. 5.4

Results of third party verification testing

A limited number of samples were also tested independently outside NGI indicating the procedures used at NGI were relevant and results are comparable.

319

5.5

Discussion of results

Offshore sampling is believed to produce some sample disturbance which has an impact on laboratory testing results, most notably on strength and stiffness. To understand the degree of disturbance and the effect of it, numerous attempts have been made to look into many of the contributors to sample disturbance at Luva, including effect of temperature during storage and transport as well as during testing, piston sampler geometry and effect of gas coming out of solution. But since it is difficult to obtain high quality samples in some soils and sites, the idea of attempting correction for the induced disturbance has been pursued at Luva. The correction procedure applied here cannot be used on a general basis. The stress history, aging and soil structure has an important impact on the stress path and hence the assumption of a close to linear elastic behaviour as presented in Figure 3. For a full description of reservations and limitations, see discussion in Berre et al (2007). 6

CONCLUSIONS

The soil investigation at the deepwater Luva field offshore Norway has confirmed seabed drilling is a competitive and efficient method of providing soils properties in deep water and for understanding in-situ soil conditions. Although this is an excellent method for in-situ testing including installation of piezometers, it is considered that the sample quality can be further improved. The comprehensive testing of samples from Luva has proved soil sample disturbance can be partly corrected given geological conditions are similar to a good reference site and stress history is relevant.

© 2011 by Taylor & Francis Group, LLC

It is also seen that undrained shear strength for the low temperature at Luva seabed is higher than what would normally be advised based on testing at room temperature. REFERENCES Berre, T, Lunne, T, Andersen, K.H., Strandvik, S. and Sjursen, M. 2007. Potential improvements of design parameters by taking block samples of soft marine Norwegian clays. Canadian Geotechnical Journal, No 44: 698–716. Kelleher, P. & Hull, T. 2008. Quality assessment of marine sediments recovered with a hydraulically tethered piston corer. Offshore Technology Conference OTC 19687, Houston, May. Kelleher, P., Low, H. E., Jones, C., Lunne, T., Strandvik, S. and Tjelta, T. 2010. Strength measurement in the very soft near seabed sediments. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Lunne, T., Berre, T.V. and Strandvik, S. 1998. Sample disturbance effects in deep water soil investigations. Conference on Offshore Site Investigation and Foundation Behaviour, SUT, London. Lunne, T., Tjelta, T.I., Walta, A. and Barwise, A. 2008. Design and testing out of Deepwater Seabed Sampler. Offshore Technology Conference 19290, Houston, May. Osborne, J.J., Halliday, T., Yetginer, A.G. & Tjelta, T.I. 2010. The future of deepwater site investigation: seabed drilling technology ?. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Tjelta, T.I. & Strout, J. 2010. Piezometer installation in deepwater Norwegian Sea. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Yetginer, A.G. & Tjelta, T.I. 2010. Seabed Drilling vs Surface Drilling – A Comparison. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth.

320

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Investigations into novel shallow penetrometers for fine-grained soils Y. Yan, D.J. White & M.F. Randolph Centre for Offshore Foundation Systems, The University of Western Australia, Perth, Australia

ABSTRACT: This paper describes the initial evolution of two novel site investigation tools – shallow penetrometers – that are designed to measure the properties of surficial seabed soils, and in particular the axial resistance between pipelines and fine-grained soils. The shallow penetrometers are toroidal and hemispherical in shape, and are pushed vertically and rotated about the vertical axis. This design allows soil-interface shearing to be performed without the inconvenience of end effects. The paper describes numerical and centrifuge modelling results which show the resistance to purely vertical or torsional loading on undrained clay. The results of a cyclic torsional test using the toroidal penetrometer are shown, including detailed measurements of pore pressure build-up and dissipation. The new tools show promise and the results highlight the need to develop effective stress-based interpretation methods, to enhance the utility of these tools.

1

INTRODUCTION

derived using classical plasticity, which can be expressed as (Merifield et al. 2009):

The toroidal and hemispherical penetrometers described in this paper are new in situ site investigation tools that aim to provide an improved understanding of the interaction between a pipeline and the near-surface soft sediments commonly encountered in deep water. These shallowly-penetrating penetrometers are a concept originally proposed by the authors to allow axial pipe-soil forces to be measured without the complication of end effects that arise if a conventional pipeline segment is used, translating axially. The toroidal device (Fig. 1a, 1b) comprises a frame with an instrumented model toroid that can be driven in the vertical and torsional directions whilst the corresponding loads and displacements are recorded. A hemispherical penetrometer, or ‘hemiball’ with the same outer diameter is shown in Fig. 1c. Shape effects raise the bearing capacity of the three-dimensional shaped penetrometer compared to a plane-strain pipe, which weakens the direct analogy that can be made between toroid-soil interaction and pipe-soil interaction. A previous finite element (FE) analysis study (Yan et al. 2010) has established the relevant response of a fully rough toroid with an aspect ratio, L/D, of two on homogeneous undrained soil. The choice of aspect ratio balances the desire for a compact instrument against the need to reduce interference across the toroid. Martin & Hazell (2005) showed that the interference from a given spacing is greatest for homogeneous soil and higher for rough footings. This indicates that the results from our FE analyses represent a conservative estimate of the degree of interference found in general conditions. The conventional approach to estimate the vertical collapse load of a shallowly-embedded pipe on undrained clay is based on fundamental solutions © 2011 by Taylor & Francis Group, LLC

321

where NcV and NswV are factors linking the shear strength, su , and soil effective self-weight, γ
, to the bearing capacity. Strength and self-weight factors were found to be independent of γ
D/su over the range 0 < γ
D/su < 3. Rigorous plasticity solutions exist for predicting the bearing capacity factor, NcV of a wishedin-place pipe (Randolph & White 2008). ABAQUS FE analyses (Merifield et al. 2009) have shown that soil heave leads to a modest increase in the NcV factor, but a value for NswV that is some 50% greater than would be estimated on the basis of Archimedes’ principle and the nominal penetration, w. An analytical solution can be derived for the undrained torsional resistance of the toroid. For soil with linearly-increasing strength with depth, and a fully rough toroid surface, the torsional resistance is:

where cosφ = 1-2w/D; Acontact = 2πDLφ; su0 is the shear strength at the surface and k is the strength gradient with depth.

Figure 2. Definition of notation for geometry of pipes and hemi-balls (left) and toroid (right).

and the associated shearing and downdrag of the seabed soil. Also, the contact area may be enhanced by heave around the shoulders although this paper ignores the heave effects and is concerned only with wished-in-place (WIP) conditions. An alternative approach for interpreting the torsional resistance is to model the interface behaviour using a simple friction coefficient, which leads to a relationship between the torsional resistance and the vertical load, namely the ratio T /VL, which depends on the embedment of the hemisphere or the embedment and geometry of the toroid. This method is favoured because it is amenable to extension into effective stress terms, via the inclusion of an excess pore pressure ratio. For slow, fully drained movements, the frictional approach is clearly preferable to a method based on undrained strength. The inclusion of pore pressure transducers on these shallow penetrometers, as shown in Figure 1, permits pore pressure effects to be identified, allowing interpretation in an effective stress framework. This paper examines the functionality of the new penetrometers through a set of centrifuge model tests and associated numerical studies, focusing on toroidsoil interaction during penetration and torsional shearing. The numerical analyses provided a basis for linking the undrained penetration resistance to the soil strength and the corresponding forces on a pipeline penetrated into the same material. While the focus of this paper is on the description of the new penetrometers, the penetration response of a model pipeline is also shown, to illustrate how the shallow penetrometers might be used for assessing the interaction between pipelines and the seabed. Although the basic interpretation framework for vertical penetration is based on an undrained total stress analysis, these results highlight the influence of pore pressure effects on the subsequent axial shearing at the penetrometer-soil (or pipe-soil) interface. Figure 1. Shallow penetrometers (a, b) toroid (c) hemi-ball.

2 To use this approach, linking su with T , it is necessary to recognise that the response is affected by any changes in strength around the toroid-soil interface due to the embedment (or pipelaying) process, © 2011 by Taylor & Francis Group, LLC

2.1

METHODS Model toroid and pipe

The model toroid employed throughout this test program was a 1:25 scale model of a potential version that

322

Table 1. Toroid and pipe dimensions (model / prototype scale). Toroid

Pipe

Lever arm, L (mm) 32/800 – Diameter, D (mm) 16/400 20/500 Vertical velocity, v (mm/s) 0.1/2.5 0.125/3.125 Range of interface shear rates vaxial = ωL vaxial * (expressed as vaxial , mm/s) 0.001-1 0.01-1 Interface shear length (mm) ±100/±2500 ±30/±750 * Varied during tests. ω: angular velocity (radians/s) ˆ Velocity shown only in model scale units

could be deployed in the field. The toroid was fabricated from aluminium and was attached to a loading arm that fits into a rotary actuator, which applies combinations of vertical and torsional loads to the model over a wide velocity range (Watson 1999). The rotary actuator is housed within an existing two-directional actuator, attached to a strongbox. The toroid and pipe dimensions are given in Table 1. 2.2 Test procedure All tests were performed at an acceleration of 25 g in the fixed beam geotechnical centrifuge at the University of Western Australia, which is described in detail by Randolph et al. (1991). The toroid was subjected to a cyclic torsional movement under displacement control before it touched the soil sample surface. This provided a correction for the modest torque applied to the toroid by the compliance of trailing wires from the PPTs. It also provided a correction for the PPT readings that arises from the curved water surface created by the radial direction of the acceleration field. The toroid or pipe was then penetrated to a predetermined position at a rate fast enough for an undrained response. Once installed, the toroid or the pipe was unloaded to represent a realistic pipeline operating weight. After a short period of consolidation, during which the dissipation of excess pore pressure at the toroid-soil or pipe-soil interface was monitored, a sequence of twisting of the toroid or axial sweeping of the pipe was applied at rates spanning a wide range, as summarised in Table 1. The toroid or pipe was then extracted at the same rate as the penetration. After each test the centrifuge was halted and the penetrometer or pipe was cleaned to remove any soil resting on the crown. Throughout the test, the pore pressure response, total vertical load and torque (toroid) or axial force (pipe) were recorded with the history of the vertical and rotational (toroid) or axial (pipe) movement.

Figure 3. Typical centrifuge T-bar test (a) penetration resistance and (b) undrained shear strength profiles.

NT = 10.5 is generally adopted for converting the measured unit bearing capacity, q to the undrained shear strength, su in the test. This constant value neglects the influence of soil buoyancy and the changing failure mechanism of soil flow around the bar. The buoyancy creates an additional component of penetration resistance that can be significant for near-surface soft soils. White et al. (2010) described an improved technique to account for these effects in the interpretation of T-bar penetrometer resistance at shallow embedment. The results of a T-bar test interpreted using the constant factor (NT = 10.5) and this more refined method are shown as Figure 3. Linear undrained strength profiles have been fitted for back-analysis of the toroid and pipe test data, as indicated in Figure 3.

2.3 Soil strength characterization A T-bar penetrometer, 5 mm in diameter (125 mm at prototype scale), was used to assess the strength profile of the clay samples. A constant bearing factor, © 2011 by Taylor & Francis Group, LLC

2.4 FE analyses of undrained penetration A set of small strain FE analyses to simulate the initial penetration of the pipe and toroid in the particular

323

linearly-fitted strength profiles shown in Figure 3 were carried out using the software ABAQUS (Dassault Systèmes 2009). The footing geometry represents the prototype size of the pipe and toroid, as given in Table 1. The radius and depth of the soil mesh extend 6.25D from the toroid or pipe axis. The unit weight of the soil was taken as 6 kN/m3 . Analyses were conducted for wished-in-place embedments of w/D = 0.1, 0.2, 0.3, 0.4 and 0.5. The soil was modelled using a linear elastic perfectly plastic constitutive law defined by the undrained Young’s modulus (Eu ), maintaining Eu /su of 500 and Poisson’s ratio (ν) equal to 0.49. Failure was according to the Tresca criterion, with the maximum shear stress limited to the undrained shear strength. Axisymmetric conditions were imposed using the equivalent mesh discretisation as for the cross-section of the threedimensional mesh described in a previous study for homogeneous soil (Yan et al. 2010). 3

Figure 4. Installation and pullout resistance of toroidal penetrometer and pipe.

FE AND CENTRIFUGE RESULTS

3.1 Vertical penetration of toroid and pipe Figure 4 summarizes the penetration and extraction resistance, V , from two toroid and two pipe tests, along with the FE results for the toroid and pipe resting on the non-homogeneous soil. Two strength profiles were used to assess the penetration resistance, including the fitted profile for the shallowest 2 m and a bilinear strength profile based on the two linear segments shown in Figure 3. Using the bilinear approach, the FE results for the smooth case generally fall within 20% of the measured data. The largest discrepancy is evident for the pipe case, which shows lower resistance than is calculated. This discrepancy could indicate downdrag of soft surface sediments, which is not considered in the wished-in-place small strain FE analyses, but it may also simply arise from the difficulty in accurately establishing the strength profile within the upper few millimetres (at model scale) of the sample. A simpler depiction of the various measurements of penetration resistance is to compare the nominal bearing pressure, V /Anom , for all three devices – T-bar, pipe and toroid – in the top 0.2 m of the seabed, as shown in Figure 5. This form of comparison highlights the potential for direct interpretations from in situ penetrometer test data to design values for pipelines. The T-bar resistance was observed to rise most rapidly, reflecting its smaller diameter and therefore shallower depth of penetration required to reach deep flow conditions, compared to both the pipe and the toroid. The toroid penetrometer and the pipe show a tight band of resistance at shallow embedment. At greater depths, the pipe resistance is 85% of the toroid resistance. This slightly higher penetration resistance for the toroid is partly linked to interference across the toroid, which has been assessed to create an increase of ∼5% compared to the plane strain pipe at the deepest embedment ratio of w/D = 0.5 (Yan et al. 2010). An additional source of discrepancy is the slightly © 2011 by Taylor & Francis Group, LLC

Figure 5. Penetration resistance of penetrometers and pipe.

larger size of the pipe, giving it a lower normalized embedment, w/D, for a given value of w.

3.2

Cyclic torsional test using toroid

Tests have been conducted in a variety of conditions to investigate axial pipe-soil interaction and torsional toroid-soil interaction. Figure 6 shows a typical example, denoted toroid test 05. This test involved penetration to an embedment of w/D = 0.28, followed by a reduction of the vertical load by a factor of 2 (Figure 4), then 6 cycles of torsional movement at various speeds under constant vertical load (Table 2). The movements are expressed both as the angle rotated through and the axial displacement of a point at the centre of the solid cross-section of the toroid (i.e. at radius L from the axis of rotation). The trajectory of the toroid is shown by this axial displacement compared to the invert embedment (Figure 6a). The time history of this movement is given in Figure 6b. The toroid settled significantly throughout the testing sequence, reaching a final embedment of more

324

Table 2.

Sequence of torsional movements (toroid test 05).

Cycle

Rate, vaxial mm/s (model)

Angle degrees

Distance mm (model / prototype)

1 2 3 4 5 6

0.001 0.01 0.1 1 1 0.01

±4 ±8 ±180 ±180 ±180 ±8

±2/55 ±4/110 ±100/2500 ±00/2500 ±100/2500 ±4/110

Figure 7. Simple normalisations of torsional response.

Figure 6. Multi-stage torsional test using toroidal penetrometer.

than twice the initial value. It appears that the torsional movement generated additional settlement. This is at least partly due to consolidation, as evidenced by the time histories of excess pore pressure generation and dissipation measured at the four PPTs located around the invert of the toroid (Figure 6c). The PPTs capture even the modest excess pore pressures present in this test, and there is excellent consistency between the 4 readings. A clear dissipation record is evident during the consolidation stage that follows the initial embedment. This provides a basis for assessments of the coefficient of consolidation, cv . Also, there is © 2011 by Taylor & Francis Group, LLC

positive excess pore pressure generated at the start of each torsional sweep, which decays during the sweep. The faster sweeps, later in the test, appear to generate higher excess pore pressure. The steady vertical load, V , is also shown in Figure 6c, illustrating the accuracy of the load control system. Two simple normalisations of the torsional response are shown in Figure 7. A total stress analysis, using the in situ undrained strength at the toroid invert for normalisation (T /su,invert Acontact L), is given in Figure 7a. A friction factor interpretation is shown in Figure 7b. This expresses the torsional resistance as a friction coefficient, assuming that the entire vertical load acts at a radius of L from the toroid axis (i.e. T /VL). Figure 7a includes a comparison with an analytical calculation of the normalised torsion based on a shear stress acting on the toroid equal to the in situ undrained strength (varying with depth). This is equivalent to an ‘α-method’ interpretation (where the interface shear stress is αsu ) using α = 1. The steady values of T /su,invert Acontact L during each cycle are collated in Figure 8. The measured data indicate that a lower shear stress than su acts at the toroid-soil interface, with the back-calculated range of α decreasing from ∼0.57 to ∼0.25 through the 6 cycles (Figure 8). This variation could be interpreted as a reduction in the soil strength through remoulding, although there appears to be minimal degradation in the first 3 cycles, despite the 212 mm of cumulative

325

design. Small-scale pilot versions of these ‘hemiball’ and toroidal devices have been fabricated and the results from a centrifuge modelling study of the toroidal penetrometer are presented, along with supported numerical results. The application of these devices requires interpretation methods to be derived, to extract pipe-soil and soil parameters from the measured behaviour. Results from centrifuge modelling illustrate the potential of these devices, but also highlight the need for effective stress-based interpretation techniques to be developed. REFERENCES

Figure 8. Back-analysis of α-values during torsional test.

shearing. An alternative interpretation is to link the decrease in α (or operative undrained strength) with the rise in excess pore pressure evident during the later, faster cycles (Figure 6). A known build-up of pore pressure can be used to derive an associated reduction in undrained shear strength. A more direct interpretation approach is to assess the corresponding change in effective stress. This is particularly convenient for this form of failure mechanism – interface shearing – since there is only one failure plane and the total stress applied to this plane is known, via V (at least approximately). The friction factor normalisation (Figure 7b) is consistent with this approach. The ratio T /VL remains approximately constant through the first 3 cycles, when minimal excess pore pressure was generated. During the final few cycles this ratio drops, which is consistent with the effective stress at the toroid-soil interface reducing due to a build-up of excess pore pressure (Figure 6c). Further analysis of this test, and others, is underway to investigate an effective stress framework for this behaviour. 4

Dassault Systèmes. 2009. “Abaqus analysis users’ manual.” Simular Corp, Providence, RI, USA. Martin, C. M. & Hazell, E. C. J. 2005. “Bearing capacity of parallel strip footings on non-homogeneous clay.” Frontiers in Offshore Geotechnics: ISFOG 2005, London. Merifield, R. S., White, D. J. & Randolph, M. F. 2009. “The effect of surface heave on the response of partiallyembedded pipelines in clay.” ASCE Journal of Geotechnical and Geoenvironmental Engineering, 135(6), 819–829. Randolph, M. & White, D. J. 2008. “Upper-bound yield envelopes for pipeline at shallow embedment in clay.” Gèotechnique, 58(4), pp. 297–301. Randolph, M. F., Jewell, R. J., Stone, K. J. L. & Brown, T. A. 1991. “Establishing a new centrifuge facility.” Proceedings of the International Conference on Centrifuge Modelling – Centrifuge 91, Broulder, Colorado, 2–9. Watson, P. G. 1999. “Performance of skirted foundations for offshore structures,” PhD thesis, The University of Western Australia. White, D. J., Gaudin, C., Boylan, N. & Zhou, H. 2010. “Interpretation of T-bar penetrometer tests at shallow embedment and in very soft soils.” Canadian Geotechnical Journal, 47(2), 218–229. Yan, Y., White, D. J. & Randolph, M. F. 2010. “Penetration resistance and stiffness factors in uniform clay for hemispherical and toroidal penetrometers.” ASCE International Journal of Geomechanics, Accepted for publication.

CONCLUSIONS

This paper has described a new class of shallow penetrometers suited to the characterisation of nearsurface seabed sediments, principally for pipeline

© 2011 by Taylor & Francis Group, LLC

326

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Seabed drilling vs surface drilling – a comparison A.G. Yetginer RPS Energy, UK

T.I. Tjelta Statoil, Norway

ABSTRACT: A deepwater site investigation was planned for the Norwegian Sea Luva Gas Field in 2009. Following an extensive market study, remotely operated seabed based drilling technology was chosen as the preferred method for conducting the soil investigation. The scope of work for the Luva Site Investigation was successfully completed. This paper evaluates the effectiveness of seabed drilling technology in terms of productivity, safety and other commercial considerations when compared with conventional drilling techniques.

1

INTRODUCTION

The number of deepwater geotechnical site investigations required each year continues to increase. The safety, effectiveness and economic feasibility of conventional “tried and tested” sampling and in-situ testing techniques, originally designed for deployment in comparatively shallow waters, has become questionable in harsher and deeper water environments. This needs to be quantified in terms of cost, productivity and safety. Over the past decade innovative solutions have been developed by the sub-sea industries to overcome engineering difficulties encountered in deepwater environments, particularly in the fields of ROV and autonomous system technologies. Additionally, throughout the past decade there has been a move for mainstream mineral and petrochemical exploitation into deeper water and as a consequence there have been technological advances. However, until recently few of these innovations have been adopted for high quality geotechnical data acquisition. Geotechnical site investigation information was required at license PL218 Luva in the Norwegian Sea in 1,300 m water depth in order to develop a thorough understanding of the ground conditions and to assist with the development of the field. In this type of water depth using a standard vessel-mounted drilling system for a geotechnical site investigation would require the following: – aluminium pipe to be deployed due to the weight restrictions on the derrick – high levels of pipe handling and long periods of time to run and pull the drill-pipe to reach the sea bottom, which would restrict the available weather window for borehole drilling. © 2011 by Taylor & Francis Group, LLC

327

The safety, effectiveness and economic feasibility of such a vessel-mounted, conventional drilling system, historically designed for deployment in comparatively shallow waters, becomes questionable in harsher and deeper water environments such as Luva. In particular where the soil investigation campaign is designed for a floating production system with seabed templates and riser bases the requirement for soil information (samples and in-situ testing) is often limited to 30–40 m depth below seabed. It is then questionable, for example in the case of Luva, if running 1.3 km pipe to drill such a shallow borehole is good practice, from an HSE perspective (where pipe handling presents a major risk element), as well as from an economic and sample quality point of view. Following an extensive market study, remotely operated seabed based drilling technology was chosen as the preferred method for conducting the soil investigation at Luva. However, the concept of remotely operated seabed drilling was not regarded as “qualified technology” at the time of the market study as such systems had not yet been used in the North Sea/ Norwegian Sea, and therefore seabed drilling was subjected to a Technology Qualification Plan (TQP) as per DNV-RP-A-203 recommendations and internal Statoil guidelines. The findings of the 2009 Luva site investigation form the basis of the Seabed Drilling Technology Qualification. In addition to the results of the 2009 campaign, previous site investigation information is available at the Luva field where the drilling and testing was performed from a dedicated geotechnical drilling vessel in 1997. Based on this additional information, the effectiveness of the chosen seabed drill for the 2009 Luva site investigation was evaluated against surface drilling performance and production rates.

– – – –

This paper will present the findings of this evaluation in terms of productivity, safety and other commercial considerations. 2

MOTIVATION AND BACKGROUND

2.3

The main advantages associated with the use of a seabed based drilling system, and hence the motivation for initiating a Technology Qualification Plan for Seabed Drilling Technology were:

Improved efficiency

Seabed drilling systems offer improved levels of efficiency compared with vessel-mounted drilling systems when operating in deep water and this is primarily due to the following:

2.4

Improved sample and data quality

As the remotely operated seabed drill is positioned on the seafloor during drilling operations it is unaffected by vessel movements and the drill head can be considered to be controlled completely independently of the vessel, which results in: – very accurate detection of mudline – significant improvement in borehole depth control (where some operators claim mm accuracy) © 2011 by Taylor & Francis Group, LLC

Increased flexibility of operation

When conducting a site investigation campaign utilising a seabed drill, operations can take place adjacent to offshore structures at a lower risk as the deployment vessel can be moved away once the drill is deployed.

– Reduced launch and recovery times for the drilling equipment, (the vessel mounted system requires the drillstring to be “built” and “de-constructed” whereas the remotely operated seabed tool is lowered to the seafloor and recovered by its umbilical). – Reduced tool run times (round trip, in and out) both for sampling and in-situ testing. Instead of tools being run from deck to seafloor and back to deck, all sampling and in-situ testing equipment is launched from and recovered to the seabed drill, which is positioned on the seafloor. – Lower levels of weather sensitivity during operations, which would allow utilisation of shorter weather windows. The most weather sensitive part of the operations with seabed drilling units is during launch and recovery; at all other times the drill can be left on the seafloor, unless there are issues with the vessel station-keeping. 2.2

HSE initiative

One of the most significant advantages of a remotely operated seabed drilling system is the elimination of the human exposure during pipe and tool handling activities, which would otherwise take place on the drill deck. Currently a high percentage of all reported major injuries offshore are associated with drilling operations. A remotely operated seabed drilling system would all but eliminate such exposure and hence the risks associated with such activities. In addition, the remoteness of the vessel from the drilling system also reduces the risks associated with presence of shallow gas.

1 Improved levels of operational efficiency compared with a vessel-mounted drilling system in deep water, which results in significant cost savings. 2 Improved level of drill head control and borehole depth accuracy, which results in better sample and data quality. 3 HSE initiative associated with moving the drilling operations down to the seafloor and away from the human interface. 4 Increased flexibility of operation. 5 Increased choice of systems and contractors for operating company. 2.1

potentially reduced levels of sample disturbance successful recovery of very soft sediments more accurate in-situ test results very accurate sampling of shallow soils.

328

2.5

Increased choice/competition in the offshore deep water site investigation market

Currently, there are a limited number of geotechnical site investigation contractors who are able to operate in 1,300 m water depth. It is believed that once the remotely operated seabed drilling technology has been qualified these systems would be able to compete with vessel mounted drilling systems in deep water, and due to their improved efficiency and data quality they would significantly improve competition levels in the offshore site investigation market, providing operating companies with an increased number of options for site investigations in deep water. 3 3.1

SEABED DRILLING VS SURFACE DRILLING: PERFORMANCE COMPARISON Comparison for deep water operations

In addition to the results of the 2009 campaign at Luva previous site investigation information is available at the Ad Notam test site where the drilling and testing was performed from a dedicated geotechnical drilling vessel in 1997. The Ad Notam boreholes are positioned in similar water depth (1270–1300 m) and in similar soil conditions to Luva at a distance of approximately 1.4 km from the Luva platform location and 5.5 km from the Luva main template location. The efficiency of the chosen seabed drill operation at Luva is therefore considered to be directly comparable with the surface

Table 1. Time taken per operation in 1,300 m water depth – Surface Drilling in soft soil conditions.

Table 2.

Surface Drilling vs Seabed Drilling – Comparison. Time per borehole (deck-to-deck) (days)

Time taken per operation Activity

Minimum Average Maximum (hh:mm) (hh:mm) (hh:mm)

Lowering & recovering 00:55 the seabed frame Running and pulling pipe 05:08 Sampling/CPT operation 00:59

01:00

01:10

05:47 01:38

07:15 03:09

drilling performance and production rates at the Ad Notam test site. The site investigation campaign at Ad Notam consisted of three (3) boreholes which were drilled to 5.0 m, 19.0 m and 14.7 m depth below mudline. The intended borehole coverage at these three locations was 100%, i.e. it consisted of continuous sampling and CPT profiling to target depth. The average, minimum and maximum times taken to lower the seabed frame, to run and pull pipe and to recover the seabed frame are presented below in Table 1 where the information was extracted from the daily progress reports of the Ad Notam site investigation. In a similar fashion, average, minimum and maximum sampling & testing operation times have been compiled from the drilling logs, which are also presented in Table 1 below. The time taken for each sampling and CPT operation has been defined as the time spent between the finish of testing/sampling at a given depth, until the start of testing/sampling at the next depth. Based on this information, it is possible to construct artificial 40 m continuous sampling, 39 m continuous CPT and 40 m composite boreholes with 5 m cycles (3 m CPT + 2 × 1 m samples) and estimate how long it would have taken to complete these boreholes if they were constructed in downhole mode utilising conventional drilling techniques. It is then possible to compare these times with the Luva seabed drill performance directly, where the scope of work consisted of four (4) approximately 40 m continuous sampling boreholes and four (4) 40m continuous CPT locations (Table 2). The values presented in Table 2 for composite borehole construction demonstrate that seabed drilling technology is 3–5 times more efficient in deep water soft soil conditions compared with surface drilling, where only shallow (up to 40 m) soil information is required. For site investigations where continuous sampling or continuous CPT profiling is required, the efficiency of the seabed drill is much more pronounced. A seabed drill is capable of performing a 40 m continuous sampling borehole 4–10 times more efficiently than a vessel-mounted system. The same applies for continuous CPT profiling where the seabed drill can perform 6–8 times more efficiently, especially in cases where it © 2011 by Taylor & Francis Group, LLC

Surface Drilling

Seabed Drilling

Activity

Min

Ave

Max

Min

Ave

Max

40 m cont. sampling 39 m cont. CPT 40 m comp. borehole

2.2

3.3

5.8

0.5

0.5

0.6

1.1

1.5

2.4

0.2

0.2

0.3

1.5

2.2

3.7

0.6*

0.7*

0.8*

* The 40 m composite borehole times for the seabed drill consist of (1) time taken for 40 m continuous sampling borehole, (2) time taken for 40 m CPT, plus (3) 20 min additional time allowed for “bumping over” between the two locations. Compared to a composite borehole performed by surface drilling the seabed drilling “composite” boreholes presented above hence provide twice as much information.

Table 3. Surface drilling site investigation data considered.

Campaign Title

Year of Campaign

Water Depth (m)

Number of boreholes

Project A Project B Project C Project D Ad Notam Project E Project F Project G Project H Project I

1997 1997 1997 1997 1997 2002 2005 2006 2006 2008

966 532 848 1351 1270–1298 1124–1622 196–200 360–371 91–114 109–111

2 2 1 1 3 2 19 27 11 9

329

may be possible to continue adding rod without having to drill out and case the hole. The minimum time estimate presented above for continuous CPT profiling with the seabed drilling technology is based on information from a location where it was possible to push to close to target depth by continually adding rods. The maximum duration presented in Table 2 for the vessel mounted system is based on information from a location where the hole was drilled out and cased and CPT operations subsequently continued to target depth.

3.2

Comparison of performance with water depth in soft soil conditions

Operational information from site investigations conducted by surface drilling in various water depths were analysed in an attempt to identify the water depth where seabed drilling becomes more efficient than surface drilling. The site investigation campaigns considered during this study are presented in Table 3, which comprised a total number of 77 boreholes in

Figure 1. Sampling/Testing Operation Time with Water Depth – Surface Drilling.

Figure 3. CPT Progress Rate with Water Depth – Comparison for a 40 m borehole.

Figure 2. Sampling Progress Rate with Water Depth – Comparison for a 40 m borehole.

Figure 4. Composite Borehole Progress Rate with Water Depth – Comparison for a 40 m borehole.

various water depths, in addition to the Luva boreholes of 2009. Based on information available from these site investigation reports a linear relationship between “time per sampling/CPT operation” and “water depth” became evident, as demonstrated in Figure 1. The times presented on this Figure correspond to the time spent between the end of the drilling at a given depth until such time when drilling recommences at the same depth. These values are therefore considered to represent the amount of time that the equipment operator requires to prepare and connect the cone equipment/ sample tube, to lower the equipment to the bottom of the borehole, to perform the test/sampling operation, to recover the tool to deck and to disconnect. Since the drilling time is not included in this “operational” time, the values presented in Figure 1 would not be expected to vary significantly based on ground conditions. The information presented in Figure 1 can then be used to construct average progress rates for a 40 m deep borehole constructed by surface drilling techniques, as presented in Figure 2, Figure 3 and Figure 4 for continuous sampling, continuous CPT and composite boreholes respectively. It is then possible to compare these production rates directly with the production rates achieved by the seabed drill during the Luva site investigation campaign in 2009. The 40 m “composite” borehole progress rate presented in Figure 4 for the seabed drill is based on (1) time taken for 40 m continuous sampling borehole, © 2011 by Taylor & Francis Group, LLC

(2) time taken for 40 m CPT, (3) 20 min additional time allowed for “bumping over” between the two locations, and (4) 4h additional time allowed for offloading of the samples. Compared to a composite borehole performed by surface drilling the seabed drilling “composite” boreholes presented above hence provide twice as much information. The 40 m continuous sampling progress rates presented in Figure 2 also include 4h additional time, which is approximately the time that is required between two deployments to offload the samples and to reload the carrousels with new sample barrels. Where the site investigation programme is carefully planned this time can be utilised for steaming between sites or to perform other types of offshore operations (e.g. geophysical survey, boxcore testing). 3.3

Discussion on findings

As demonstrated in Figures 2–4 above the performance break-even point between surface drilling and seabed based drilling is expected to be at around 150 m water depth for both continuous sampling and composite boreholes to 40 m bml. Seabed based technology is expected to be more efficient for continuous CPT operations to 40 m bml in all water depths. It should however be noted that the production rate comparison presented in Figures 2–4 is only relevant for very soft to soft soil conditions, such as those typically, but not necessarily always, present in the deep

330

water environment. The efficiency of both the seabed drilling and the surface drilling technology is reduced in more complex ground conditions, where the former is possibly affected more than the latter. On the other hand, some of the currently available seabed drilling systems have the capability to switch to coring mode within the borehole, which presents a major advantage over surface drilling technology. Encountering a boulder within the borehole may cause the borehole to be prematurely terminated with conventional drilling techniques, whereas a suitably equipped seabed drill may be able to core past the obstruction and continue with testing and sampling operations below that depth. 3.4 Other considerations Even though the results of the performance comparison above appear to be very much in favour of seabed drilling technology, the advantages provided by surface drilling should on no account be discounted. One of the biggest advantages provided by surface drilling is the fact that it delivers real time information during the sampling process and hence the scope of work within the borehole can be altered as needed as the borehole progresses. Similarly, since the samples are recovered to deck as the borehole progresses it is possible to perform a significant amount of laboratory testing with a good offshore laboratory onboard (undrained unconsolidated triaxial tests, miniature lab vane tests and other classification tests) as further sampling takes place, which is not possible to achieve with seabed drilling technology and which would most likely need to be delayed until after the completion of the site investigation programme on seabed drilling campaigns. Hence on site investigation campaigns where complex ground conditions are involved and critical decisions may need to be made promptly offshore during the site investigation campaign (e.g. on site investigations conducted prior to jack-up emplacement) the surface drilling technology may present a more suitable solution. In comparison, on campaigns where the ground conditions and the local geology are predicted

© 2011 by Taylor & Francis Group, LLC

to be more uniform and the results of the lab testing are not as time-critical, the seabed drilling technology is likely to provide a more efficient alternative to surface drilling at almost any water depth.

4

CONCLUSIONS

The findings of the Luva site investigation campaign were analysed in an attempt to draw a comparison between conventional vessel-based drilling techniques, “surface drilling”, and remotely operated seabed-based drilling technology, “seabed drilling”. Both techniques have their advantages and disadvantages which have been briefly discussed in this paper. A comparison of the production rates from Luva with a nearby site where surface drilling techniques were utilised revealed that seabed drilling technology is potentially 3–5 times more efficient than surface drilling in 1,300 m water depth and soft soil conditions. Further analyses were carried out in order to establish the minimum water depth beyond which seabed drilling becomes more efficient than surface drilling and this water depth was found to be at approximately 150 m for sampling and composite boreholes to 40 m bml, and possibly even shallower considering the reservations made in the comparison.

ACKNOWLEDGEMENTS The work presented here was commissioned by Luva Project within Statoil. The authors would like to acknowledge the support and permission to publish this information by Statoil and the partners ExxonMobil and ConocoPhillips.

331

REFERENCE Det Norske Veritas (DNV). 2001. Qualification procedures for new technology. Recommended Practice DNV-RPA203.

4 Soil characterisation and modelling

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Rheological behaviour of soft clays P.E.L. de Santa Maria Acergy Brasil S.A.

I.S.M. Martins COPPE-UFRJ

F.C.M. de Santa Maria Eletrobras Termonuclear S.A.

ABSTRACT: The main objective of this paper is to draw attention to evidences of the viscous resistance in the behaviour of saturated clays. The Principle of the Effective Stress and its corollaries are initially discussed, followed by a revision of the main concepts involving viscosity of soils. Further on, some laboratory test results are presented, evidencing the effects of a viscous component in the soil resistance to deformation under a strain rate field.

1

INTRODUCTION

The Effective Stress Principle, enunciated by Terzaghi (1936), constitutes the theoretical basis of the Soil Mechanics framework. Atkinson & Bransby (1978) enunciated three corollaries: Corollary 1: “The engineering behaviour of two soils with the same structure and mineralogy will be the same if they have the same effective stress”. Corollary 2: “If a soil is loaded or unloaded without any change of volume and without any distortion there will be no change of effective stress”. Corollary 3: “Soil will expand in volume (and weaken) or compress (and strengthen) if the pore pressure alone is raised or lowered”. Martins (1992) presented examples showing that these corollaries are not always true. These evidences may be seen in undrained shear strength tests, where two samples of the same soil present different strains for the same effective stress but different strain rates; in relaxation tests, where variation in effective stress is observed with no strain and in a one-dimensional consolidation test where the measured pore pressure increased after closing the drainage (Lima 1993 and Martins et al. 1997). To extend the scope of the Principle, Martins (1992) enunciated the Expanded Principle of Effective Stress, valid for saturated soils without acceleration, hereafter presented in two parts: 1st Part: “In any plane of a saturated soil element subjected to a normal stress σ and a shearing stress τ, there will be internally resisting to σ the sum (σf
+ σv
+ u), where σf
is the frictional component and σv
the viscous component of the normal effective stress and u the pore © 2011 by Taylor & Francis Group, LLC

335

pressure; and resisting to τ the sum (τf + τv ), where τf is the frictional and τv the viscous components of the shear resistance”. The following equations may be written:

where φmob is the mobilised angle of internal friction and ηn and ηs the coefficients of viscosity of the soil for longitudinal strain rate (dε/dt) and shear strain rate (dγ/dt) respectively. 2nd Part: “Whenever a variation in the mobilized angle of internal friction occurs, there will always be shearing strains and reciprocally, whenever shearing strains occur, there will always be a variation in the mobilized angle of internal friction (undrained conditions).” Having in mind that laboratory and in situ tests imply strain rates in the soil, a viscous resistance are therefore always present in these tests and the strain rate associated to any particular test may have a significant influence on the result. In other words, if a test is carried out on two similar samples of the same soil using different strain rates, two different values of the concerned parameter will be obtained. Such apparent discrepancy results from the strain rate effect. A physical understanding and an adequate quantification of the viscous phenomenon allows a proper interpretation of test results, leading to basic parameters of behaviour, irrespective of the strain rate. The aim of this paper is to draw attention to evidences of the viscous resistance in the behaviour of saturated clays, presenting experimental results that may contribute to the knowledge of the viscous nature of

these soils. In particular, marine clay sediments interacting with most of the offshore structures are very soft and therefore displaying low frictional strength. In this case, the viscous resistance may play an important role in the behaviour of the system soil-structure. 2 VISCOSITY Many real bodies display viscous behaviour distinct from the Newton law (Suklje 1969), generalized in the following equations:

where τi , ηs and dγ/dt are the stress, viscosity coefficient and strain rate associated with shear strain and σi , ηn and dε/dt the same variables associated with longitudinal strain. This distinct behaviour, known as non-linear or non-Newtonian, manifests itself under the form of a variation in the viscosity coefficient. Undrained triaxial tests and oedometric and hydrostatic consolidation tests have shown evidences of non-Newtonian behaviour of soils, as further on presented in this paper.

Figure 1. Schematic undrained triaxial test results.

3 VISCOUS BEHAVIOUR OF SOILS A saturated soil is a two-phase system consisting of (1) solid particles enveloped by an adsorbed water film strongly adhered to them, called solid water, and (2) water filling the void volume of the soil structure (Terzaghi 1941). This water displays variable viscosity, from very high values in the near vicinity of the particles, decreasing with increasing distance from them. For constant loading rate, the resistance to deformation comes from both frictional and viscous phenomena (Taylor 1942). The first one arises from the solid water contacts and the second one arises from the viscous behaviour of the water. 3.1

Figure 2. Representation of the viscosity jump.

curves represent tests with ε = 0, which could not be practically feasible. However, it is not difficult to infer them from real test results once the viscous resistance is well known, as illustrated in Figures 1 and 2. The fundamental point that arises from the previous consideration is the importance of discounting the viscous component (η.˙εt ) from the total resistance when interpreting strength tests, leading to basic curves and basic angles of internal friction (αb ) as illustrated in Figure 3.

Soils under undrained shearing stresses

Lacerda (1976), Martins (1992), Guimarães (2000), showed that typical undrained triaxial tests featured the pattern of behaviour schematically represented in Figure 1, where the strain rate ε˙ t1 < ε˙ t2 . It has also been observed that the pore pressure u does not vary significantly as the strain rate increases. Another interesting feature apparent from the results is the “viscosity jump” at the very beginning of the test, illustrated in Figure 2. The stress path A-A
(viscosity jump) takes place theoretically with ε = 0 and u = 0, but with finite ε˙ . In fact, this is the viscous resistance corresponding to the strain rate associated to the test. This resistance is mobilised at the very instant the top-cap touches the sample. It is worth calling attention to the stressstrain curve and stress path ABC in Figure 1. These two © 2011 by Taylor & Francis Group, LLC

3.2

Soils under compression

Taylor & Merchant (1940) proposed a consolidation theory, called Theory A, which included the influence of the secondary compression. In this paper, the authors considered that primary consolidation was a hydrodynamic phenomenon governed by a strictly elastic mechanical behaviour. The secondary consolidation was a viscoelastic phenomenon governed by Kelvin’s rheological model (Suklje 1969, Findley et al. 1976). For a given normally consolidated soil, the frictional component of the effective stress is only a function of

336

Table 1. Artificial soil characteristics. wL (%)

wP (%)

IP (%)

Gs

92.5

21.8

70.7

2.62

Maria 2002). The initial mobilization of the viscous resistance in one-dimensional consolidation tests is accounted for by the initial strain rate field in the sample, which is also affected by the compressibility of the water (Tsytovich & Zaretsky 1969). 4

All the experimental evidences hereafter presented were inferred from tests carried out at the Clay Rheology Laboratory of COPPE-UFRJ. This laboratory operates in an insulated room provided with a temperature control system.

Figure 3. Basic stress path and angle of internal friction.

4.1

Figure 4. Schematic representation of the pore pressure variation with time for both measured and theoretical (without viscosity) values.

the loading and the void ratio. It is believed that the viscous component is, for the same soil, a function of the void ratio and the strain rate. Thus, for a fixed loading and void ratio, and admitting the validity of the Effective Stress Principle, any observed variation in the pore pressure may be accounted for by the existence of a viscous component σv
. Furthermore, in these conditions, as a consequence of equation (1), one may write:

Equation (5) would allow the evaluation of the intensity of σv
associated to a particular strain rate and void ratio. One could also estimate σv
from a onedimensional consolidation test calculating the pore pressure by means of an adequate Soil Mechanics mathematical model, where the viscosity is not considered, and then subtracting the measured pore pressure from this value, as schematically represented in Figure 4. The typical progress of pore pressure with time displayed in Figure 4 has been observed in laboratory tests (Tsytovich & Zaretsky 1969 and Santa © 2011 by Taylor & Francis Group, LLC

LABORATORY EVIDENCES

Undrained triaxial tests

Aiming at qualitative evaluation of the viscous resistance of a clayey soil, Guimarães (2000) carried out triaxial compression CIU tests. The samples were hydrostatically consolidated at three stresses, 95 kPa, 140 kPa and 190 kPa, and saturated under a back-pressure of 50 kPa. The consolidation phase ended at a volumetric strain rate of approximately 4(±1)×10−8 s−1 . The shearing phase was carried out at two strain rates, ε˙ = 0.1%/min and ε˙ = 0.01%/min, aiming at assessing the influence of this parameter in the results. The samples were 7 cm in diameter and 14 cm high and consisted of an artificial soil obtained from consolidating slurry prepared with 20% of bentonite and 80% of kaolin. The Atterberg limits and the average specific gravity of the solid particles are presented in Table 1. According to item 3.1, the key point to interpret an undrained triaxial test result is the evaluation of the viscosity jump, which is the viscous resistance of the soil sample. However the illustration of the viscosity jump in Figure 2 is not realistic because the contact soil × top-cap does not happen instantaneously, but gradually owing to the irregularities of the sample top face and fluctuations on the apparatus speed. Nevertheless, it is not difficult to obtain the value of the viscous resistance from a simple geometric construction. Figure 5 shows the stress path of three CIU tests and the corresponding basic path (in dashed lines) obtained from subtracting the viscous component from the overall resistance. Figure 6 displays how the viscosity resistance varies with the strain rate, suggesting a non-Newtonian behaviour. Table 2 presents the interpreted data from the tests. It is noticeable that in this case the basic or true angle of friction φb
is between 39% and 53% of the classical φ
, obtained from total resistance.

337

Figure 5. Stress paths of three CIU triaxial tests and corresponding basic paths in dashed lines (after Guimarães 2000).

Figure 7a. Decrease in the pore pressure immediately after he opening of the drain – hydrostatic consolidation tests (after Thomasi 2000).

Figure 7b. Increase in the pore pressure after closing the drain – hydrostatic consolidation test (after Thomasi 2000). Figure 6. Variation of the viscous component of the shear stress with strain rate – CIU triaxial tests (p
e = hydrostatic consolidation stress). Table 2. Values of the strength parameters of CIU tests.

Test

Cons. Stress (kPa)

φ


φb


p
f (kPa)

qf (kPa)

CP1 CP8 CP9

95 140 190

16.8◦ 17.0◦ 17.0◦

6.5◦ 8.4◦ 9.0◦

83 120 161

24 35 47

4.2

Consolidation tests

Both hydrostatic and one-dimensional consolidation tests were performed at the Clay Rheology Laboratory, aiming at improving the understanding of the influence of the viscous resistance in the behaviour of soils under volumetric strains. 4.2.1 Hydrostatic compression tests Thomasi (2000) performed hydrostatic consolidation tests using same soil and sample size as Guimarães (2000). The tests were performed with a back-pressure of 50 kPa and consolidation stresses of 90 kPa and 140 kPa. At the end of primary consolidation, the drain was closed and the pore pressure measured. The test results indicated the following features of behaviour, showed in Figures 7a and 7b. • At the beginning of the tests a decrease in pore pres-

sure was observed immediately after the opening of © 2011 by Taylor & Francis Group, LLC

338

the drain, rising again to a maximum value (Fig. 7a). After this, the pore pressure decreased steadily until almost zero (not represented in Figure 7a). • At the end of primary consolidation, when the drain was closed, the pore pressure increased gradually (Fig. 7b) until stabilization (not represented in Figure 7b). The first feature described above suggests that the high strain rate verified in the very beginning of the test resulted in significant values of the viscous component of the normal effective stress. Having in mind that the void ratio underwent almost no variation and admitting the validity of the Principle of Effective Stress (Terzaghi 1936), the only possible behaviour would be a transient reduction of the pore pressure. The second feature is a consequence of the same phenomenon manifesting itself in a reverse way. At the instant the drain was closed, the viscous resistance associated to ε˙ v became null, generating an increase in pore pressure. Figure 8 illustrates the σv
× ε˙ v behaviour for the consolidation stress of 90 kPa. 4.2.2 One-dimensional compression tests Santa Maria (2002) carried out oedometric consolidation tests in a K0 cell primarily intending to identify the variation of K0 during the secondary consolidation. The cell was provided with a top drainage and a pore pressure transducer at the bottom. Incremental and constant rate of displacement tests were performed on remoulded samples of Sarapuí clay, whose Atterberg limits and specific gravity of the solid particles are shown in Table 3.

Figure 8. Variation of the viscous component of the hydrostatic effective stress with volumetric strain rate – hydrostatic consolidation test (after Thomasi 2000). Table 3.

Figure 10. Curves of pore pressure versus time corresponding to 5 steps of loading, unloading and reloading – oedometric consolidation test (after Santa Maria 2002).

Sarapuí clay characteristics.

wL (%)

wP (%)

IP (%)

Gs

123

46

77

2.68

Figure 11. Viscous resistance versus strain rate corresponding to all steps of loading – oedometric consolidation test (after Santa Maria 2002).

Figure 9. Experimental data and Burgers model function fitted for relaxation of vertical effective stress - oedometric consolidation test (after Santa Maria 2002).

At the end of a constant rate of displacement test, a stage of stress relaxation was performed. The vertical stress at the top of the sample experienced a decrease of around 44% of the maximum stress (585 kPa to 327 kPa) in about 578 hours. It was observed that Burgers rheological model (Suklje 1969, Findley et al. 1976) fitted very well the measured variation of the vertical stress with time (relaxation), as presented in Figure 9. Figure 10 shows the results of the pore pressure readings at the base of the sample versus time for all steps of loading and unloading for an incremental loading test, following the pattern of behaviour discussed in item 3.2. Applying the methodology proposed in 3.2, represented in Figure 4 and by equation (5), the average value of the viscous resistance was obtained. Diagram of Figure 11 was produced using the viscous resistance normalized with respect to the increment of the total stress and the corresponding average strain rate. This diagram represents all steps of loading. It is noticeable that the shape of the curve suggests the existence of a limit value for σv
corresponding to a © 2011 by Taylor & Francis Group, LLC

rate of about 0.25%/min (valid for this particular soil). The non-Newtonian aspect of this diagram confirms the results obtained by Taylor (1942) for oedometric tests and Thomasi (2000) for hydrostatic consolidation tests (Fig. 8).

5

CONCLUSIONS

1. There are consistent evidences of a viscous component in the resistance of clayey soils, function of the strain rate and the void ratio. This component can be inferred from current laboratory tests. 2. The assessment of the viscous component of the soil resistance in laboratory and in situ tests allows the definition of only one set of “basic” curves and strength parameters, functions only of the frictional resistance and independent of the strain rate associated to the corresponding test. 3. The effective Stress Principle Corollaries (Atkinson & Bransby 1978) are not always true. The classical Effective Stress Principle was extended (Martins 1992) to take into account the viscous component of the soil resistance. 4. As far as viscosity is concerned, the soil behaves following a non-Newtonian pattern, with the viscosity coefficient η displaying a functional dependence on the strain rate. 5. Considering that Offshore Engineering frequently deals with soil-structure interaction, usually

339

involving soft soils, it is of major importance to improve the understanding of the viscous component of the overall resistance associated to this interaction.

ACKNOWLEDGEMENTS The authors would like to thank Acergy Brasil S.A. for the support and assistance with the preparation of this paper and also the Brazilian Research Council (CNPq) for the financial support to the experimental work.

REFERENCES Atkinson, J. H. & Bransby, P. L. 1978. The mechanics of soils: an introduction to critical state soil mechanics. McGrawHill Book Company (UK) Limited. London, 375p. Findley, W. N., Lai, J. S. & Onaran, K. 1976. Creep and relaxation of nonlinear viscoelastic materials. Dover Publications, Inc., New York. Guimarães, P.F. 2000. Study of the influence of a viscous component in the shear strength of saturated clays, in Portuguese, M.Sc. Dissertation, COPPE/UFRJ, Rio de Janeiro, RJ, Brasil. Lacerda, W. A. 1976. Stress relaxation and creep effects on soil deformation, Ph.D. Thesis, University of California, Berkeley. Lima, G.P. 1993. A study of non-linear one-dimensional consolidation, in Portuguese, M.Sc. Dissertation. COPPEUFRJ, Rio de Janeiro, RJ, Brasil.

© 2011 by Taylor & Francis Group, LLC

Martins, I. S. M. 1992. Fundaments of a model of behaviour of saturated clayey soil, in Portuguese, D.Sc. Thesis, COPPE/UFRJ, Rio de Janeiro, RJ, Brazil. Martins, I. S. M., Santa Maria P. E. L. & Lacerda, W. A 1997. A brief review about the most significant results of COPPE research on rheological behaviour of saturated clays subjected to one-dimensional strain, In: proceedings of the International Symposium on Recent Developments in Soil and Pavement Mechanics, Rio de Janeiro, Brazil. Santa Maria, F. C. M. 2002. Rheological experimental study of the coefficient of earth pressure at rest, K0 , in Portuguese, D.Sc. Thesis., COPPE-UFRJ, Rio de Janeiro, RJ, Brazil. Suklje, L. 1969. Rheological aspects of soil mechanics.WileyInterscience, division of John Wiley & Sons Ltd. Taylor, D.W. & Merchant, W. 1940. A theory of clay consolidation accounting for secondary compression. Journal of Mathematics and Physics, 19(3): pp.167–185. Taylor, D.W. 1942, Research on Consolidation of Clays, Department of Civil and Sanitary Engineering, Massachusetts Institute of Technology, Serial 82, August. Terzaghi, K. 1936. The shearing resistance of saturated soil and the angle between the planes of shear, In: Proceedings of the 1st ICSMFE, V. 1, pp. 54–56, Massachusetts, p.45–65. Terzaghi, K. 1941. Undisturbed clay samples and undisturbed clays, In: Contributions to soil Mechanics, 1941–1953, p.45–65. Thomasi, L. 2000. About the existence of a viscous component on the normal effective stress, in Portuguese, M.Sc. Dissertation, COPPE/UFRJ, Rio de Janeiro, RJ, Brazil. Tsytovich, N. A. & Zaretsky, Yu. K. 1969. The development of the theory of soil consolidation in the U.S.S.R., 1917–1967. Géotechnique, 19. No. 3, pp. 357–375.

340

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

A three-dimensional finite element study of the direct simple shear test J.P. Doherty & M. Fahey The University of Western Australia

ABSTRACT: A number of different versions of direct simple shear (DSS) apparatus have been devised over the past few decades. The DSS apparatus at the University of Western Australia (UWA) is similar to the Norwegian Geotechnical Institute (NGI) DSS, except that the specimen is contained in a pressurised cell (like a triaxial specimen). Specimens are consolidated by increasing the vertical load and horizontal stress independently, generally with a back pressure to ensure saturation. An undrained shear test is carried out by displacing the bottom drive horizontally, while maintaining constant height and volume. With no means of generating complementary shear stresses on the vertical faces of the specimen, the mode of shearing cannot be true “simple shear”. In fact, the stress state in the specimen is complicated, and this leads to some uncertainty in how to interpret the test results. The paper describes a finite element study of the UWA simple shear test, carried out using the ABAQUS finite element software. Soil models used include an elastic-plastic (Tresca) model, an elastic Mohr Coulomb model, and a soft-soil hardening model (a Cam Clay type model). The results show the strengths and limitations of the UWA simple shear test, and indicate how the test results should be interpreted.

1

INTRODUCTION

2 THE UWA DSS APPARATUS

A number of different direct simple shear (DSS) apparatus have been developed over the past few decades (e.g. by NGI, the University of Cambridge, University of California at Berkeley). In none of these devices can true simple shear conditions be imposed – i.e. a prismoidal specimen subjected to perfect plane strain conditions, with independent control of the vertical and horizontal normal stresses, and the ability to apply shear stress on the top (and bottom) surface and an equal complementary shear stress on the vertical (“front” and “back” surfaces). The final version of the Cambridge DSS apparatus (Budhu, 1979) comes closest to this ideal. None of the DSS apparatus using a cylindrical specimen (those similar to the NGI apparatus) can impose the complementary shear stress, and this, coupled with having a cylindrical specimen, means that the stress state in the apparatus when shear stress is applied is quite complex. Nevertheless, it is common to interpret the test using simplifying assumptions – particularly that the stress state can be determined from the boundary stresses or forces, and that simple shear conditions exist, even though there is no complementary boundary shear stresses. In this paper, a three-dimensional (3-D) finite element analysis of the UWA DSS is conducted using ABAQUS (HKS 2006) to determine if stresses measured on the boundary can be used to determine the strength parameters of a typical specimen. Modelling is carried out using an elastic-plastic (Tresca) model and a Modified Cam Clay (MCC) model. © 2011 by Taylor & Francis Group, LLC

The UWA DSS is based on the UC Berkeley DSS, in that it tests a cylindrical specimen enclosed in a latex membrane within a pressurised cell, just like a triaxial specimen. In contrast, the NGI-type DSS uses wire winding around the specimen to provide the lateral confinement. With independent control of the vertical and horizontal stresses, it is possible in theory to apply any desired initial consolidation stresses to the specimen in the UWA DSS, though there are limitations to this, as will be explained later. One of the major advantages of the UWA (and Berkeley) apparatus is the ability to apply an elevated back pressure (ensuring full saturation), and the consequent ability to measure the pore pressure changes accurately in an undrained test. Hence it is possible to determine the effective stresses as well as the total stresses in an undrained test. In a typical UWA test, consolidation stresses are applied in the desired ratio of horizontal to vertical effective stress. An undrained shearing test consists of moving the base carriage horizontally, applying shear to the specimen, while maintaining a constant specimen height. It is also common practice to maintain the total vertical stress constant by varying the horizontal stress (the cell pressure). The test is interpreted assuming that the stress state in the specimen is true simple shear, such that the horizontal shear force divided by the specimen crosssectional area gives the shear stress τzx , where x is the direction of horizontal movement, and z is the vertical direction. On the assumption of perfect simple shear,

341

Figure 2. Stress path for horizontal plane (z-plane) from DSS test on normally consolidated Boston blue clay (Wroth, 1987).

Figure 1. Mohr circles for ideal drained DSS.

the “plane strain” stress invariants t and s (or s
) can be determined:

s
and t represent the centre and radius of a Mohr circle, respectively. Hence the maximum shear stress reached (τmax ) is the peak value of t, which therefore represents the undrained shear strength su . A t − s
stress path represents the locus of the “top” point of the Mohr circle as the stress state evolves during shearing (this corresponds to the path To –T in Figure 1). For a frictional soil, at critical state the t − s
stress path intersects a line through the origin inclined at α to the horizon

), where φcv is the critical tal, where α = tan −1 ( sin φcv state friction angle.

3

PREVIOUS STUDIES

3.1 Theoretical and experimental studies There has been much discussion on the stress states within a DSS specimen, and what the results of the test mean in terms of undrained shear strength su and effective stress parameters. In particular, Wroth (1984, 1987) attempted to trace out the stress paths followed in DSS (and direct shear) tests, referring to the work of de Josselin de Jong (1972). Due to the means of applying the consolidation stresses in the NGI device, and by choice in the UWA device, most DSS tests start from an anisotropic stress state (with σh
= Ko σv
, with Ko < 1). In an ideal drained DSS test, where σv
and σh
are kept constant, the stress paths can be represented on a Mohr circle diagram, as
). This shown in Figure 1 (where σv
= σzz
and σh
= σxx diagram represents (to scale) a DSS test in which the initial stresses (on horizontal and vertical planes, represented by points A and B) are 100 kPa and 40 kPa (i.e. Ko = 0.4), and where φ
= 35◦ . Applying shear stress τzx to plane A (and complementary τxz to plane B) changes the stress state until points A∗ and B∗ are © 2011 by Taylor & Francis Group, LLC

Figure 3. Stress path for horizontal plane (z-plane) from DSS test on normally kaolin (Wroth, 1987).

reached, where the Mohr circle touches the failure surface (OF–OF∗ ). The “pole of planes” is point P, such that the orientations of the planes of maximum stress obliquity are indicated by lines PF and PF∗ , whereas the orientation of the planes of maximum shear stress τmax are indicated by lines PT and PT∗ . Thus, failure occurs on almost vertical planes (PF∗ ) and planes at about 42◦ to the horizontal (PF). In particular, the friction angle is defined by the stresses on plane F (F∗ ), and it is clear that horizontal planes (A∗ ) are not involved in the failure at all – i.e. tan−1 (τzx /σzz
) is not equal to φ
. Coincidentally, plane B* is almost coincident with the plane of failure (F∗ ), so that a simple and almost correct interpretation of the friction angle would in this case be obtained as
tan−1 (τxz /σxx ). Wroth (1987) presented data from an undrained DSS test on normally consolidated Boston Blue Clay by Ladd and Edgars (1972) in an NGI-type apparatus. Figure 2 shows the stress path for the horizontal plane (z-plane) for this test. The stress on vertical planes (xplanes) is not measured, so the corresponding x-plane stress path cannot be plotted. Large positive pore pressures are generated, which allows the effective stress path to migrate to the left (ABC), and the ultimate stress state on the horizontal plane (point C) appears to define φ
. Wroth (1987) also showed results from the Cambridge DSS by Borin (1973) reproduced as Figure 3. One of the strengths of the Cambridge DSS is clear from this figure – the ability to measure the stresses on both faces, using contact stress transducers. This allows the full stress path to be determined, allowing the Mohr circle at failure to be plotted, as shown. In

342

Figure 4. Extended modified Cam-clay yield surface in deviatoric plane (after HKS 2006).

Figure 5. The finite element mesh.

this case, the failure condition is reached on vertical planes (T), and not on the horizontal planes (B).

For triaxial conditions, q = σ1 − σ3 . The parameters λ and e1 control the slope and location of the isotropic compression line (as in conventional MCC), respectively, and the slope of the unload-reload line is controlled by κ.

3.2 Numerical (FE) modelling FE modelling of the DSS has been carried out previously (e.g. Budhu and Britto 1987; Potts et al. 1987), and these studies have provided interesting insights into various forms of the DSS. However, in none of these cases has the modelling being 3-D, and none of them have involved a cylindrical specimen. Thus, we believe the results presented in this paper represent the first attempt to model a cylindrical DSS test using 3-D FE with the MCC model. 4 THE SOIL MODELS In this study, two different soil models have been used: an elastic perfectly plastic (Tresca) model, and a Modified Cam Clay (MCC) model. The Tresca model is well known, and will not be described further. The MCC model used is the “extended modified Camclay” soil model (EMCC) available in ABAQUS. This model provides two extensions on the classical MCC model developed by Roscoe and Burland (1968). The parameter β provides the first extension by controlling the shape of the yield surface on the “wet side” of critical state. However, in the modelling that follows, β = 1, which results in the standard modified Cam-clay ellipse. The second extension is provided by the parameter K, which is the ratio the critical state shear stress in triaxial extension and triaxial compression. The parameter must be in the range 0.778 ≤ K ≤ 1.0, and changes the shape of the yield surface in the deviatoric plane (π-plane) as shown in Figure 4. Here, K = 1, giving the same strength in all loading paths (DSS and triaxial compression and extension). Stress states are expressed in terms of the invariants q and p
, where:

© 2011 by Taylor & Francis Group, LLC

5 THE FINITE ELEMENT MODEL The finite element mesh (and the coordinate system) used to model the DSS is shown in Figure 5. Due to symmetry, half the specimen was modelled. The mesh comprised 5,420 20-noded brick elements with 27point Gauss integration. With respect to the global x, y, z coordinate system, the plane of symmetry has a constant value of y. The top and bottom surfaces have constant values of z. The diameter of the specimen was 72 mm and the height 28 mm – typical of the values used in the UWA DSS. For the Tresca model, the parameters used were a Tresca “su ” = 30 kPa, Young’s modulus E = 10 MPa (i.e. E/su = 333), and Poisson’s ratio ν = 0.49. For the EMCC model, the soil parameters used are listed in Table 1. The values correspond to those of reconstituted kaolin clay (Stewart 1992). The DSS was simulated in the following 3 stages: 1. “Wish” in place a normally-consolidated specimen, with σv
= σh
= 1 kPa. 2. Apply σv
= 100 kPa to the top boundary and σh
= 75.4 kPa to the radial boundary to simulate Ko loading. 3. Prescribe horizontal displacement in the x–direction while fixing displacement in the y– and z–directions, with drainage prevented. For each stage, zero x, y and z displacements were prescribed at all nodes on the base, with zero ydisplacement on the plane of symmetry. In stage 2, the ratio σh
/σv
was based on the theoretical Ko value for normally consolidated MCC (Wood 1990):

343

Table 1.

Parameters

Parameter

Value

λ κ M ν β κ e1

0.205 0.044 0.92 0.3 1.0 1.0 2.140

where  = (λ − κ)/λ and ηKnc is the q/p
ratio for one-dimensional compression (i.e for Ko loading conditions) and is related to Ko nc :

Substituting  = 0.785, M = 0.92 and ν = 0.3 into Equation 3 and numerically solving gives ηKnc = 0.3. Substituting this into Equation 4 gives Ko nc = 0.754. This value was confirmed by conducting a “true” Ko loading (consolidation) step in which the vertical pressure was increased while horizontal displacements on the radial boundary were fixed. The shearing stage (stage 3) was conducted assuming undrained conditions. However, a high permeability k was used to ensure that internal pore pressures were always uniform. During the shearing stage, displacements in the x and z-directions on the top cap were fixed (i.e. zero), while a uniform non-zero displacement was prescribed in the y-direction. 6

RESULTS

6.1 Consolidation phase In theory, a strength of the UWA-type DSS is its ability to apply vertical and horizontal stress that are independent of each other, allowing specimens to be consolidated to stress states other than “true Ko ”, but this is not the case in practice. The DSS top and bottom caps are deliberately made perfectly rough, to ensure that no slip occurs between the specimen and these caps when the shear displacement is applied, and so none can occur in the consolidation phase either. Thus, no horizontal (radial) strain can occur in the soil immediately adjacent to the ends in consolidation, so in this region, “true Ko ” conditions apply, irrespective of the actual ratio between the applied vertical stress and cell pressure. This was explored by carrying out the consolidation loading stage using different ratios of horizontal to vertical stress, using the EMCC model and model parameters described above. This showed that when the horizontal consolidation stress was chosen to be Ko times the vertical consolidation stress, the resulting stress state was almost perfectly homogeneous, with the correct Ko being achieved at all points within the © 2011 by Taylor & Francis Group, LLC


Figure 6. Contours of σxx for isotropic consolidation to 100 kPa.

specimen. (For the chosen EMCC parameters used, the Ko value is about 0.75). However, when values other than the true Ko are used, the resulting internal stress state is highly inhomogeneous. This is well illustrated in the stress contour plot in Figure 6. This shows one quarter of the specimen, following consolidation to σv
= σh
= 100 kPa, where σv
is the total applied vertical force divided by the top surface area, and σh
is the uniform radial stress applied to the cylindrical boundary. The stress contours shown
(these are not radial effective stresses, so are for σxx axial symmetry is not to be expected, as is clearly the case). The contours on the left-hand vertical face (the
y–plane) show that there is a region where σxx = σh
right at the external boundary, extending in to less than ¼ of the radius at the mid-height, but with this zone reducing in penetration as the end caps are approached.
Within the main body of the specimen, σxx ≈ 75 kPa, which represents “true Ko ” conditions. Figure 6 shows that even though isotropic boundary stresses have been applied, the interior of the specimen is under approximately Ko stresses. Therefore, applying consolidation stresses other than at a “true Ko ” ratio is futile, and could produce misleading results in the shearing phase. In the Tresca model with Poisson’s ratio = 0.49, the “elastic” Ko value is 0.96, so that applying isotropic “consolidation” stresses in this case produces almost a homogeneous (isotropic) stress state. (The result presented later will be for Ko = 0.96). 6.2

Shearing phase: Tresca

Following initial compression with σv = 100 kPa and Ko = 0.96, the Tresca specimen (with su = 30 kPa) was sheared, to produce the τ − σ
stress paths shown in Figure 7(a). The theoretical Mohr circle at failure is also shown, as is the theoretical t − s
stress path. Two (practically coincident) stress-strain curves are shown in Figure 7(b), one derived from the horizontal shear force divided by the specimen area (in the same manner as an actual test would be interpreted), and the other represents the maximum shear stress (t) calculated using Equation 1, with the usual assumption about complementary shear stress. As expected, τzx

344

Figure 7. Results of modelling for the Tresca case: (a) stress paths and (b) stress-strain curves.

and t are practically identical, since Ko is very close to 1.0. Thus, the “test” reproduces the Tresca su , with a small degree of underprediction. 6.3

Figure 8. Model output for EMCC: (a) stress-strain (τzx /suTC ) curves; (b) corresponding stress paths; and (c) orientation of Plane T in the specimen.

Shearing phase: EMCC

The undrained shear strength in triaxial compression (suTC ) for the EMMC is obtained as follows:

where p
= σv
(1 + 2Ko )/3 = 83.6 kPa (for σv
= 100 kPa and Ko = 0.754) and p
c is:

where η = q/p
where q = 100–75.4 = 24.6 kPa. Substituting these values into Equation 5 give suTC = © 2011 by Taylor & Francis Group, LLC

24.09 kPa. This value was confirmed for the EMCC model in ABAQUS by simulating a triaxial compression test using a single 8-noded axisymmetric continuum element. This gave suTC = 23.46 kPa, which is within 3% of the analytical solution. The results of the shearing phase for the EMCC model are shown in Figure 8. Figure 8(a) shows shear stress versus shear strain curves, where the shear stress has been normalised by the theoretical shear strength in a triaxial test (suTC ). The “single Gauss point” curve is for a point near the centre of the specimen, showing that the model gives about the correct local strength. (Though the specimen is undrained in an overall sense, internal water flows can occur – k being chosen deliberately to be high to equalise internal pore

345

pressures – and hence minor local non-zero volumetric strains can occur). The “boundary” τzx /suTC plot shown is obtained by dividing the total horizontal force applied to the top by the specimen cross-sectional area (as is done in interpreting an actual DSS test in the UWA apparatus). Similarly, the “boundary t” plot is the value of t obtained from the top-cap force and the cell pressure, in accordance with Equation 1. The other curves in Figure 8(a) are obtained from the average values of τzx on planes within 3 mm of the “top” and “bottom”, and a plane near the “middle” of the specimen, as indicated in the legend. Figure 8(b) shows corresponding normalised stress paths for the “top”, “bottom” and “middle” planes, and the stress paths obtained from the top-cap force and applied boundary horizontal stress is also shown. Also shown are a t − s
plot (from the boundary stresses), and the Mohr circles (from the boundary stresses) for the start and end of the test. This plot shows that that the correct value of φ
(or
φcv ) is obtained from the top-cap τzx , and the maximum shear stress (i.e. su ) is given by the maximum value of t. As expected, τzx max = tmax . Maximum stress obliquity occurs on a horizontal plane (z-plane) for this case, as in the DSS test in Figure 2, and not on a vertical plane (x-plane), as in Figure 3. The maximum shear stress occurs on Plane T (Figure 8(b)), with the orientation of this plane shown on a cross-section of the specimen in Figure 8(c). This orientation is found from a “pole of planes” construction as in Figure 1, with the pole being found from a horizontal line through point F. An inclined failure plane such as Plane T is generally observed in undrained shear tests in the UWA DSS, suggesting that failure occurs on the plane of τmax rather than on the plane of maximum stress obliquity (F in Figure 8(b)). This is analogous to suggesting that failure in an undrained triaxial test always occurs on planes at ±45◦ to the horizontal, rather on planes at ±(45◦ + φ
/2), and this raises interesting questions is meant by “failure” in such tests (triaxial and DSS). Wroth (1987) discussed this issue, pointing out that even when “failure” in a typical undrained triaxial test occurs on the plane of maximum stress obliquity rather than on the plane of τmax , it is still conventional to define the undrained shear strength suTC as being equal to the τmax = q/2 from such tests, which is actually greater than the strength mobilised on the failure plane. 7

CONCLUSION

The modelling presented in this paper suggests that the results from the UWA DSS apparatus are rather more

© 2011 by Taylor & Francis Group, LLC

representative of ideal DSS than might be expected, given the complications of a cylindrical specimen, and the lack of complementary shear stresses on the “front” and “back” faces of the specimen.The results show that specimens consolidated with Ko consolidation stresses are likely to be more homogeneous at the end of consolidation that those consolidated to any other stress ratio, and hence it seems that only “true Ko ” consolidation should be used. The results also indicate that for the limited study performed (normally consolidated specimens with full internal drainage), the “standard” interpretation of the test (using Equation 1) gives cor
, and that a correct value of su can be rect values of φcv obtained from a t − γ plot. This work is continuing, as there are still many issues that need to be explored. The next stage will involve overconsolidated specimens, where overconsolidation is achieved under true Ko conditions in the consolidation phase. The effect of the initial anisotropic stress state resulting from applied stresses that are not “true Ko ” will also be investigated, as will the UWA standard procedure of increasing the cell pressure to achieve constant total vertical stress. This plot shows that τzx from the top-cap force gives a measure of su within about 10% of the theoretical value of suTC , but a better approximation of su is provided by the “boundary t” plot.

REFERENCES Budhu, M. 1979. Simple shear deformation of sands. PhD Thesis, University of Cambridge, UK. Budhu, M. & Britto,A.M. 1987. Numerical analysis of soils in simple shear devices. Soils & Foundations, 27(2) 31–41. de Josselin de Jong, G. 1972. Proc. Roscoe Memorial Symp. Stress-Strain Behaviour of Soils (ed. R. H. G. Parry), discussion on session II, 258–261. Cambridge: Foulis. HKS (2006). ABAQUS Users’ Manual, Version 6.5, Hibbit, Karlsson and Sorensen, Inc. Potts, D.M., Dounias, G.T. & Vaughan, P.R. 1987. Finite element analysis of the direct shear test. Géotechnique 37(1), 11–23. Roscoe, K. H. & Burland, J.B. 1968. On the generalised stress–strain behaviour of ‘wet’ clay. In: Heyman, J., Leckie, F.A. (Eds.), Engineering Plasticity. Cambridge University Press, pp. 535–609. Stewart, D.P. (1992). Lateral loading of piled bridge abutments due to embankment construction. PhD Thesis, The University of Western Australia. Wood D.M. 1990. Soil behaviour and critical state soil mechanics. Cambridge University. Press. Wroth, C.P. 1984. The interpretation of in situ soil tests. 24th Rankine Lecture. Géotechnique 34(4), 449–489. Wroth, C.P. 1987. The behaviour of normally consolidated clay as observed in undrained direct shear tests. Géotechnique 37(1), 37–43

346

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Repeated loading and unloading of the seabed H.J.E. Hu, K.K. Tho, C.T. Gan, A.C. Palmer & C.F. Leung Department of Civil Engineering, National University of Singapore, Singapore

ABSTRACT: In many contexts, offshore systems repeatedly load and unload the seabed in the same place. A catenary riser, for example, is repeatedly lifted from the seabed and lowered back, and progressively indents the bed, to the extent that it may cut a trench 2 m deep. This is vitally important to the mechanical behaviour of the riser, because the region near the touchdown point is a fatigue hotspot where the stress range is particularly high. Moreover, a deep trench constrains the riser from moving sideways, which can lead to large lateral curvatures if the floater moves off station. Similar problems occur during laybarge and reelship pipelaying, and when jackup spudcans are lifted and lowered. Our research applies a combination of centrifuge modelling, numerical modelling by ABAQUS and plasticity theory. It shows that this kind of interaction with the seabed cannot usefully be modelled by soil springs and that the interaction is much more complex. Repeated loading and unloading do not lead to shakedown to a condition at which the seabed response is elastic, and hysteresis continues to absorb energy from the system, though that may have a usefully favourable effect on the system dynamics. Another option is to pave the seabed to eliminate the formation of deep trenches. 1

INTRODUCTION

1.1 Background Traditional geotechnics is mostly concerned with loads that are applied only once, though there are instances of repeated loading, such as wheel/soil interaction and of course earthquakes. In contrast, offshore geotechnics almost always has to deal with cyclic loading, generally generated by waves, which act directly on the seabed itself (Damgaard and Palmer, 2001) and indirectly through piles, seabed-founded structures such as jackups (Osborne, 2005), ice (Palmer and Niedoroda, 2005), risers and pipelines (Palmer and King, 2008). This paper examines a particularly severe case, where a pipeline or a riser touches down on the seabed and is repeatedly lifted and lowered. 1.2 Context and motivation Figure 1(a) is a sketch of a floating production system, in this instance a SPAR, connected to the seabed by a mooring system and a steel catenary riser (SCR). The mooring system is inevitably fIexible, and allows some movement of the SPAR. If the SPAR moves to the right, the tension in the SCR increases, the riser is picked up, and the touchdown point (TDP) moves to the left. If the SPAR moves to the left, the tension drops, and the TDP moves to the right. Close to the TDP, the riser is alternately picked up and set down. A similar situation arises during laybarge pipelaying, sketched in Figure 1(b). Deepwater laybarges are dynamically positioned, move in response to waves, and sometimes wander away from their intended position. The TDP moves along as the barge progresses. A

© 2011 by Taylor & Francis Group, LLC

Figure 1. Schematic representation of (a) SCR connecting a pipeline on the seabed to a floating platform; (b) Pipeline during pipelaying.

typical lay rate is 4 km/day, and so a particular length of pipe does not usually spend much time close to the TDP, but sometimes laying is interrupted by mechanical breakdown on the barge, by a severe sea state, or by lack of pipe, and then the same length of pipe may be close to the TDP for hours or days. The practical significance of seabed/pipe interaction close to the TDP is related to the mechanics of a suspended length of pipe (Palmer et al., 1974). Most of the suspended span is essentially a catenary, in which the curvature is determined by the interaction between the submerged weight and the tension, and is

347

where ψ is the angle between the pipe axis and the horizontal, S is distance measured along the pipe,

for this reason the region close to the TDP is a fatigue hotspot. The actual response of the seabed to repeated picking up and setting down has a major influence on bending moments in the fatigue hotspot, and therefore on the fatigue life. That is sometimes taken into account by a ‘touchdown factor’ that multiplies w by 2 or 3, but that is a crudely oversimplified approach, and there is little real justification for the factor. Neither the rigid nor the rigid-plastic idealisation is a good one. Our modelling set out to investigate the true behaviour, particularly if a number of loading cycles degrades the indentation and uplift resistances and leads to a reduction of the forces between the pipeline and the soil.

Figure 2. Distributions of curvature in suspended pipeline induced by movement of touchdown point, to the right and left in Figure 1. (a) on rigid seabed; (b) deformable seabed.

2 w submerged weight per unit length, and U is the horizontal tension at the surface. Close to the TDP, however, there is a structural boundary layer in which the flexural rigidity of the pipe is important, because the bending moment has to be continuous and the curvature in the suspended span must match the curvature on the seabed. On a rigid seabed, the curvature within the boundary layer drops exponentially from w/U to√zero, there is a concentrated force of magnitude w (F/U) between the seabed and the pipe at the touchdown point, where in addition F is the flexural rigidity of the pipe. The √ characteristic length is (F/U ). For a large-diameter 762 mm (30 inch) pipeline F is 6.5 × 105 kN m2 , w is 1.1 kN/m, and so if U is 350 kN the concentrated force is 47.4 kN (4.7 tonnes) and the characteristic length is 43 m. A real seabed cannot withstand a concentrated force, and so the seabed indents and spreads out the load. The case where the seabed is linear-elastic is of little or no practical interest, because elastic deformations are so small. The case where the seabed can be idealised as rigid–perfectly plastic can be treated analytically (Palmer, 2008). Again there is a boundary layer, both above and below the mudline. The horizontal distance Z from the point where the pipe reaches the mudline to the point of maximum indentation is

where r is the indentation resistance per unit length; w/r must be smaller than 1, because otherwise the pipe would sink indefinitely. Figure 2 plots schematically the distribution of bending moment in the boundary layer close to the TDP for a rigid seabed (a) and a rigid-plastic seabed (b). The bending moments are smaller than they are in the suspended span, but as the TDP moves back and forth the range over which the bending moment changes is much larger than it is in the suspended span. Fatigue damage primarily depends on stress range, and © 2011 by Taylor & Francis Group, LLC

EXPERIMENTAL SETUP, SAMPLE PREPARATION AND PROCEDURES

The centrifuge model tests were conducted on the beam centrifuge at the National University of Singapore (NUS), and simulated repeated vertical movement of a length of riser. The model pipe was made from a hollow stainless steel tube with 3 mm wall thickness, with polytetrafluoroethylene (PTFE) end caps fitted at both ends to reduce the side friction between the pipe and the perspex plates on the inner walls of the chamber. The outer diameter of the model pipe was 50 mm, and its overall length was 300 mm, so that it fitted perfectly into the model container. Two 7 mm aluminium rods were attached to the studs welded on top of the model pipe. The other end of each rod was attached to a load cell. The load cell readings were checked to confirm that they record equal forces. All the centrifuge tests were carried out under a 20-g condition. In this paper, all results are presented at prototype scale in accordance with Taylor (1994), unless otherwise stated. Malaysian kaolin clay was selected for this study. Its properties are reported by Goh (2003). The dry kaolin powder was mixed with water at a water content of 120%. Before the slurry was poured, 25 mm of saturated sand was placed on the base of the container as a drainage layer and silicon grease was applied on the entire vertical internal wall to reduce the friction between the wall and the soil. The clay slurry was carefully poured into the model container half filled with water and then subjected to self-weight consolidation in the centrifuge up to 95% consolidation. A T-bar penetrometer with a bar diameter of 5 mm and length of 25 mm was used to determine the profiles of undrained shear strength of the model soil sample. The undrained shear strength, su, was calculated from the net penetration resistance, q, as su = q/Nt, where Nt is typically taken as 10.5, which is within the range derived from plasticity theory for the flow of soil around a deeply embedded cylinder (Randolph and Houlsby, 1984). The pipe penetrated from 0D invert embedment up to the maximum embedment of 3.5D. Upon reaching

348

Figure 4. Force–displacement relationship.

Figure 3. Undrained shear strength profiles.

the target embedment, it was extracted immediately until negligible resistance was recorded before being repenetrated for further load cycles. The pipe was driven into the model soil at 1 mm/s, and extracted at the same rate as soon as it reached the predetermined depth. This displacement rate makes vD/cv equal to 80 which is to ensure that the strain rate is carried out in an undrained condition. (Palmer, 1999, and Randolph and House, 2001) After the model pipe was completely lifted out of the model ground, it was penetrated to the same final depth again. This repetitive penetration and extraction continued until the force-displacement response stabilised. Figure 3 shows the undrained shear strength profiles of the model ground. The measured undisturbed shear strength increases almost linearly with depth and can be approximated as su = 1.55z.

3

CYCLIC TEST RESULTS

Figure 4 shows that the force–displacement relationship for the repetitive process of penetration and extraction. The force presented is the net vertical load per unit length, exclusive of the submerged weight of the model pipe. In what follows the net force between the pipe and the soil is denoted as negative when it is compressive (as when the pipe is first pushed in the soil) and positive when it is tensile (as when the pipe is pulled out of the soil). A compressive load of 46 kN/m was applied to drive the pipeline down to the target embedment depth: this is the maximum penetration load. After reaching the target embedment depth, the pipe is extracted immediately at the same rate. It is evident in Figure 4 that the uplift resistance of the pipe was mobilised instantaneously and reached the maximum of 14 kN/m at breakout. Upon further extraction, the uplift resistance decreased gradually to zero. The load reaches zero at the mudline, which implies that a trench of 0.5D depth was formed underneath the pipe after the initial penetration. © 2011 by Taylor & Francis Group, LLC

Figure 5. Normalised maximum penetration and uplift resistance degradation curve for the first 20 cycles.

Subsequently, the cyclic process of penetration and extraction continues with constant target embedment depth until it completes 50 cycles. From the second cycle onward, there is no penetration resistance until the pipe centre was embedded at the mudline level which confirms the presence of a trench beneath the pipe. As shown in Figure 4, while the number of cycle increases, it is clear that the maximum penetration resistance and maximum extraction resistance decreases, probably because of severe remoulding of the soil due to the large amplitude cyclic displacement and cavity collapse. Figure 5 is a plot of normalised maximum penetration and uplift resistance over the first 20 cycles. The maximum resistance in each cycle is normalised with the respective maximum penetration resistance of the first cycle. The resistance degrades sharply within the first few cycles, but stabilises after about 12 load cycles. It is also evident that there is a 27% difference between the penetration and uplift resistance during the steady state. Hodder et al. (2008) stated that the asymmetry of the cyclic load-displacement response could be accounted for by the soil buoyancy component of resistance. For clay with effective unit weight of

349

As the experimental result shows, soil at the SCR touchdown zone experiences degradation as the number of load cycles increases. The degrading component in the pipe-soil tensile force appears to be due to the diminishing suction. Soil degradation would reduce the SCR fatigue damage. As shown in the test result, the load penetration response approaches steady state after 12 load cycles. SCR touchdown is a high-cycle fatigue problem, and so it is not important to model the soil strength degradation in the first 12 cycles, particularly before the first severe storm occurs a real SCR touchdown is loaded by many small waves that induce small deformation cycles. Applying the loadpenetration profile at the steady state would further reduce the conservatism of an SCR design.

Figure 6. Normalised penetration and extraction resistance degradation curve for spudcan and T-bar.

4 6 kN/m3, the soil buoyancy would be about 4.7 kN/m (∼10% of the penetration resistance) when the pipe is fully submerged into the ground. The asymmetry of the load-displacement profile is still not fully accounted for. It might also due to the difference between the soil failure mechanisms during the penetration and uplift processes. Additional cyclic T-bar tests and a repeated loading of a spudcan tests were carried out in normallyconsolidated clay. The T-bar test was conducted in the drum centrifuge of The University of Western Australia (UWA), whereas the spudcan test was performed in NUS using Malaysian kaolin clay. The characteristic properties of the clay sample used in UWA can be found in Bienen et al. (2009). In the cyclicT-bar test, theT-bar was penetrated to 40 DT-bar and subsequently extracted to 30 DT-bar, where DT-bar is the diameter of T-bar. The cyclic test continued until the load-displacement response reached steady-state. Figure 6 plots the normalised penetration and extraction resistance degradation curve for spudcan and T-bar. The normalised penetration load degradation curve of T-bar is comparable with the pipe test. Chung et al. (2006) found T-bars with aspect ratios (length/diameter) from 4 to 10 showed very similar penetration resistances. The aspect ratio of the T-bar and the model pipe are 4 and 6 respectively, which is sufficient to simulate a plane strain condition and thus, give a comparable penetration response. However, the degradation of uplift resistance in the T-bar cyclic tests does not differ much from the degradation of penetration resistance as observed in the pipe test. It indicates that the pipe undergoes cycling in shallow depth or deeply embedded manner exhibit different soil failure mechanism. On the other hand, the spudcan test shows similar load degradation behaviour for cycle of up to 2 with slightly lower degradation ratio compared to the model pipe. The circular spudcan is axisymmetric rather than plane strain, and has a relatively shallow penetration depth of about 0.7 spudcan diameter. This may result in lower extraction resistance as less volume of soil above the spudcan. © 2011 by Taylor & Francis Group, LLC

NUMERICAL SIMULATION OF CYCLIC RESPONSE

The numerical analysis of an object penetrating deep into the seabed is a fundamentally challenging problem due to severe mesh distortion resulting from very large soil deformation, complex contact condition and nonlinear soil behaviour. In a conventional Lagrangian based large strain large deformation finite element analysis, severe mesh distortion occurs during the analysis leading to progressive deterioration in solution accuracy and the subsequent imminent termination due to negative volume phenomenon. In this study, an Eulerian based finite element technique in which the finite element mesh is kept stationary throughout the analysis and the material is allowed to move independently of the element nodal points is employed. By taking advantage of symmetry, only half of the domain is modelled. In order to minimize the boundary effect, a 6.0 m width by 7.0 m height of soil domain is included in the model. A strength gradient of 3.48 kPa per m depth is considered. Rough contact condition is adopted. In the first penetration, the pipe is advanced to a depth of 5D, measured from the pipe invert. Subsequent to that, 10 cycles of uplift and re-penetration are performed at constant amplitude of 0.5D. The Tresca constitutive model with an empirical strain softening model (Wang et al., 2009) as defined by Equation 3 is adopted.

350

where St is the true sensitivity and taken to be 4.2; ξ denotes the accumulated plastic strain and ξ95 represents the value of ξ at which soil has undergone 95% of total strength reduction and taken to be 30. suo is the initial undisturbed undrained shear strength. The numerically generated normalized loaddisplacement curve is presented in Figure 7. The experimental data corresponding to a centrifuge test performed with clay of OCR = 5 is plotted on the same

not represent the strength degradation characteristic of this particular clay used in the centrifuge experiments. In Figure 8, the equivalent plastic strain contour corresponding to a numerical model without considering strain softening is plotted. The area shown in grey denotes region of soil which has accumulated 200% equivalent plastic shear strain. It can be observed that a sizeable zone beneath the pipe has accumulated substantial equivalent plastic shear strain and this will significantly reduces the bearing capacity in clay with high sensitivity.

5

Driving a truck across the natural land surface almost always creates ruts, and repeated driving across the same surface deepens the ruts and makes the truck difficult to steer and to drive. In the same way, repeated loading around the TDP indents the seabed and creates a deep rut, which has a significant influence on the fatigue response. An unfortunate secondary consequence is that if the floater or the laybarge moves sideways (perpendicular to the plane of Figure 1), the pipe may be trapped in the deep rut, and that creates large bending moments at the point where the pipe comes out of the rut. In the case of roads on land, the obvious solution is to pave the surface with a stronger and harder material that will not be indented. There seems no reason why this solution should not be applied to the seabed. Similarly, it at first seemed strange to use wheels for equipment that has to move across the seabed, but seabed ploughs often have wheels and work well. One design solution is to pave the seabed with flexible mattresses. A typical mattress consists of concrete units, sometimes hexagonal and sometimes rectangular, typically 0.75 m across, tied together with polymer rope. Mattresses are routinely applied to span correction (Palmer and King, 2008), prevention of upheaval buckling, dropped object protection, and scour protection. Most of them are quite small, typically 4 m square, but in one instance a flexible mattress 60 m long and 5 m broad was successfully deployed as part of a defence project. A long mattress can be rolled up, lowered to the seabed at the TDP, and unrolled before the SCR was installed. It would eliminate uncertainty about cyclic loading close to the TDP.

Figure 7. Experimental and numerical load-penetration curves.

Figure 8. Equivalent plastic strain contour for numerical model without incorporating strain softening.

figure for comparison. For the first penetration up to a depth of 5D, the numerical result agrees rather well with the experimental data. With an increasing number of uplift-repenetration cycles, the numerical model correctly predicted that both the uplift and penetration resistances decrease due to increasing amount of plastic strain being accumulated in the numerical model. After 8 cycles, the load-penetration curves appear to stabilize towards a steady-state. However, the trends of the numerical load-penetration curves deviate substantially from that observed in the experiments. It appears that the empirical strain softening model defined by Equation 3 does © 2011 by Taylor & Francis Group, LLC

MITIGATION MEASURE

6

CONCLUSIONS

An accurate assessment of the vertical pipe response is critical to the long-term fatigue analysis at the touchdown zone of the steel catenary risers. A centrifuge model test on a pipe section shows that the repeated loading on the seabed significantly degrades the undrained shear strength of the seabed. Further centrifuge model tests conducted with T-bar penetrometer and spudcan have confirmed this observation.

351

The trends of the numerically predicted loadpenetration curve deviate from the experimental observation. Therefore, the empirical strain softening model adopted in the numerical model might not represent the degradation characteristic of this particular clay used in the model tests. ACKNOWLEDGEMENTS The authors acknowledge the research funding provided by A*STAR SERC (R-264-000-225-305) and MPA (R-264-000-225-490). The first author would like to gratefully thank the National University of Singapore for the award of a research scholarship. Assistance provided by the centrifuge technician is acknowledged. The third author also would like to thank Professor Mark Cassidy for his permission to publish the T-bar test result. REFERENCES Bienen, B., Gaudin, C. and Cassidy, M.J. (2009). The influence of pull-out load on the efficiency of jetting during spudcan extraction. Applied Ocean Research, 3, 202–211. Chung, S.F., Randolph, M.F. and Schneider, J.A. 2006. Effect of penetration rate on penetrometer resistance in clay. Journal of Geotechnical and Geoenvironmental Engineering, 1188–1196. Damgaard, J.S. and Palmer, A.C. 2001. Pipeline stability on a mobile and liquefied seabed: a discussion of magnitudes and engineering implications. Proceedings, 20th International conference on Offshore Mechanics and Arctic Engineering, Rio de Janeiro, 2001. Goh, T.L. 2003. Stabilisation of an excavation by an embedded improved soil later. PhD thesis, National University of Singapore.

© 2011 by Taylor & Francis Group, LLC

Osborne, J.J. 2005. Are we good or are we lucky? Managing the mudline risk. OGP/CORE Workshop – the Jackup Drilling Options & Ingredient for Success, National University of Singapore.Palmer, A.C., Hutchinson, G. and Ells, J.E. 1974. Configuration of submarine pipelines during laying operations, American Society of Mechanical Engineers, Journal of Engineering for Industry, 96, 1112–1118. Palmer, A.C. and Niedoroda, A.W. 2005. Ice gouging and pipelines: unresolved questions, Proceedings, Eighteenth International Conference on Port and Ocean Engineering under Arctic Conditions, Potsdam, NY, 1, 11–21, 2005. Palmer, A.C. and King, R.A. 2008. Subsea Pipeline Engineering. Pennwell, Tulsa, OK, second edition. Palmer, A.C. 2008. Touchdown indentation of the seabed. Applied Ocean Research, 30 (3) 235–238. Palmer, A.C. 1999. Speed effects in cutting and ploughing. Geotechnique, 49, (3) 285–294. Randolph, M.F. and Houlsby, G.T. 1984. The limiting pressure on a circular pile loaded laterally in cohesive soil. Geotechnique 34 (4) 613–623. Randolph, M.F. & House, A.R. 2001. The complementary roles of physical and computational modeling. International Journal of Physical Modelling in Geotechnics, Vol. 1, No. 1, pp. 1–8. Roscoe, K. H., Burland, J. B. 1968. On the generalized stress – strain behaviour of wet clay. Engineering Plasticity, (eds. J. Heyman, F. A. Leckie), Cambridge Univ. Press 535–609 Stewart, D.P. and Randolph, M.F. 1991. A new site investigation tool for the centrifuge. Proc. Int. Conf. on Centrifuge Modelling, Boulder, Colorado, 1991, 531–538. Taylor, R.N. (1995), Geotechnical Centrifuge Technology. London: Blackie Academic and Professional. Wang, D., White, D. J. and Randolph, M. F. 2009. Numerical simulation of dynamic embedment during pipe laying on soft clay, Proceedings, 28th International Conference on Ocean, Offshore and Arctic Engineering, Honolulu, USA, 31 May – 5 June, 2009.

352

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

A new interpretation of the simple shear test H.A. Joer, C.T. Erbrich & S.S. Sharma Advanced Geomechanics (AG), Perth Australia

ABSTRACT: The simple shear test is becoming more popular and attractive in offshore soil characterization due to the relatively small size of the sample required and the mode of shearing which closely resembles the deformation condition imposed under many offshore foundation systems.The traditional method of interpretating a simple shear test assumes that complementary shear stress act on the vertical boundary of the sample which is not generally the case especially at large strain level. A new interpretation of the simple shear test is proposed in this paper based on the observed response from a large number of tests performed in the last 15 years. This new interpretation method can be used to estimate the normal and shear stresses acting on a pre-defined failure plane, selected based on the observed failure mechanism through the sample. This provides additional insight into certain potentially important aspects of soil behaviour such as undrained shear strength, progressive failure, and the soil friction angle. Test results obtained from different soil types (clay to cemented calcareous soils) are presented using the new interpretation method and the traditional interpretation.

1

INTRODUCTION

The undrained shear strength of soil is a crucial parameter used in the offshore industry and is commonly determined in the laboratory using various types of testing devices, such as the triaxial or the simple shear. Although the triaxial test is well established and accepted in industry, the simple shear test is becoming more popular and attractive. This is mainly due to the relatively small required sample size and the mode of shearing of the sample which closely resembles the deformation condition imposed under many offshore foundation systems. In general, the interpretation of a simple shear test assumes that complementary shear stresses act on the vertical boundaries of the sample. However, this is not possible in practice. In addition, examination of a large number of test results obtained using the simple shear apparatus at the University of Western Australia (UWA) on many samples revealed a different failure mode in most cases compared to that assumed in the traditional interpretation method. This paper presents a new method (the “AG method”) for the interpretation of simple shear test results. The AG method is developed based on the observed failure response of different types of materials tested over the last 15 years in the simple shear apparatus at UWA. Typical responses obtained in different types of soils (clay, carbonate silty sand and cemented carbonate soil) using the traditional and AG method are presented in this paper. It should be appreciated that for both the traditional and AG interpretation methods presented in this paper, it is assumed that the stresses acting in the sample are uniform. However, it is widely discussed in the © 2011 by Taylor & Francis Group, LLC

353

literature that the stresses acting in the sample may not be as uniform as assumed (Airey and Wood 1987; Budhu, 1985; Joer et al., 1992; Reyno et al., 2005). The implication of such stress non-uniformity is beyond the scope of this paper. 2

EQUIPMENTS AND TEST CONDITIONS

Over the years, several different types of simple shear devices have been developed. The most commonly used in industry are the SGI apparatus (Kjellman, 1951), the Geonor apparatus (NGI) and the Berkley type simple shear apparatus. In the SGI apparatus (Kjellman, 1951) a series of concentric rings are used around the sample in order to transfer the shear strain from the top to the bottom of the sample, whereas the Geonor apparatus (Bjerrum and Landva, 1966) uses a membrane reinforced by helical wire to constrain the sample. In commercially available SGI and NGI type apparatus, tests can be performed under constant volume condition (no lateral strain) but the horizontal effective stress cannot be controlled or measured. In the UWA apparatus an unreinforced membrane is used to confine the sample. This means that the initial boundary conditions imposed on a sample are similar to those of a triaxial test (i.e both the horizontal effective stress, σh
and vertical effective stress, σv
applied to the boundary of the sample can be independently controlled and measured), which allows tests to be performed under controlled stress and deformation conditions. However, the unreinforced membrane does offer a lesser overall restraint to the vertical boundaries of the sample.

(shear band; highlighted by white dotted lines) is most pronounced in the cemented material. Where a clear failure plane develops as described above, it is more appropriate to consider the stresses that actually act on the inclined failure plane, rather than focus on the shear stresses applied by the apparatus parallel to the sample ends, as in the traditional interpretation In addition, the traditional interpretation assumes that a complementary shear stress acts on the vertical boundaries of the sample, although it is clear that such a shear stress cannot be imposed with the ’Berkeley’ type (or the NGI type) of test apparatus. Hence the new interpretation considers an alternative mechanism for balancing the shear stress imposed across the end platens. Based on the above considerations Figure 3 presents a free body diagram of the sample and the stresses that are actually believed to act on the boundaries. The applied conditions on the outside boundary of the sample are the average vertical effective stress (σv
), the average horizontal effective stress (σh
) and the shearing load (Fs ) across the platens. A differential vertical stress (σv
) is assumed to act across the width of the sample in order to balance the moment generated by Fs across the platens. Importantly, whilst this will modify the local vertical stress distribution in the sample, the average vertical stress across the whole width remains unaltered.

Figure 1. Typical failure mode (Wood, 1990).

2.2 Traditional interpretation For the traditional interpretation, the imposed stresses acting on the outside boundary of the sample are used directly, with the assumption that complementary shear stresses arise on the vertical sample boundaries. The following two equations show the stress conditions assumed around the sample: Figure 2. Typical failure planes in simple shear.

2.1 Typical failure modes Wood (1990) presented different modes of failure which may occur in simple shear tests as shown on Figure 1. In Mode A, horizontal rupture planes are generated, while in Mode B inclined rupture planes are formed during shearing of the specimen. These two failure modes are possible in the SGI and Geonor type of device where the cylindrical shape of the specimen is maintained constant and the use of concentric rings or reinforced membrane imposes relatively uniform deformations over the height of the sample. However, recent studies have demonstrated that the ‘Berkeley’ type simple shear apparatus used at UWA usually results in a diagonal failure plane through the sample, generally from the back of the top platen to the front of the bottom platen. Figure 2 presents typical photographs of such observed failure planes, as obtained for a cemented soil and for a silty sand sample. As expected, the developed failure plane © 2011 by Taylor & Francis Group, LLC

Using these stresses, the mean stress (s
) and the plane strain shear stress (t) are deduced as follows:

An estimate of the effective friction angle (φ
) may then be deduced as follows:

Where α is the plane strain friction angle determined using the t-s plot.

354

applied on the sample. As noted earlier, the differential vertical stress (σv
) averages to zero across the failure plane and hence does not influence the average resistance that is mobilised along the failure plane. During the shearing phase, the general deformations sustained by the sample are horizontal displacement of the top platen relative to the bottom platen (h) and reduction (compression) or increase (dilation) of the sample’s height (v). Considering the geometry of the sample with the failure plane extending from the back of the top platen to the front of the bottom platen, the angle of the failure plane θ can be defined as follows:

The cross sectional area of the failure plane is given by the following equation:

The friction angle (φ
) along the failure plane is the angle between the resulting load (R) and the normal to the failure plane (Figure 3c) and can be calculated as follows:

The stresses acting on the failure plane can therefore be estimated as following:

Figure 3. Illustration of the boundary conditions of the sample.

2.3 Proposed new interpretation In the proposed “AG method”, a ‘best fit’ linear failure plane is determined based on the actual failure plane observed in the sample after completion of the test, as shown for example on Figure 2. Considering the free body diagram shown on Figure 3a, the normal (σn
) and shear (τfail ) stresses acting on the failure plane are calculated based on the geometry of the sample (ie. orientation of the failure plane) and the stresses on the outside boundary of the sample. The horizontal load (Fh ) and vertical load (Fv ) acting on the failure plane can be estimated from the applied effective stresses as follows:

Where, Dsample is the diameter of the sample, Heff is the effective height and Fs is the horizontal load © 2011 by Taylor & Francis Group, LLC

3 TEST RESULTS 3.1 Typical test results Test results obtained from different types of soils (clay to cemented calcareous soils) were used to assess the difference between the traditional and AG interpretation methods.The simple shear results presented in this paper were all carried out at UWA on samples with initial diameter to height ratio of about 2. End platens with a 5 mm long perimeter ‘skirt’ were also used to prevent specimen slip across the platen interface. All the test were performed under undrained loading conditions by maintaining a constant sample height (v = 0) and constant total vertical stress (σv = 0) throughout the loading stage. This is generally achieved by stopping the vertical motor during shearing in order to maintain a constant sample height while adjusting the cell pressure to maintain constant vertical stress.

355

Figure 4. Typical simple shear results (cemented soil). Figure 6. Typical simple shear results (kaolin clay).

Figure 5. Typical simple shear results (carbonate silty sand).

Figure 4 presents plots of the shear stress (τfail and τxy ) obtained from both interpretation methods versus the shear strain (γ). Similar plots for a silty sand sample and a Kaolin clay sample are presented on Figure 5 and Figure 6, respectively. Comparing results obtained from the traditional and AG methods, it can be noted that the peak undrained shear strength derived from τfail may be higher than the peak τxy (as for cemented soil and kaolin clay) or about the same (as for carbonate silty sand). In some cases it can be lower as well. However it is worth noting that for the cemented soil and the carbonate silty sand, postpeak strain softening is much more evident when using the AG method, leading to lower ‘residual’ strengths at the end of the test. It is worth mentioning that τfail as such has no special meaning until the shear plane forms, but is important for the peak and post-peak response. With the AG method it is important to appreciate that at the start of shearing τfail is not zero for any of the tests shown on Figure 4 to Figure 6. This is because anisotropic consolidation was imposed in all cases which results in a shear stress mobilised on the ultimate failure plane during the consolidation stage. It should also be noted that at the start of loading, the vertical and horizontal stress are very close to the principal stresses. However, as τxy is increased the principal stress directions continuously change © 2011 by Taylor & Francis Group, LLC

356

Figure 7. Effective stress paths (cemented soil).

(Joer et al., 1998; Arthur et al., 1980) and the vertical sides tilt. 3.2 Assessment of friction angle For the traditional interpretation method, equation 4 may be used to derive friction angles from simple shear data, provided complementary stresses are assumed to exist. However, the friction angle may be directly estimated using the AG method without any such artificial assumptions:

Figure 9. Effective stress paths (kaolin clay). Figure 8. Effective stress paths (carbonate silty sand).

The stress paths obtained using the traditional and new interpretation methods are presented on Figure 7 to Figure 9 for the cemented sand, carbonate silty sand and kaolin clay discussed above. The results are presented in terms of shear stress (τfail ) versus normal stress (σn
) using the AG method and in terms of shear stress (t) versus mean stress (s
) using the traditional interpretation method. Best fit friction angles (φ
) deduced assuming no cohesion intercept are also shown on Figure 7 to Figure 9. In general, it can be noted that the φ
values obtained using the traditional interpretation method are higher than the values obtained using the AG method. It should be appreciated that a very high peak fric
) is shown for the cemented soil. This tion angle (φpeak is due to the true cohesion resulting from the cementation. The magnitude of this true cohesion could not be determined in the absence of more test results but any increase above zero, as assumed, would reduce the
true φpeak . The residual friction angle obtained at the
end of the test (φres ) was computed to be 43◦ or 59◦ with the AG and traditional methods respectively. The
obtained using the AG method is consisvalue of φres tent with values for similar material in the absence of cementation. © 2011 by Taylor & Francis Group, LLC

For the uncemented carbonate silty sand (Figure 9),
a φpeak value of 46◦ was determined using the AG method, which is in line with expectation for this kind
of material. This contrasts with the excessive φpeak value of 63◦ deduced using the traditional method. However, for kaolin clay (Figure 10) generally sim
values of 33◦ and 34◦ were obtained using ilar φpeak the AG and traditional methods, respectively, which are both consistent with the values reported in Lehane et. al. (2009). An alternative measure of friction angle may also be determined using conventional shear stress and ver
tical effective stress (tan φ
= τxy /σvo ). Peak friction angles of 50◦ , 39◦ and 29◦ were obtained for cemented soil, carbonate silty sand and kaolin clay, respectively with this approach. However, the peak friction angles were mobilized at the end of the test irrespective of the type of material, which is not consistent with expectation, especially for cemented soil and carbonate silty sand.

3.3

Friction angle comparison; simple shear and triaxial tests

Figure 10 presents typical stress paths obtained from monotonic simple shear tests performed on calcareous silty sand from offshore Western Australia. Also

357

an assumed diagonal failure plane through the sample and without accounting for unrealistic complementary shear stresses. These assumptions are supported by experimental data collected over the past 15 years, using the Berkley type simple shear apparatus developed at UWA. The results obtained from different soil types are presented using both the new AG and traditional interpretation methods. For sands and cemented soils, the mechanism with a well defined diagonal failure plane appears generally more appropriate. The choice of the interpretation method to be used should be selected on a case-by case basis based on the observed deformation mode in each case. However, the actual failure mechanism obtained in a soft clay sample is possibly a combination of both a discrete diagonal shear plane and the conventional type of simple shear deformation. This may explain why the friction angles interpreted from the AG and traditional methods are similar in this case. REFERENCES

Figure 10. Effective stress paths.

shown are the average peak friction angle as determined from triaxial tests (the dashed line). Both the traditional and AG interpretations of friction angle are presented. It can be seen that the simple shear stress paths obtained using the AG method are consistent with the average peak value obtained from triaxial tests (Figure 10a). However, the traditional interpretation of the same simple shear data (Figure 10b) implies a friction angle that is well above the average peak value obtained from the triaxial tests. A large number of simple shear and triaxial tests have been performed by AG over the years on various materials. Triaxial friction angles and simple shear friction angles calculated using τfail have generally been found to be consistent. 4

Airey, D. W. & Wood, D. M. 1987. An evaluation of direct simple shear tests on clay. Géotechnique 37 (1), 25–35. Arthur, J.R.F., Chua, K.S., Dunstan T. and Rodriguez del C, J.I., 1980. Principal Stress Rotating: A missing Parameter. Journal of Geotechnical Engineering, ASCE, 106, CT4, 419–433. Bjerrum, L. & Landva, A. 1966. Direct simple shear tests on a Norwegian quick clay. Géotechnique 16 (1), 1–20. Budhu, M. 1984. Non-uniformities imposed by simple shear apparatus. Canadian Geotechnical Journal 20, 125–137. Budhu, M. 1985. Lateral stresses observed in two simple shear apparatus. Journal of Geotechnical Engineering, ASCE, 111, GT6, 698–711. Joer, H.A., Lanier, J., Desrues, J. & Flavigny E., 1992. “1γ2ε”: A new shear apparatus to study the behaviour of granular materials. Geotechnical Testing Journal ASTM, 15 (2), 129–137. Joer, H.A., Lanier J. and Fahey M. 1998. Deformation of granular materials due to rotation of principal axes. Géotechnique 48 (5), 605–619. Kjellman, W. 1951. Testing the shear strength of clay in Sweden. Géotechnique 2 (3), 225–232. Lehane, B.M., o’Loughlin, C.D., Gaudin, C. and Randolph, M.F. 2009. Rate effect of penetrometer resistance in Kaolin. Géotechnique 59 (1), 41–52. Reyno, A. J., Airey, D. W. & Taiebat, H. A. 2005. Influence of height and boundary conditions in simple shear tests. In Frontiers in Offshore Geotechnics ISFOG 2005, Proc. Int. Symp., Perth, 19–21 September 2005. Rotterdam, Balkema. Wood, D. M. 1990. Soil behaviour and critical state soil mechanics. Cambridge University press. ISBN0-52133249-4.

CONCLUSIONS

A new method of interpreting the results from simple shear results has been proposed in this paper, based on

© 2011 by Taylor & Francis Group, LLC

358

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Physical modelling of the crushing behaviour of granular materials H.A. Joer & S.S. Sharma Advanced Geomechanics (AG), Perth Australia

ABSTRACT: The grading curves for the granular materials which are susceptible to crushing continuously change during loading and deformation. Such microscopic changes significantly affect the engineering behaviour of these materials. This paper presents the laboratory test data obtained from high pressure oedometer and triaxial tests performed on two different types of calcareous soil that represent different formation conditions and the identical glass balls of two different diameters. The crushing behaviour is then quantified using modified Hardin criterion based on results obtained from PSD analyses carried out on the original soils and the soils after testing. The results clearly show that the relative breakage is related to void ratio of the sample at the end of the tests and can be uniquely represented using the maximum breakage (Br max ) and minimum void ratio corresponding to onset of breakage (em ) both of which are found to be related to the breakage potential of the soil.

1

INTRODUCTION

The crushability of soil can have important consequences for engineering design. This is particularly the case for calcareous sediments which are generally comprised of the remains of marine organisms that tend to crush relatively easily under loading and deformation compared to their siliceous counterparts. The low friction capacity of piles driven into calcareous sediments is a typical example which is attributed to the grain rearrangement and crushing of the particles during pile penetration. In order to examine the crushing behaviour of different granular materials, oedometer and triaxial tests were carried out on two different types of calcareous soils, a coastal aeolian calcareous soil taken off the beach of Ledge Point (LP), 100 km north of Perth and a fine-grained calcareous soil recovered from the North West Shelf of Western Australia (NR). In addition, oedometer tests were performed on identical glass balls of two different sizes of 0.5 mm and 5 mm diameters. The oedometer tests were performed to high pressures (55 MPa for soils and 117 MPa for glass balls), while consolidated isotropic undrained (CIU) and drained (CID) triaxial tests were undertaken on samples prepared at various densities and at different stress levels. Particle size distribution (PSD) analyses were carried out on the original soils and all the soils and glass balls carefully recovered after testing. This paper is comprised of two parts. In the first part, the experimental results obtained from the oedometer and triaxial testing are presented along with the PSD results obtained from both the original and tested materials. In the second part, the relative breakage of the materials was examined using the Hardin criterion originally proposed by Hardin (1985) and later modified by Joer et al (1997). Based on the © 2011 by Taylor & Francis Group, LLC

test results, a simplified framework to describe the crushing behaviour of different granular materials is proposed. 2

MATERIALS TESTED

The original grading curves of the LP and NR soils and the two uniform size glass balls are shown on Figure 1. It should be noted that two different batches of LP soils were used for this study. The old batch was used for oedometer tests and the new batch was used for triaxial tests. Figure 1 clearly shows the uniform grain size of the LP soil, with a median grain size (D50 ) of about 0.32 mm, where NR soil contains a large range of particle size with a D50 of about 0.1 mm. Based on the Unified Soil Classification System, the LP soil can be classified as sand (SP) and NR soil as well graded silty sand (SM). Figure 2 shows photographs obtained from microscopic analyses of the LP and NR soils. It can be observed that LP soil comprised mainly coarse rounded particles, while NR soils comprised particles of various shape identified as skeletal grains, detrital grains, coated grains, pallets and lumps.

359

3 TEST RESULTS 3.1

1-D compression response

One dimensional compression tests were carried out on both LP and NR soils and on glass balls of 0.5 mm and 5 mm diameters. The tests were performed using a high pressure oedometer mould. The vertical load was applied using a 100 kN capacity compression load frame and the displacement was measured using an LVDT attached to the loading rod. The tests were

Figure 1. Original gradation curves for soils and glass balls.

Figure 3. Compression response of (a) LP soil and (b) NR soil.

maximum vertical stress of 55 MPa are presented on Figure 3. The results clearly show decreasing void ratio with increasing stress level with the tests performed at different initial void ratio converge to a unique compression line at high stress level. This response is similar to that observed for other soils available in the literature (Pestana & Whittle, 1995). Figure 4 shows the compression response of the glass balls of 0.5 mm and 5 mm diameters. The results showed an initial stiff elastic behaviour up to a yield point. This was followed up by large deformations with practically constant stress, before the stress increased again. The transition from elastic to plastic behaviour is sharper than the response observed for LP and NR soils. The yield stress was reached earlier (14 MPa) for the 5 mm diameter balls compared to 0.5 mm diameter balls (40 MPa).

Figure 2. Micrographs of (a) LP Soil and (b) NR Soil (magnification × 50).

terminated at different stress levels with the maximum stress varying from about 1.8 MPa to 55 MPa for soils and from 9.5 MPa to 117 MPa for glass balls. Typical results obtained from LP and NR soils are presented on Figure 3 in terms of void ratio versus vertical effective stress. It should be noted that only the results obtained from the tests performed to the © 2011 by Taylor & Francis Group, LLC

3.2

Monotonic shearing response

A series of CID and CIU triaxial tests were carried out on the LP soil at three different initial densities (ρdi ) of 1.23 t/m3 , 1.32 t/m3 and 1.4 t/m3 and at confining stresses (p
o ) of 100 kPa, 500 kPa and 1,000 kPa.

360

Figure 4. Compression response of glass balls.

Figure 6. CID test results for LP soil (ρdi = 1.32 t/m3 ).

results from tests performed at ρdi of 1.32 t/m3 are presented here, while results from other tests can be found elsewhere (Joer, 2002). In general, the CIU tests show an initial contractive response (increase in pore pressure) until a phase transformation point, which is followed by dilative response (decrease in pore pressure), which continued until the peak deviator stress was mobilised at large strain (greater than about 25%). This response is similar to other most commonly available granular materials (e.g. Hyodo et al 1998). The CID tests show contractive response throughout the loading stage for the test performed at 1000 kPa. However, the test performed 100 kPa and 500 kPa showed initial contractive response followed by dilative response with the level of compression dependent on the stress level.

Figure 5. CIU test results for LP soil (ρdi = 1.32 t/m3 ).

3.3 Typical results obtained from these tests are presented on Figure 5 (CIU) and Figure 6 (CID). The CIU results are presented in terms of deviator stress (q = σ1
− σ3
) and excess pore pressure (u) versus axial strain (ε1 ), while the CID test results are presented in terms q versus ε1 and volumetric strain (εv ) versus ε1 . It should be noted that only the © 2011 by Taylor & Francis Group, LLC

361

Crushing response

In order to examine the crushing behavior of soils and glass balls, PSD analyses were carried out for all the samples both prior to and after testing. Figure 7 shows the PSD curves for the LP and NR original soils and for the soils after testing in the oedometer to different stress levels. Similar plots for glass balls are presented on Figure 8. The results

Figure 9. PSD for LP soil before and after CIU and CID tests.

conditions. Typical results for the original soils and the soils at the end of the shearing tests are presented on Figure 9 for the medium dense LP soil. In general the PSD curves show that D50 reduces with increasing initial confining stress level and vice versa. Similar results were also obtained for the tests performed on loose and dense LP soil (Joer, 2002). 4 ASSESSMENT OF CRUSHING 4.1 Figure 7. PSD curves for LP and NR soils before and after oedometer tests.

Definition of relative breakage

The degree of crushability was quantified using the relative breakage (Br ) originally proposed by Hardin (1985) and later modified by Joer et al (1997) as:

where Bp is the breakage potential and is defined as the area between the original grading curve of the soil and the size of particles of 0.075 mm for Hardin method (Hardin 1985) and 0.005 mm for modified Hardin method (Joer et al 1997), whereas Bt is defined as the area between the original grading curve and the final grading curve. These two quantities are illustrated in Figure 10 for modified Hardin method. In this paper, the crushability of different materials was assessed using the modified Hardin method (Br mod ) proposed by Joer et al. (1997). 4.2 Figure 8. PSD for glass balls before and after oedometer tests.

clearly show that D50 of the materials reduces with increasing stress levels. PSD analyses were also carried out for all the samples subjected to CIU and CID triaxial loading © 2011 by Taylor & Francis Group, LLC

Influence of state on crushing

The results obtained from the LP and NR soils are presented on Figure 11 in terms of modified relative breakage (Br mod ) versus the void ratio at the end of the tests (efinal ). It should be noted that for the oedometer test efinal is the void ratio corresponding to the maximum vertical stress at which the test was terminated, whereas for CIU and CID test efinal is the void ratio at the end of the shearing (ε1 ranging between 31% and

362

Figure 12. Modified relative breakage for glass balls. Table 1.

Figure 10. Definition of relative breakage.

Summary of Br max and em for soils and Glass Balls.

Materials

Br max

em

LP NR 0.5 mm Glass Ball 5 mm Glass Ball

0.62 0.44 0.68 0.90

1.29 0.81 0.64 0.71

Figure 11. Modified relative breakage for LP and NR soils.

32%). Given that all the tests were performed to very large strain, it is believed that efinal is close to the critical state void ratio for CIU and CID tests. Similar plots for the test performed on glass balls are presented on Figure 12. The results clearly show a linear relationship between Br mod and efinal for both the LP and NR soils and also for the glass balls. Furthermore, it is interesting to note that Br mod and efinal relationship was found to be unique for a particular soil irrespective of the type of tests (compression or shearing and also between undrained or drained shearing). Further investigation indicates that a linear relationship between Br mod and efinal can be established. Figure 14 represented a schematic diagram representing a linear relationship between Br mod and efinal and the two parameters Br max and em . Where, Br max is the maximum breakage which would occur should the final void ratio reaches zero (efinal = 0) and em is the minimum void ratio at the onset of the breakage (Br mod = 0). It should be noted that efinal may be © 2011 by Taylor & Francis Group, LLC

Figure 13. Relationship between Br mod and efinal .

greater than em and in that case no breakage would occur (Br mod = 0 for efinal > em ). Values of Br max and em deduced from the tests presented on Figure 11 (LP and NR soils) and on Figure 12 (glass balls) are summarised in Table 1. Normalized breakage (Br mod /Br max ) versus normalized void ratio (efinal /em ) are plotted on Figure 13 for all the tested materials. It is interesting to note that all the test results fall within a narrow band and can be represented using the following relationships:

363

4.3

Influence of fines content

In order to investigate the effect of fines content on the crushing behaviour, oedometer tests to a maximum stress of 117 MPa were performed on samples

Figure 14. Relationship efinal /em .

between

Br mod /Br max

(CIU and CID) tests. Tests were performed on two different types of calcareous soil (LP and NR) and identical glass balls of two different sizes of 0.5 mm and 5 mm diameters. In general, the behaviour of both the LP and NR soils was similar to that reported in the literature for other granular materials while the compression response of glass balls was found to be similar to that of structured soils. However, both the LP and NR soils and the glass balls showed high susceptibility to crushing during loading at high stresses. The crushing behaviour of the soils and glass balls was investigated using the PSD results obtained for both the original soils and the soils after testing. In general relative breakage of the soils was found to increase with increasing stress level. However, careful examination of the results clearly shows that the breakage of the soils is uniquely related to void ratio of the sample at the end of the tests. Further analysis indicates the maximum breakage (Br max ) and minimum void ratio corresponding to onset of breakage (em ), both of which are related to breakage potential of the soils, are unique parameters for each granular material and can be used to normalize the results obtained from different granular materials. The breakage of granular materials (as modeled by the mixtures of two sized glass balls) is progressive by nature. The tests performed on the mixture of the two size balls presented above show that the particles which constitute the majority of the sample tend to break first.

and

ACKNOWLEDGEMENTS The work described here was carried out while the authors were working at the Centre for Offshore Foundation Systems, funded through the Australian Research Council’s Research Centres Program.

Figure 15. Relative breakage for glass balls.

prepared by mixing different percentages of 0.5 mm and 5 mm glass balls. Values of Br mod obtained from these tests are presented on Figure 14 together with the initial void ratio (eini ) of the tested samples. This figure shows that both Br mod and einitial reduce for samples containing 0 to 50% of 0.5 mm glass balls and increase for samples containing 50% to 100% of 0.5 mm glass balls. This indicates that in the mixture of 0 to 50% of 0.5 mm balls, the larger particles tend to experience breakage first and the smaller particles tend to fill the voids with minimum contact. As the percentage of smaller particles increases over 50%, this trend is reversed and the smaller particles will experience breakage before the larger particles. 5

CONCLUSIONS

The crushing behaviour of granular materials was investigated through a series of oedometer and triaxial

© 2011 by Taylor & Francis Group, LLC

364

REFERENCES Hardin, B.O. (1985). Crushing of soil particles. Journal of Geotechnical Engineering, ASCE, 111 (10), 1177–1192. Hyodo, M., Hyde, A.F.L & Aramaki, N. (1998). Liquefaction of crushable soils. Géotechnique 48 (4), 527–543. Joer, H.A. (2002). Physical Modelling of the crushing behaviour of soils. Intl. Sym. On Geotechnical Engineering and Hi-technology. Ube Yamguchi, Japan, Joer, H.A., Ismail, M.A. & Randolph, M.F. (1997). Compressibility and crushability of calcareous soils. Proc. of 15th Australasian Conf. on the Mechanics of Structures and Materials, Melbourne, 8–10 December, 514–524. Pestana, J. M. & Whittle A.J. (1995). Compression model for cohesionless soils. Géotechnique 45 (4), 611–631.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

New evidence for the origin and behaviour of deep ocean ‘crusts’ M.Y-H. Kuo & M.D. Bolton University of Cambridge, Cambridge, UK

A.J. Hill BP Exploration, Sunbury, UK

M.J. Rattley Fugro Ltd, Wallingford, UK

ABSTRACT: In situ tests in deep water West African clays show crust-like shear strengths within the top few metres of sediment. Typical strength profiles show su rising from mud-line to 10 kPa to 15 kPa before dropping back to normally consolidated strengths of 3 kPa to 4 kPa by 1.5 m to 2 m depth. A Cam-shear device is used to better understand the mechanical behaviour of undisturbed crust samples under pipelines. Extremely variable peak and residual shear strengths are observed for a range of pipeline consolidation stresses and test shear rates, with residual strengths approximating zero. ESEM of undisturbed samples and wet-sieved samples from various core depths show the presence of numerous randomly-located groups of invertebrate faecal pellets. It is therefore proposed that the cause of strength variability during shear testing and, indeed, of the crust’s origin, is the presence of random groups of faecal pellets within the sediment. 1 1.1

INTRODUCTION Background

As oil and gas exploration continues to expand into deeper waters off the west coast of Africa, the phenomenon of deep sea clay crusts is becoming more familiar to industry and academic research. Installation of oil flowlines in these areas causes partial embedment into the crust material resulting in the need to understand the medium- and long-term behaviour of these sediments under hot pipeline conditions. Installed pipelines undergo several hundreds of thermal cycles during their operational life. Pipeline designers therefore require an understanding of the undrained shear strength (su ), which governs depth of pipe embedment, and the soil-pipe coefficient of friction (µ) along axial sections of pipe adjacent to designed-in buckles, which governs the amount of self-weight anchorage. 1.2 West African clays The clays of interest lie in water depths of between 1000 m and 2000 m and exhibit very high water contents and liquid limits as shown in Figure 1. Organic matter content ranges from 3% to 6% (Thomas et al., 2005), representing an environment with a high level of organic input. The Benguela Current bringing nutrients from the Benguela Upwelling to the south may influence the areas of interest, as may the terrestrial input from the Congo River to the north. © 2011 by Taylor & Francis Group, LLC

Figure 1. Typical shear strength profiles from T-bar testing and variation in water content with depth.

The strength of the crust was initially identified through in situ cone penetration and T-bar penetration tests used to investigate proposed pipeline alignments. Typical undrained strength profiles are shown in Figure 1; values rise to 10 kPa to 15 kPa between 0.5 m and 1 m depth before reducing sharply. Had the strength profile from Figure 1 been obtained from a terrestrial mud, the favoured geotechnical origin would have been shallow overconsolidation due to desiccation. This hypothesis is apparently not applicable in deep water. 1.3 The crust: hypothesis-testing Previous work in these sediments has cited “enhanced chemical activity of the upper layers” (De Gennaro et al., 2005 p1068) as a plausible explanation for

365

the presence of crusts. Such activity may be in the form of chemical precipitation or osmosis (e.g. Ehlers et al., 2005; Sultan et al., 2001). Rigid cementation of the sediment by precipitates (chemical or biological) could result in increased sediment shear strength whilst maintaining high water contents. However, extensive microscopy during the current investigation has not identified any such precipitation. Another possible explanation for the crust origin may be the presence of micro-organisms such as bacteria which are in great abundance in marine sediments. Bacteria may either be transported through the water column or be active within the sediment. Turley et al. (1995), Ransom et al. (1997) and Bhaskar and Bhosle (2005) consider the formation of aggregates within the water column by the production of extra-cellular polymeric substances (EPS), which may be produced by bacteria. Parkes et al. (2000) suggest that bacterial numbers may exceed 109 cells per cubic centimetre of sediment within the top metre. Their presence within the clay matrix may hypothetically influence sediment mechanical properties such as permeability and shear strength through the production of EPS. These studies have considered the water-sediment interface whereas the interest of the current investigation is material from significantly greater depth. By undertaking comparative shear tests on sterile and bacterially-inoculated samples, Kuo and Bolton (2009a) showed that the presence of the bacterium Marinobacter aquaeolei only had a minor influence on shear strength. A third possibility is the presence of burrowing invertebrates which heavily populate the upper metres of deep ocean sediments. Offshore investigations undertaken by several authors highlight the influence of such organisms in both shallow and deep waters. Rowden et al. (1998) observed the presence of an increase in mean particle size due to the “aggregation of fine particles into faecal pellets” (Rowden et al., 1998 p1354). Meadows et al. (1994) completed a detailed examination of invertebrate bioturbation in cores taken from the central Southern Pacific Ocean at water depths from 5000 to 5300 m. It was demonstrated that a significant number of open burrows ranging in diameter from 0.5 cm to 4.5 cm were present to at least 0.4 m depth. Burrows were found to be oriented both vertically and horizontally, with the most abundant being the smallest diameter burrows. Undrained shear strength was measured using a Geonor fall cone penetrometer at 0.05m intervals down-core. Results from these tests showed a rapid increase in su in several test locations, with values of up to 12 kPa at 0.4 m depth. Variations in strength were attributed to the effects of bioturbation, whereas the increase in strength with depth was suggested by Meadows and Meadows (1994) to be the result of normal consolidation of the clay due to increasing vertical effective stress. The strengths measured by Meadows et al. appear to be similar to those in Figure 1, demonstrating a degree of overconsolidation. It is suggested that the effects of bioturbation may control the trend line of strength, not just the fluctuations. The current

© 2011 by Taylor & Francis Group, LLC

investigation suggests that faecal pellets from burrowing invertebrates may be the major contributing factor to increased sediment strength within the crust. No hypothesis for the disappearance of crustal strength at 1m to 2m depth has yet been advanced. The following are suggested: 1. Crushing of pellets due to in situ vertical effective stress, equivalent to approximately 3 kPa at 1m depth, resulting in ‘resedimentation’ of originally pelletised material; and 2. Continuation of biological activity with smaller organisms or bacteria breaking-down pelletised material into smaller agglomerates and/or removing mucin membranes, therefore allowing resedimentation. If loss of crustal strength is associated with the breakdown of pellets with increasing depth, then a reduction in the percentage of pelletised material should be observed. Oedometer tests may provide evidence for crushing of pellets due to increasing vertical effective stress. Previous laboratory shear testing of West African offshore clays was conducted on fully remoulded samples. They were reconsolidated to vertical effective stresses considerably higher than in situ values, and then allowed to swell under vertical stresses equivalent to those applied by a pipeline, prior to shearing. Undrained shear strengths approximated those found at shallow depth in the crusts. Access to new undisturbed box core samples presented the opportunity to quantify the influence of sediment structure on shear strength and other mechanical properties within the crust, by comparing with remoulded and reconsolidated strengths. Visual inspection of undisturbed cores, with optical and microscopic imaging, also allowed quantification of the factors influencing sediment structure, including evidence of micro-organisms and of invertebrate activity. Environmental scanning electron microscopy (ESEM) and confocal microscopy may, for example, show the presence of polysaccharide (EPS) webs produced by bacteria within the crust, or the presence of mucous-derived membranes and macro-voids associated with bioturbated sediment. Optical microscopy, X-ray computer tomography (CT) and ESEM may also show the presence of bioturbation in the form of burrows and pelletised sediment. The goal of this research is to provide evidence for the biological origin of deep ocean crusts, and to explain their mechanical behaviour in terms of their unusual natural structure. 2 2.1

METHODOLOGY Cam-shear testing

To better understand the mechanics of the soil-pipeline interface, a series of interface shear tests were completed. The Cam-shear apparatus (Bolton et al., 2007)

366

Figure 3. Soil-soil peak strengths for variable shear speeds.

2.3 Figure 2. Cam-shear apparatus showing PTFE split-plane shear box, pipeline coating material and actuator.

shown in Figure 2 was used to create stress and interface conditions for testing soil behaviour at shallow pipeline embedment. In addition to strength measurements, interface friction values for different pipeline coating roughnesses are of interest. Shear tests were undertaken for vertical effective stresses of 2 kPa to 6 kPa using the following procedure: 1. Extrude undisturbed core sample from core liner and trim to 75 mm diameter sample prior to extrusion into PTFE shear box; 2. Place shear box onto wet pipeline coating interface and seal around edge of box to permit only upwards drainage during consolidation; 3. Place interface material and sealed sample into waterbath before applying required vertical effective stress (equivalent to pipeline stress) via filter paper and porous disk; and allow to consolidate for at least 30 hours; and, 4. Interface material and shear box are removed from waterbath and attached to actuator for shear testing. Each sample was subject to three shear speeds: 0.05 mm/s, 0.5 mm/s and 0.005 mm/s, corresponding to approximate descriptions of medium, fast and slow shearing. A shearing speed of 0.005 mm/s was initially considered ‘slow enough’ to allow the determination of a ‘drained’ shear strength value. Internal (soil-soil) shear strengths were then measured by shearing at the split-plane of the shear box. 2.2 One dimensional compression Two oedometer tests were undertaken on undisturbed samples from 0.25 m and 0.3 m depth to determine the preconsolidation stress and subsequent v-log p
behaviour. Sample fabric was imaged using ESEM and optical microscopy and the percentage of pelletised material determined prior to and after consolidation to observe the effect of vertical effective stress on the sediment structure, in particular, crushing of pelletised material.

© 2011 by Taylor & Francis Group, LLC

X-ray Computer Tomography (CT) and ESEM

To provide a three dimensional, non-intrusive image of core samples, an X-ray CT imager was used. This allows samples of up to 85 mm diameter and 80 mm height to be imaged while remaining within the core liner. The resolution that can be achieved with samples of this volume is limited. However, internal structure including burrows and macro-voids of millimetre scale can still be detected. ESEM was used to image both fresh and autoclaved bulk samples and individual pellets in addition to preand post-test samples. All samples were imaged in moist conditions with vapour pressure and back scatter detectors to pick up secondary electrons. 3 3.1

RESULTS AND DISCUSSION Cam-shear test results

Previous testing of reconstituted and reconsolidated samples undertaken by Bolton et al. (2007) suggested a good fit to the Cam-clay model proposed by Schofield and Wroth (1968). The results obtained from current testing of undisturbed samples are therefore also interpreted using the Cam-clay model as a framework. This allows a direct comparison and quantification of the influence of ‘natural structure’ present in undisturbed samples on the shearing characteristics of the samples. Figure 3 shows the soil-soil results from Cam-shear tests on natural samples sheared beyond peak strength (to about 12mm) for a range of vertical effective stresses and shear speeds. Bolton et al. (2007, 2009) observed that an ultimate drained friction factor µ* (soil-soil shear strength divided by normal stress) of 0.6 was typical for West African clays. By accepting this value for µ*, the following observations are made for Figure 3: • The apparent preconsolidation stress σ
*c (backcalculated from corresponding Cam-clay yield surfaces for each point) may range from 6 kPa to 35 kPa for the samples tested. • An arithmetic mean of approximately σ
*c,mean = 16 kPa is suggested. • No rate-dependant trends can be inferred from the current data.

367

Figure 4. Interface peak shear strength results for smooth and rough coatings and variable shear speeds.

Figure 5. Interface residual shear strength results for smooth and rough coatings and variable shear speeds.

The variation in σc
∗ may be due to the presence of randomly distributed groups of pellets within the sample. Figure 4 presents the results of maximum strength values for interface shear tests undertaken on samples for a range of shear speeds. Two coating roughnesses were used: ‘smooth’ (5 µm) and ‘rough’ (95 µm). An adhesion factor α (shear strength observed for interface sliding divided by maximum shear strength) was assumed to be 0.8 following Bolton et al. (2007) resulting in µ = 0.8 µ∗ = 0.48. The following observations are made for Figure 4:

• σrmc ≈ 0.8σc,mean is found to define a reasonable

Figure 6. Oedometer test results for sample depths of 0.25 m and 0.3 m with comparison with De Gennaro et al. (2005).

upper bound Cam-clay surface for τmax .

• A surprising lower bound of zero is found for rough

interfaces. • Somewhat counter-intuitively, shearing on the

smooth interface produces larger strengths than rough interfaces. • Shear rate does not have a significant influence on τmax . Figure 5 presents residual strength values, τres for the corresponding τmax values in Figure 4. Residual strength was measured after shearing the samples for a distance of approximately 50 mm, at which point the sample at the soil-coating interface should be in a remoulded state. Observations made from Figure 5 are: • Residual strengths τres on a rough interface may be

as small as zero at vertical effective stresses within the range tested. • Residual strengths for a smooth interface approximate peak strengths τmax . • Shear rate appears to influence τres only for larger vertical effective stresses. 3.2

Oedometer test results

Figure 6 shows the results of two oedometer tests undertaken on undisturbed samples from 0.25 m and 0.3 m depth. These samples were subject to staged loading taking the vertical effective stress to greater than 10 kPa, approximating the in situ overburden stress at 3m depth. For comparison, results obtained by © 2011 by Taylor & Francis Group, LLC

368

Figure 7. Comparison of sample fabric pre- (left) and post(right) oedometer testing.

De Gennaro et al. (2005) from an oedometer test on a sample at 4 m depth are shown. A normal compression line determined by De Gennaro, et al. (2005) is also shown. A precompression vertical effective stress of approximately 1kPa is found for the tested samples. Figure 7 shows typical photographs of samples taken pre- and post-compression. Whereas the largest pellets have been crushed, the post-compression photograph still displays a ‘granular’ appearance. Wet sieving was undertaken of trimmings before testing and of samples after testing to determine the percentage of pellets (Figure 8). The following points are made: • Prior to testing, a significant proportion (>50%) of

the material comprises pellets; • The proportion of pellets after testing was not

significantly different; and,

Figure 9. Photograph (left) and X-ray (right) of undisturbed sample showing pellets (∼1.5 mm diameter) and macro voids.

Figure 8. Variation in percentage of pellet sizes prior and post oedometer testing. • Larger pellets may be crushed resulting in a greater

number of smaller fragments. 3.3

CT imaging and ESEM for internal structure

Figures 9 and 10 show the presence of voids at different scales within undisturbed samples and washed samples of pellets. Water-filled macro-voids (0.1 mm to > 1 mm size; Figure 9) are found adjacent to pellets and micro-voids (<1 µm to 3 µm size; Figure 10) are observed within the pellets. ESEM imaging with the variable pressure secondary electron (VPSE) detector allows the identification of ‘wet textures’ including some evidence for mucin membranes surrounding pellets. By imaging with a back scatter detector in a wet chamber, information about the structure of pellets can be resolved, showing the arrangement of clay platelets at the surface of pellets.

Figure 10. ESEM of pellet aggregates >50 µm showing micro-voids: VPSE (top), back scatter SE detectors (bottom).

3.4 Discussion of key observations Faecal pellets are described by zoologists as being “compacted” (overconsolidated or desiccated in geotechnical terms) and bound with mucous within the gut of their depositors. Their inherent robustness is demonstrated by survival during one-dimensional compression and the wet sieving process. So the crust may indeed have its origins in desiccation. Cam-shear test results shown in Figures 3 to 5 demonstrate the variability in shear strength that is © 2011 by Taylor & Francis Group, LLC

369

obtained from testing undisturbed samples. This variability is attributed to the random distribution of faecal pellets found at the sheared surface. Observed macrovoids between pellets provide a source of pore water. Shearing against a rough interface may disintegrate pellets, thereby temporarily reducing intergranular stresses and increasing pore pressures until excess water can escape. In the mean time, “hydroplaning” can occur on the trapped water film, eliminating interface friction. The abundance of faecal pellets may therefore be used to explain the shear behaviour seen in Figures 4 and 5. Even slow shearing at 0.005 mm/s does not guarantee that the peak strength will be fully drained. The pore pressures caused by pellet crushing

are slow to dissipate, even though extensive open burrow networks are observed in core samples (Kuo & Bolton, 2009b). Oedometer tests show that elements of pelletised structure remain even at vertical effective stresses exceeding the overburden stress at the base of the crust. Pellet percentages at sediment depths greater than 1 m, however, do decrease to 20% for pellets larger than 63 µm (Kuo & Bolton, 2009b). If increase in stress is not the cause, other factors such as a continuation of biological activity at greater depth may be responsible for the gradual destructuring of pellets. Given that the sedimentation process in these water depths may be at least three orders of magnitude slower than the bioturbation processes that may initially produce the crust, it is also feasible to suggest over the same time-frame but at a greater depth, a different species of invertebrates are working to remove the crust. Removal of the crust may be manifested through reworking of larger pellets into smaller agglomerates or complete destructuring of pellets by removal of mucins that bind pellets together. This may then allow ‘resedimentation’ of material within pellets that were initially overconsolidated. 4

CONCLUSIONS

This paper presents recent Cam-shear testing of undisturbed samples of West African clays from circa 1500 m water depth and from 0 m to 0.5 m soil depth. Based on CT, ESEM, optical imaging, and wet sieving, it is shown that over 50% of the undisturbed samples are in the structured form of robust faecal pellets, the proportion of which is sufficiently high to influence the mechanical behaviour. Though scatter is observed in Cam-shear data, this may simply reflect the random disposition of faecal pellets within the test samples. Very low friction values can be found especially when these natural samples are sheared against rough pipe coatings, presumably due to the destruction of fabric by asperities and the creation of excess pore pressures. Faecal pellets are therefore suggested to be the origin both of high crust strength and the significant variability of both peak and residual interface friction. It is hypothesised that a continuation of biological activity at greater depth may be a major contributing factor to the crust’s demise. ACKNOWLEDGEMENTS

REFERENCES Bolton M.D., Ganesan S.A. & White D.J. 2007. CUTS Report no. SC-CUTS-0609-R2, University of Cambridge. Bolton M.D., Ganesan S.A. & White D.J. 2009. CUTS Report no. SC-CUTS-0705-R1, University of Cambridge. Bhaskar, P.V. & Bhosle, N.B. 2005. Microbial extracellular polymeric substances in marine biogeochemical processes. Current Science, 88(1):45–53. De Gennaro, V., Delage, P., & Puech, A. 2005. On the compressibility of deepwater sediments of the Gulf of Guinea in deepwater, ISFOG 2005, Gourvenec&Cassidy (eds) Taylor & Francis Group, London. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origin of near-seafloor “crust zones” in deepwater, ISFOG 2005, Gourvenec&Cassidy (eds) Taylor&Francis Gp, Lon. Kuo, M.Y-H. & Bolton, M.D. 2009a. Soil Characterization of Deep Sea West African Clays: Is Biology a Source of Mechanical Strength? Proc. 18th ISOPE Conf., Japan. Kuo, M.Y-H. & Bolton, M.D. 2009b. On the Origin and behaviour of deep ocean “crusts”. Under review. Meadows, A. & Meadows, P.S. 1994. Bioturbation in deepsea Pacific sediments. J of Geol. Soc., 151:361–375. Meadows, P.S., ReicheltA.C., MeadowsA. & Waterworth J.S. 1994. Microbial and meiofaunal abundance, redox potential, pH and shear strength profiles in deep sea Pacific sediments. J of Geol. Soc., 151:377–390. Parkes, R.J., Cragg, B.A. & Wellsbury, P. 2000. Recent studies on bacterial populations and processes in subseafloor sediments: A review. Hydrogeology Journal, 8:11–28. Puech, A., Colliat, J.L., Nauroy, J-F. & Meunier, J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments, ISFOG 2005, Gourvenec&Cassidy (eds) Taylor & Francis Group, London. Ransom, B., Bennett, R.H., Baerwald, R. & Shea, K. 1997. TEM study of in situ organic matter on continental margins: Occurrence and the “monolayer” hypothesis.Mar.Geol.,138: 1–9. Rowden, A. Jago, C. & Jones, S. 1998. Influence of benthic macrofauna on the geotechnical & geophysical properties of surficial sediment, North Sea. Cont.Shelf Res., 18: 1347–1363. Schofield A.N. & Wroth C.P. 1968. Critical state soil mechanics. McGraw-Hill. Sultan, N., Cochonat, P., Cauquil, E. & Colliat, J.L. 2001. Apparent overconsolidation and failure mechanisms in marine sediments. Proc. OTRC Intern. Conf. Houston, Tx. Thomas, F., Rebours, B., Nauroy, J-F. & Meunier, J. 2005. Mineralogical characteristics of the Gulf of Guinea deep water sediments, ISFOG 2005–Gourvenec&Cassidy (eds) Taylor & Francis Group, London. Turley, C.M., Lochte, K. & Lampitt, R.S. 1995. Transformations of biogenic particles during sedimentation in the Northeastern Atlantic. Phil. Trans. of Royal Soc of London Series Biol. Sci., 348(1324): 179–189.

The authors are grateful to BP Exploration for providing natural core samples for testing.

© 2011 by Taylor & Francis Group, LLC

370

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Soil unit weight estimated from CPTu in offshore soils P.W. Mayne Georgia Institute of Technology, Atlanta, Georgia USA

J. Peuchen & D. Bouwmeester Fugro, Leidschendam, The Netherlands

ABSTRACT: Empirical expressions relating soil unit weight with Cone Penetration Test (CPT) measurements have been explored using complementary sets of laboratory and field data from 14 offshore sites. The most significant trends found that the total unit weight (γt ) increases with measured CPT resistances, yet also decreases with plasticity characteristics. A prior study from 44 sites (2 offshore and 42 onshore soils) showed a strong relationship between unit weight, sleeve friction, and effective overburden stress. This trend was further corroborated with an offshore data set. For normally-consolidated clays that constituted 12 of the offshore sites, the cone resistance increased linearly with depth below seafloor and defined a new parameter (designated mq = qt /z). For these cases, γt was linked to the slope parameter mq and cone resistance (either total, net, or effective resistance), or alternatively, sleeve friction. 1

INTRODUCTION

A knowledge or assessment of the profile of soil unit weight is a necessary first step in the evaluation of geotechnical parameters, including strength (τmax ), small-strain shear modulus (G0 ), stress history (OCR), and lateral stress state (K0 ). For the post-processing of Cone Penetration Test (CPT) measurements, many relationships are based on the net cone resistance, qn = qt − σvo , where qt = total (corrected) cone resistance and σvo is the total vertical overburden stress. In fact, σvo is the accumulation of unit weights times layer thickness with depth. The relative effect of σvo on qn is significant particularly for normally consolidated and slightly overconsolidated soils. While a few prior studies have assigned soil unit weights on the basis of CPT charts using soil behavioural types (SBT) or soil classification zones (e.g. Larsson & Muladi´c 1991; Lunne et al. 1997), the direct assessment of unit weight from measured CPT resistances (qt , fs , u2 ) has not been fully addressed.

2

DATABASE PREPARATION

2.1 Offshore sites A database was prepared from 14 offshore deposits that were subjected to comprehensive in-situ testing, sample logging and laboratory testing. The offshore sites were located in: south Atlantic Ocean (2 sites), Gulf of Mexico (3 sites), east Indian Ocean, west Australia, west Africa, Mediterranean Sea, Baltic Sea, Caspian Sea, and the North Sea. At two sites (North Sea and Australia), separate upper and lower layers © 2011 by Taylor & Francis Group, LLC

371

Figure 1. Summary of measured unit weight profiles for 14 offshore deposits.

were identified, thus giving a total of 14 geomaterials for study. Water depths at the sites varied from as low as 40 m to as deep as 2400 m. The majority of data were acquired in clay soils that permitted the acquisition of undisturbed tube samples. Figure 1 shows the summary profile of (saturated) unit weights with depth below seafloor. The unit weights range from about 12 kN/m3 to 20 kN/m3 (mean γt = 16.2 kN/m3 ). The reference profiles of unit weights were evaluated using mass-volume measurements from undisturbed tube samples, as well as calculated values from index water contents, assuming full saturation.

Figure 2. Example profile of unit weight by tube sample measurements and from water contents at Gulf of Mexico Site 3. Figure 4. Summary of cone resistances for 14 offshore sites.

Figure 3. Mean values of plasticity index vs. liquid limit for the offshore clay sites.

In general, both methods provided very comparable values of γt with depth. An example from the Gulf of Mexico Site 3 is shown in Figure 2, showing good agreement. Laboratory testing was used to provide index parameters at the offshore sites. The value ranges (mean) of the index parameters are as follows: density of solids (mean: ρd = 2.72 g/cc) from 2.65 to 2.76 g/cc; natural water content (w = 73%) from 23% to 200%; liquid limit (LL = 77%) from 34% to 174%; plasticity index (PI = 49%) from 20% to 108%. In a conventional plot of PI vs. LL, the mean values of data fell near to the A-Line, as seen in Figure 3. For these offshore soils, the measured calcium carbonate contents ranged from <1% to 72% with a mean value of 22%.

2.2 Offshore CPT soundings At corresponding depths of the laboratory results, corrected piezocone test (CPTu) measurements of cone resistance (qt ), sleeve friction (fs ), and shoulder © 2011 by Taylor & Francis Group, LLC

372

Figure 5. Summary of sleeve frictions at the offshore sites.

pore water pressures (u2 ) were matched for analysis. Figure 4 presents the qt profiles for these sites and groups the majority of data into the normallyconsolidated category. Cone resistances of overconsolidated (OC) clays in some cases reached 8 MPa, while for normally-consolidated (NC) clays, the values of cone resistances qt ≤ 2 MPa for depths of 40 m or less. The corresponding sleeve friction readings for these sites are presented in Figure 5. For OC deposits, fs values up to 120 kPa were observed, while for NC geomaterials, values of sleeve friction were fs ≤ 50 kPa for depths less than 40 m. Although not presented herein, the corresponding shoulder porewater pressures (u2 ) ranged up to 5 MPa for OC clays and for the NC series, all u2 readings ≤1.2 MPa at depths less than 40 m.

Figure 6. Unit weight versus effective cone resistance for offshore deposits showing influence of plasticity index.

Note that total or corrected resistances are employed in the correlations and relationships for the offshore series (Lunne, et al. 1997; Mayne 2007). Specifically, the following expressions were used for obtaining total cone resistance (qt ) and total sleeve friction (fst ):

where qc = measured cone resistance, anet = net area ratio for cone, fs = measured sleeve friction, bnet = friction sleeve factor, and u2 = measured porewater pressure. Notably, ISO (2005) recommends u2 measurement in the gap between the cone and the friction sleeve. This is not done in practice and the correction of cone resistance assumes the measured u2 value at the CPT shoulder location to be the same as the pressure in the gap (ISO 2005). This assumption is probably reasonable for offshore sites but may overpredict qt at shallow depth for onshore sites. An approximate correction of sleeve friction requires simultaneous measurements of both u2 and u3 , corresponding to shoulder and above the sleeve, respectively by a special dual-element or triple-element piezocone. This is not practical and therefore Equation (1b) has been offered as a pragmatic approximate correction (e.g. Larsson & Mulabi´c 1991; Mayne 2007). In the case of the offshore sites, nine were tested using a 15 cm2 Fugro penetrometer (anet = 0.58; bnet = 0.015) and five were tested using a 10 cm2 Fugro penetrometer (anet = 0.75; bnet = 0).

Figure 7. Relationship for unit weight from CPT sleeve friction derived from 44 various soil sites (after Mayne, Peuchen, and Bouwmeester 2010).

The latter of these two trends is presented in Figure 6. As the PI is not normally known during CPT, the above is of limited practical use. Of course, a methodology that can be used in a variety of soil types, including clays, silts, sands, and mixed soils of varying consistencies would be more useful. In a prior study of data from 44 soils (42 onshore and 2 offshore sites), it was found that a generalized trend related the soil unit weight to sleeve friction and effective overburden stress (Mayne et al. 2010). A dimensionless expression for this trend is given by:

2.3 General regression trends Processing of the data was handled by graphing, simple statistics, and multiple regression analyses, too numerous to review herein.Two of the more significant trends of that review are given by the following expressions (coef. of determination: R2 = 0.859 and R2 = 0.856, respectively):

© 2011 by Taylor & Francis Group, LLC

373

where γw = unit weight of water and σatm = atmospheric pressure (=1 bar = 100 kPa). The general trend and corresponding statistics are shown in Figure 7. Note that the standard error of the dependent variable (S.E.Y.) indicates the expected variance of the statistical trend, here shown as ±1.08 kN/m3 . Notably, the methodology does not apply well to diatomaceous geomaterials and may be of limited use in highly structured soils. There appears to be additional effects related to soil plasticity characteristics, however, the quantification was complicated because of combinatorial influences of overconsolidation, ageing, and cementation. Since Equation (4) contains an effective overburden stress term, it is necessary to assume the first value of unit weight, which can later be upgraded once the full profile is assessed (i.e. iteration).

Figure 8. Combined onshore and offshore data sets for unit
weight as function of fst and σvo .

Figure 10. Onshore and offshore CPTu data sets plotted on nine- part zone soil behavioural type chart.

Figure 9. Statistics on combined onshore and offshore CPTu data sets (58 sites total = 42 onshore and 16 offshore).

2.4

soil formation have shown rather good agreement in comparisons of fst profiles in a variety of soil types, including: stiff clays (Tumay et al. 1998), natural sands (Wise et al. 1999), silts (Finke et al. 2001), as well as soft to hard clays (Titi et al. 2000; Powell & Lunne, 2005). For the case of soft sensitive fine-grained soils, the sleeve friction values may be less certain. Here, CPT soil behavioral type (SBT) charts may be used. The full sets of data considered herein are shown on a nine-zone SBT chart developed by Robertson (1990), as presented in Figure 10. A CPT material index (Ic ) defines the zones (Robertson 2004), which can be calculated from:

Combined onshore and offshore data sets

The data from both onshore and offshore soils can be combined, as presented in Figure 8, showing excellent agreement between the two sets. Statistics on the measured versus predicted results are shown in Figure 8 showing a best-fit line with intercept = 0. It has been observed that the sleeve friction in soft sensitive clays may exhibit very low readings with high uncertainty (e.g., Lunne et al. 1997), thus some discussion is warranted with regard to the reliability of CPT sleeve friction readings. Notably, some variances may be due to limits in load sensing accuracies (e.g., Peuchen 2000). Other differences can be attributed in part to inevitable transient, differential pressures acting in the gaps at the top and bottom of the friction sleeve and to permissible geometric tolerances for the cone tip, the pore pressure filter and the friction sleeve (ISO 2005). Nevertheless, a number of studies using different types and brands of penetrometers in the same © 2011 by Taylor & Francis Group, LLC

374


where Qt = (qt − σvo )/σvo = normalized net cone resistance and Fr = fst /(qt − σvo ) = normalized friction ratio in percent. See Robertson (2009) for additional details, particularly for improved stress normalization in sands and silts. The majority (92%) of the offshore data fall into zone 3 (clays), while the remaining 8% of offshore data fall into zones 3 to 5 (clayey silts to silt mixes to sand mixes). The CPT data from primarily onshore sites fall between zone 2 (organic) up to zone 6 (sands). For the pre-screening of sensitive fine-grained soils, zone 1 may be identified when:

Figure 11. Regression fittings of cone resistances with depth below seafloor for the upper clay of North Sea site.

Figure 12. Trend of mq = q/z with liquid limit for NC clays.

3 TRENDS SPECIFIC TO SOFT CLAYS 3.1 Cone resistance profiles in NC deposits For soft Normally-Consolidated (NC) offshore clays, the data indicate some additional trends of interest. The measured cone resistances clearly show a linear increase with depth, as evidenced by Figure 4. For each of the 12 NC clay deposits, least squares (y = mx + b) and best-fit regression lines (y = mx) were fitted for the following CPT resistances with depth: (a) total cone resistance, qt ; (b) net cone resistance, qn = qt − σvo ; and (c) effective cone resistance, qe = qt − u2 . In all cases, very small intercepts were obtained, thus allowing a best-fit line (b = 0) to be assigned. Figure 10 shows an example of the regression fittings to the CPT data from the upper clay (0 m to 15 m below seafloor) for the North Sea location. A characteristic parameter from this type of analysis is the slope of resistance with depth. For simplicity, we can define:

For the upper clay of the North Sea site, the characteristic value of mq = 46 kN/m3 is found from the qt versus depth plot. Of particular additional interest is that the slope parameters from qt , qn , and qe vs. depth all decrease with soil plasticity characteristics, thus could be used as a surrogate in the aforementioned Section 2.3 correlative relationships. Figure 12 shows the trend of mq with liquid limit from the NC set of offshore clays. For the cone resistances, the following trends were determined between the water content indices, respectively (liquid limit, R2 = 0.88; plasticity index, R2 = 0.85; and natural water content, R2 = 0.82):

where LL, PI, and w are in percent. © 2011 by Taylor & Francis Group, LLC

375

Figure 13. Direct soil unit weight assessment from CPTu results in NC offshore clays.

3.2

CPT γt relationships for NC clays

Combining the aforementioned trends between total unit weight, plasticity, and cone resistance with the observed relationships between cone resistance, depth, and plasticity for normally consolidated clays allows for a direct assessment of soil unit weight from CPT. Figure 13 shows one such derivation that now ties together the soil unit weight in terms of net cone resistance and slope parameter mq . As the relationship in Figure 13 is in terms of the net cone resistance, the initial first value of unit weight is assumed for evaluating the initial overburden σvo term in order to get the calculations of γt started. Later, that assumed value is updated as necessary, once the remaining profile of γt has been obtained. Using the slope parameter defined from measured cone

resistances (mq = qt /z), the four defined empirical trends for total unit weight (γt in kN/m3 ) are:

where the CPT resistances are all in kPa and the slope parameter mq is input in units of kN/m3 . For the ratio of measured to predicted values, the corresponding coefficients of variation ( = standard deviation/mean) for these four relationships are 5.3, 5.5, 5.7, and 5.5 percent, respectively.

4

CONCLUSIONS

Results from 14 offshore deposits have confirmed a generalized trend that relates total unit weight to the CPT sleeve friction and effective overburden stress. The associated relationship was originally derived from a database compiled from 42 onshore and 2 offshore geomaterials, including: clays, silts, sands, and mixed soils, excluding diatomaceous soils and sensitive clays. The offshore series were predominantly characterized as normally-consolidated clays that showed cone resistance increasing linearly with depth and represented by a single slope parameter: mq = (qt /z). For NC clays, the unit weight can be related to slope mq and CPT resistance, as expressed in terms of either total, net, or effective cone resistance, or alternately, sleeve friction.

ACKNOWLEDGMENTS Sponsorship for this study was provided by Fugro to enhance efficient, safe, and sustainable geotechnical practice.

© 2011 by Taylor & Francis Group, LLC

REFERENCES Finke, K., Mayne, P.W. and Klopp, R.A. 2001. Piezocone penetration testing in Atlantic Piedmont residuum. J. Geotechical and Geoenvironmental Engineering 127(1): 48–54. ISO International Organization for Standardization (2005), Geotechnical Investigation and Testing – Field Testing – Part 1: Electrical Cone and Piezocone Penetration Tests, Draft International Standard ISO/DIS 22476-1:2005. Larsson, R. & Mulabdi´c, M. 1991. Piezocone Tests in Clays. Report No. 42, Swedish Geotechnical Institute, Linköping, 240 p. Lunne, T., Robertson, P.K. & Powell, J.J.M. 1997. Cone PenetrationTesting in Geotechnical Practice. EF Spon/ Blackie Academic, Routledge Publishers, London, 312 p. Mayne, P.W. 2007. NCHRP Synthesis 368: Cone penetration testing. Transportation Research Board, Washington, DC: 118 p. available from: www.trb.org Mayne, P.W., Peuchen, J., & Bouwmeester, D. 2010. Soil unit weight estimation from CPTs. Proceedings, 2nd International Symposium on Cone Penetration Testing, Vol. 2 (CD Paper 2-05), Huntington Beach, California: 169-176. Peuchen, J. (2000), Deepwater cone penetration tests. Offshore Technology Conference, OTC Paper 12094, Houston, TX. Powell, J.J.M. and Lunne, T. 2005. A comparison of different size piezocones in UK clays. Proc. 16th ICSMGE, Vol. 1 (Osaka), Millpress, Rotterdam: 729–734. Robertson, P.K. 1990. Soil classification using the cone penetration test. Canadian Geotechnical J. 27(1): 151–158. Robertson, P.K. 2004. Evaluating soil liquefaction and postearthquake deformations using the CPT. Geotechnical & Geophysical Site Characterization, Vol. 1 (Proc. ISC-2, Porto), Millpress, Rotterdam: 233–249. Robertson, P.K. 2009. Interpretation of cone penetration tests: a unified approach. Canadian Geot. J. 46 (11); 1337–1355. Titi, H.H., Mohammad, L.N. and Tumay, M.T. 2000. Miniature cone penetration tests in soft and stiff clays. ASTM Geotechnical Testing J. 23 (4): 432–443. Tumay, M.T., Kurup, P.U. and Boggess, R.L. 1998. A continuous intrusion electronic miniature cone penetration test system. Geotechnical Site Characterization, Vol. 2 (Proc. ISC-1, Atlanta), Balkema, Rotterdam: 1183–1188. Wise, C.M., Mayne, P.W. and Schneider, J.A. 1999. Prototype piezovibrocone for evaluating soil liquefaction susceptibility. Earthquake Geotechnical Engineering, Vol. 2 (Proc. IS-Lisbon), Balkema, Rotterdam: 537–542.

376

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Strain rate dependent simple shear behaviour of deepwater sediments in offshore Angola M.J. Rattley Fugro GeoConsulting, Wallingford, UK

A.J. Hill BP Exploration, Sunbury-on-Thames, UK

S. Thomas & B. Sampurno Fugro GeoConsulting, Wallingford, UK

ABSTRACT: This paper presents the preliminary results of laboratory testing on deepwater soils to determine the influence of applied strain rates on their constant volume deformation characteristics. The results of direct simple shear testing are interpreted with reference to the lithological and mineralogical components of the sediments and their arrangement in terms of soil structure. The results are examined with reference to existing laboratory consolidation test data and previous investigations into strain rate dependent behaviour. Conclusions are drawn in relation to the potential influence on the derivation of parameters for use in foundation design.

1

INTRODUCTION

undrained strength per log cycle increase in strain rate,

1.1 Background For many offshore foundations the loading period considered during design may be significantly different to that applied in laboratory tests from which the soil parameters are determined. The rheological behaviour is, then, of engineering significance and it has long been established that the stress-strain response of finegrained soil is time dependent (e.g. Bjerrum 1973, Sheahan et al. 1996). The evolution of the mechanical properties of clay with time may be better understood through laboratory testing methods incorporating stress-relaxation, creep at constant stress or strain rate variations. To this end, previous studies have focused primarily on the evaluation of rate dependence in triaxial and one-dimensional loading tests, with the latter being predominantly related to the investigation of secondary compression.

1.2 Rate dependence of undrained shear strength Strain rate effects have a particular importance in relation to the undrained strength of clay. It has been demonstrated by Graham et al. (1983) and Lefebvre and LeBoeuf (1987), amongst others, that the undrained shear strength of clay may increase by about 9–20% for a 10-fold increase in strain rate. Sheahan et al. (1996) proposed the general strain rate parameter ρε˙ a0 to describe the increase in © 2011 by Taylor & Francis Group, LLC

where su0 is the value of su at the reference strain rate, ε˙ a0 and su is the increase in undrained shear strength for an increase in strain rate  log ε˙ a . It was found in the study of Sheahan et al. (1996), based on results of triaxial tests on resedimented Boston Blue Clay, that ρε˙ a0 decreased with increasing overconsolidation ratio (OCR), although the trend was found to be less apparent at higher strain rates. The strain rate effect appears to be independent of stress level (at OCR = 1) and mode of shearing in laboratory testing (Graham et al. 1983). The influence of water content and plasticity index is less certain in the literature. Given that the rate dependent response is linked to the viscosity of the water present in the pores of the clay and the absorbed water envelope of the clay particles (Low 1961) it might be expected that higher water contents would lead to more pronounced increases due to viscous behaviour. The increased thickness of the absorbed water envelope present in high plasticity clays may also lead to higher viscosity, where the envelope itself is thought to be more viscous than the free water in the pores. This reasoning is generally supported in the data presented by Briaud & Garland (1985), compiled

377

Table 1.

General properties of the selected samples.

Sample

Depth (m)

γ’ (kN/m3 )

w (%)

wL (%)

Ip (%)

St

SC1 SC2

8.6 16.2

2.4 2.6

215 187

190 185

115 110

3 4

moisture contents above the liquid limit and organic contents of approximately 5–7%. X-ray examination and analysis of the clay activity suggests that the predominant clay minerals are smectite-rich illite and kaolinite. The average index properties of the samples are given in Table 1. 2.2

One dimensional compression

from some 150 undrained laboratory tests, although it should be noted that the available data at higher values of water content and plasticity index are somewhat limited.

A significant number of incremental loading (IL) oedometer and constant rate of strain (CRS) consolidation tests have been performed as part of a general geotechnical investigation of the OA soils in the region of interest. Examples of these results for samples taken at adjacent locations, at corresponding depths, are presented here to aid in the analysis of the strain rate dependent shear behaviour.

1.3

2.3

Figure 1. Typical plasticity index and undrained strength profiles for the offshore Angola sediments of interest.

Rate dependence of excess pore water pressure

The magnitude of excess pore pressures generated during undrained shearing will influence the measured shear strength. For this reason the response of excess pore water pressure to changes in strain rate is also of interest and it has been demonstrated that excess pore pressures in triaxial compression decrease with increasing strain rate (Sheahan et al. 1996). This trend appears to relate mainly to normally consolidated clays but is also observed in overconsolidated clays at high strain rates. The generation of excess pore pressure has been observed to be almost strain rate independent in structured clay (Lefebvre & LeBoeuf 1987). The pore pressure response to changes in strain rate during simple shear seems to be less well reported in the literature. 2 2.1

EXPERIMENTAL PROGRAMME Offshore Angola sediments

The offshore Angola (OA) clays exhibit physical characteristics which are different from other marine sediments. These properties include high water content, w (in the region of 200% up to 20 m depth), extremely high plasticity index, Ip (up to 150% at the seabed) and generally low values of undrained shear strength, su , Figure 1. In general, the undrained strengths of the clays are thought to be governed by their post-sedimentation history (Puech et al. 2005). The samples of OA clay selected for testing were recovered from depths of approximately 8.6 m and 16.2 m below seabed using a STACOR® piston corer (Borel et al. 2005). The clay samples are very soft with © 2011 by Taylor & Francis Group, LLC

Direct simple shear testing

The results of eleven consolidated, constant volume direct simple shear (DSS) tests performed on specimens from each STACOR® sample are presented. The DSS test is intended to simulate plane strain conditions in which radial deformation is prevented during consolidation (ensuring the K0 condition) but the specimen is allowed to deform in simple shear. The constant volume condition is achieved by continuous adjustment of the vertical stress acting on the horizontal surface during shearing, so that the height of the specimen remains constant. The change in vertical stress during shearing is considered to be equal to the change in pore-water pressure which would have occurred in a truly undrained test (Bjerrum & Landva 1966). DSS tests were carried out on 67 mm diameter, 19 mm high specimens, in the GDS Instruments apparatus. The system utilises a stack of low friction, Teflon coated steel rings placed over the membrane-enclosed specimen to provide the required lateral restraint. For each specimen the vertical consolidation stresses were representative of the average estimated in-situ conditions at the sample depth essentially following a normally to lightly overconsolidated stress history (OCR ≈ 1.2). Specimens were sheared at constant shear strain rates between 0.003 and 300%/hour. 3 3.1

EXPERIMENTAL RESULTS Constant rate of strain DSS

The results of all DSS tests undertaken are presented in Table 2. The maximum consolidation stress in each
test case is σvcmax . Figure 2 shows the variation of

378

Table 2.

Results of direct simple shear (DSS) tests. At peak shear stress

Test No.

Depth (m)


σvcmax (kPa)


σvc (kPa)

OCR

γ˙ (%/hr)

γ (%)

τ (kPa)

u (kPa)

τ/σ’vc

u/σ’vc

τ/σv


αp (◦)

SC1_1 SC1_2 SC1_3 SC1_4 SC1_5 SC1_6 SC2_1 SC2_2 SC2_3 SC2_4 SC2_5 SC2_6

8.62 8.65 8.74 8.78 8.70 8.80 16.17 16.23 16.29 16.33 16.36 16.33

26 26 26 26 26 26 50 50 50 50 50 50

22 22 22 22 22 22 40 40 40 40 40 40

1.2 1.2 1.2 1.2 1.2 1.2 1.25 1.25 1.25 1.25 1.25 1.25

0.003 0.03 0.3 3 30 300 0.03 0.3 3 30 300 0.8S L

4.4 6.0 6.0 6.6 8.1 6.8 6.3 4.0 6.1 7.3 6.7 –

5.1 5.3 6.4 7.6 8.3 10.3 10.7 12.0 13.6 16.5 18.9 10.0

5.7 8.1 5.8 5.2 4.7 1.3 9.8 6.8 7.6 6.7 3.7

0.23 0.24 0.29 0.34 0.38 0.47 0.27 0.30 0.34 0.41 0.47

0.26 0.37 0.26 0.24 0.21 0.06 0.25 0.17 0.19 0.17 0.09

0.32 0.38 0.40 0.45 0.48 0.50 0.35 0.36 0.42 0.50 0.52

17.5 21.0 21.7 24.3 25.7 26.5 19.5 19.9 22.7 26.4 27.4

Figure 2. Variation of (a) normalised shear stress (b) normalised excess pore pressure and (c) normalised shear stress (with respect to σv
) with shear strain. Sample SC2, depth 16.2 m.

shear stress, normalised with respect to both preshear
consolidation stress (σvc ) and effective vertical stress (σv
), and change in vertical stress required to maintain constant specimen volume. As noted previously the © 2011 by Taylor & Francis Group, LLC

379

latter corresponds to the excess pore pressure induced in an equivalent undrained shear test and is therefore referred to as u. The excess pore pressure has also
for dimensionless comparison. been normalised by σvc Data are presented in Figure 2 for all tests performed on sample SC2 at depths from 16.17 m. For all tests the point corresponding to the mobilisation of peak shear stress is indicated on each plot. A clear trend is observed for increasing shear stress with increasing strain rate (γ). ˙ This is illustrated in Figure 2(a), which also indicates that there is no clear strain rate effect on the shear strain at peak shear stress. It should be noted that this may be related to small differences in the internal structure of the individual specimens which, externally, appeared to have consistent composition. It is also worthwhile to note that, for each test undertaken, the shear strain at peak shear stress is less than that at which clay samples are generally thought to exhibit non-uniform straining due to the development of localised deformations or ruptures (Airey & Wood, 1987). In addition, the trend for slight strain softening shown in Figure 2(a) and approximately constant stress ratio with increased shear strain noted from Figure 2(c) would tend to indicate a tendency toward post-failure development of horizontal ruptures which is generally not consistent with the expected behaviour for high plasticity clay. Figure 2(b), shows a trend for overall reduction in the development of excess pore pressure as the applied strain rate increases. This trend is not well defined at γ < 1% and the limited number of results available also indicate that the variation of u may become strain rate independent at γ˙ < 0.03 %/hour. Reduction in excess pore pressure is thought to be the primary cause of enhanced shear strength and so this assumption is also supported by the similar peak shear stresses recorded for tests SC1_1 and SC1_2 sheared at γ˙ = 0.003 and 0.03 %/hour respectively. It should be noted that the strain rate conventionally used in the DSS test is selected so as to prevent any rise in pore pressure during shear under constant volume.

Figure 3. Normalised stress paths from constant volume DSS tests. Sample SC2, depth 16.2 m.

Figure 4. Compression behaviour from incremental loading (IL) and constant rate of strain (CRS) consolidation tests.

The highest strain rates applied in this study are up to 100 times faster than the reference rate and as such the excess pore pressure variations, which are inferred rather than directly measured, should be viewed with some caution. The normalised stress paths for tests undertaken on specimens from sample SC2 at depths from 16.17 m are presented in Figure 3. Owing to the test arrangement employed, the stress paths shown relate to the variation of shear stress and vertical stress on the horizontal plane only. Plane stain friction angles at peak stress (αp ) are dervied from Equation 2,

and are shown, in Table 2, to increase with applied strain rate for both samples tested. If the assumption were to be made that the horizontal plane is the plane of maximum obliquity at large strains, then the friction angle relating to this state may also be estimated. However, as the extent to which localised deformation may have occurred within each specimen at large shear strain is largely unknown, these angles are not reported here. For clarity the results of sample SC1 are not shown in Figures 2 and 3, although the trends discussed are consistent for each data set. 3.2

Compression behaviour

In Figure 4, IL and CRS one-dimensional compression curves are compared for specimens recovered from depths between 15.5 and 15.7 m at a location adjacent to that of the STACOR® samples. The curves are indicative of the high compressibility of the OA clay but also highlight the influence of strain rate on the stress-compressibility relationship. The CRS curve shown in Figure 4 is principally representative of the primary consolidation curve, © 2011 by Taylor & Francis Group, LLC

380

Figure 5. Consolidation behaviour under a constant incremental stress greater than pc ’ (oedometer test, depth 15.5 m).

whereas the incremental loading curve will incorporate both primary and secondary compression strains, the latter being creep under constant stress. The vertical strain, εv , log time relationship is shown for a specimen under constant stresses equal to, and in excess of, the estimated preconsolidation stress in Figure 5. The coefficient of secondary consolidation with respect to strain, Cαε , is used as a reference for describing creep behaviour in clay and has been shown to vary with applied effective stress, reaching an approximately constant value beyond the preconsolidation stress pc ’ (Mesri & Godlewski 1977). The results indicate that the high level of compressibility in primary consolidation is also reflected in the secondary consolidation behaviour. In agreement with the findings

4.2

Empirical relationship

Leroueil et al. (1985) proposed a model which relates the viscous properties of a soil to its primary deformation behaviour:

where ε˙ 0 is a reference strain rate and m is a constant describing the rate dependence of the soil. The parameter m has been shown to be equal to the ratio Cαε /Ccε for a range of different clays in one dimensional compression. The CRS data presented in Figure 4 have been modified using Equation 4 and the corrected result is also plotted in Figure 4. Since the strain rate during compression is not constant, the reference strain rate in this case has been taken to be equal to the value derived at the end of primary consolidation in the incremental test. Good agreement is observed between the IL and corrected CRS results supporting the validity of Equation 4 in describing the rate dependence in compression of the OA clay. If it is considered that there is a common mechanism controlling creep and rate dependency then the relationship derived to describe one phenomenon may be equally applied to the other. Equation 4 may then be re-written to relate to the behaviour of the OA clay during constant volume shear as follows,

Figure 6. Variation of normalised shear strength with shear strain rate for DSS tests on OA clay specimens.

of Mesri & Castro (1987) the ratio of Cαε to the compression index, Ccε , is observed to be approximately constant for the OA soils tested. At the depths of interest for this study the values of Cαε / Ccε lie within the range 0.06–0.065.

4 ANALYSIS OF RESULTS 4.1

Shear strength

The variation of normalised constant volume shear strength (suDSS /su0 ) with shear strain rate is shown in Figure 6 for both samples tested. A clear trend for increased shear strength with shear strain rate is evident from the results. The overall magnitude of the increase may be interpreted through use of the strain rate parameter proposed by Sheahan et al. (1996), given in Equation 1, modified to apply to the DSS test as shown in Equation 3.

where su0 is the shear strength taken at a strain rate of 3%/hour, in line with current practice for direct simple shear testing. Values of ργ0 ˙ calculated from the results presented exhibit a trend for increasing with shear strain rate up to a maximum of approximately 19% per log cycle. This rate dependence is at the upper limit of that derived for other soil types from previously published experimental studies. This may be related to the thickness of the absorbed water envelope of the OA clays and a viscous mechanism which is enhanced (or less suppressed) as strain rate increases. © 2011 by Taylor & Francis Group, LLC

where it is assumed that the observed stress-strainstrain rate relation is consistent for viscous behaviour under creep and during constant volume deformation and may be described empirically. The relevance of this assumption in describing the measured rate dependence is demonstrated in Figure 6 where Equation 5 is plotted alongside the results of the DSS tests undertaken. The parameter m has been taken to be equal to the average value of the ratio Cαε / Ccε , established previously for the OA clays of interest as 0.063. This value represents an approximate 16% increase in shear strength per log cycle increase in shear strain rate. 4.3

Implications for foundation designs

The influence of rate of loading on foundation performance is not always accounted for in design. DNV (2002) presents one of the most comprehensive codifications of rate effects on plate anchors, based on a series of anchor pull out tests at Onsøy. Despite those specific criteria, the recommendations provide a framework for assessment of rate effects for other foundation types and in different soils. Clearly, to properly account for the variations in mobilised shear strength requires understanding of the loading rate in the field as well as the loading rate

381

and promote the applicability of a unique stress-strainstrain rate relation for the OA clay which is likely related to the ratio of its mineralogical components. This relationship is only valid under conditions where strains are increasing and as such cannot be used to model relaxation behaviour. Design factors describing reduction in shear strength under slow loading are often based on assumed “typical” limiting values and not on site specific tests which, as this paper demonstrates, can be informative. Advantages may also be gained from the increase in available soil strength for rapid loading, such as short-term environmental events (for example the “squalls” experienced offshore Angola). Provided tests are performed in a properly controlled manner, it should be possible to optimise foundation designs in this way, according to the measured strain rate dependent relation for the soil. Figure 7. Variation of normalised shear strength with time to failure for DSS tests on OA clay specimens.

specified in the laboratory. Rather than expressing the loading rate as a strain rate, it may be simpler to consider time to failure (Tf ) which can be compared to the loading duration of the design event. There is a limit to the applicability of the results of these tests. Eventually, at slow loading rates, drainage effects will start to influence the strength of the soil. However, within the range of loading durations shown in Figure 7 (i.e. up to one or two weeks), it is suggested that an undrained assumption is valid. It is becoming more routine in design to take account of the impact of slow loading rates on shear strength to account for creep effects. Loading of subsea manifold foundations and sustained tension on moorings or riser bases in deepwater may require shear strength reduction for a reliable foundation design. Typically this ranges from 60–80% of the reference strength. To establish these factors sustained load testing may be undertaken and the results of a number of such tests have been included in Figure 7 to highlight the apparent agreement between strain rate and creep related time to failure behaviour. 5

CONCLUSIONS

The results presented in this paper have demonstrated that higher undrained strengths are measured at higher strain rates for the OA clays under DSS conditions. These increases are primarily a result of a reduction in the generation of excess pore pressures during shearing although may also be related to some increases in the effective stress friction angle mobilised at peak shear stress. The mechanisms controlling this rate dependency appear to be consistent for both strain rate and strain creep behaviour, as described during 1-D compression,

© 2011 by Taylor & Francis Group, LLC

REFERENCES Airey, D.W. & Wood, D.M. 1987. An evaluation of direct simple shear tests on clay. Geotechnique 37(1): 25–35. Bjerrum, L. & Landva, A. 1966. Direct simple shear tests on a Norwegian quick clay. Geotechnique 16(1): 1–20. Bjerrum, L. 1973. Problems of soil mechanics and construction on soft clays. Proc. 8th Int. Conf. Soil Mech. Moscow 3: 111–159. Borel, D., Puech, A. Dendani, H. & Colliat, J-L. 2005. Deepwater geotechnical site investigation practice in the Gulf og Guinea. Proc. Int. Symp. on Frontiers in Offshore Geotech. Perth: 921–926. Briaud, J-L. & Garland, E. 1985. Loading rate method for pile response in clay. J. Geotech. Eng. 111(3): 319–335. Det Norsk Veritas. 2002. Design and installation of plate anchors in clay. DNV recommended pactice RP-E302. Hovik. Graham, J., Crooks, J.H.A. & Bell, A.L. 1983. Time effects on the stress-strain behaviour of natural soft clays. Geotechnique 33(3): 327–340. Lefebvre, G. & LeBoeuf, D. 1987. Rate effects and cyclic loading of sensitive clays. J. Geotech. Eng. 113(5): 476–489. Leroueil, S., Kabbaj, M., Tavenas, F. & Bouchard, R. 1985. Stress-strain-strain rate relation for the compressibility of sensitive natural clays. Geotechnique 35(2): 159–180. Low, P.F. 1961. The physical chemistry of clay-water interaction. Advances in Agronomy 13(1): 269–327. Mesri, G. & Godlewski, P.M. 1977. Time and stresscompressibility inter-relationship. J. Geotech. Eng. 103(5): 417–430. Mesri, G. & Castro, A. 1987. Cα /Cc concept and K0 during secondary compression. J. Geotech. Eng. 113(3): 230–247. Puech, A., Colliat, J-L., Nauroy, J-F., & Meunier, J. 2005. Some geotechnical specifities of Gulf og Guinea deepwater sediments. Proc. Int. Symp. on Frontiers in Offshore Geotech. Perth: 1047–1053. Sheahan, T.C., Ladd, C.C. & Germaine, J.T. 1996. Rate dependant undrained shear behaviour of saturated clay. J. Geotech. Eng. 122(2): 99–108.

382

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Simplified calibration procedure for a high-cycle accumulation model based on cyclic triaxial tests on 22 sands T. Wichtmann, A. Niemunis & Th. Triantafyllidis Institute of Soil Mechanics and Rock Mechanics, Karlsruhe Institute of Technology

ABSTRACT: A simplified calibration procedure for the material constants used in the authors’ high-cycle accumulation model has been improved based on data from more than 350 drained cyclic triaxial tests performed on 22 clean quartz sands with different grain size distribution curves. The simplified method allows the estimation of a set of parameters from characteristics of the grain size distribution curve (mean grain size, coefficient of uniformity) and index quantities (minimum and maximum void ratio). 1

INTRODUCTION

The high-cycle accumulation (HCA) model proposed by Niemunis et al. (2005) predicts the accumulation of permanent deformations or the build-up of excess pore water pressure due to a cyclic loading with many cycles (N > 103 ) of small to intermediate strain amplitudes (εampl < 10−3 ). The model can be used for example for the prediction of permanent deformations of offshore wind power plant (OWPP) foundations (Wichtmann et al., 2010b). The determination of the material constants of the HCA model (Wichtmann et al., 2010a) is quite laborious. Drained cyclic triaxial tests with different amplitudes, densities and average stresses are necessary. Regarding the large number of OWPPs in a wind park and the layered soil, an experimental determination of the constants for each OWPP foundation and each soil type would be tedious. Therefore, a simplified calibration procedure has been already proposed by Wichtmann et al. (2009) based on cyclic triaxial tests on eight quartz sands with different grain size distribution curves. Correlations of the HCA model constants with index properties (mean grain size d50 , coefficient of uniformity Cu , minimum void ratio emin ) have been developed for that purpose. However, some of the correlations showed a significant amount of scatter. Therefore, 14 more grain size distribution curves with linear shape (in the semi-logarithmic scale) and with different mean grain sizes and coefficients of uniformity were tested in order to improve the correlations and to adapt them to a wider range of d50 - and Cu -values. The present paper reports on this effort. 2 TESTED MATERIALS AND TESTING PROCEDURES The 14 tested grain size distribution curves are shown in Figure 1. They were mixed from a natural quartz © 2011 by Taylor & Francis Group, LLC

Figure 1. Tested grain size distribution curves.

sand with subangular grain shape. The sands and gravels L1 to L7 (Figure 1a) have mean grain sizes in the range 0.1 mm ≤ d50 ≤ 3.5 mm and the same coefficient of uniformity Cu = d60 /d10 = 1.5. The materials L4 and L10 to L16 (Figure 1b) have the same mean grain size d50 = 0.6 mm while Cu varies between 1.5 and 8. The samples with a diameter of 10 cm and a height of 20 cm were prepared by dry air pluviation and afterwards saturated with de-aired water. They were consolidated for one hour at the average stress. Due to large

383

Figure 2. Accumulation curves εacc (N ) in tests with different stress amplitudes qampl (all tests: pav = 200 kPa, ηav = 0.75), thick solid curves = recalculation with HCA model.

Figure 3. Accumulated strain εacc /f¯e as a function of strain amplitude ε¯ ampl (all tests: pav = 200 kPa, ηav = 0.75).

deformations the first irregular cycle was applied with a low loading frequency of 0.01 Hz while f = 1 Hz was used for the subsequent 100,000 regular cycles. The only exception was the fine sand L1 were 2,000 or 10,000 regular cycles were tested with frequencies of 0.01 or 0.1 Hz, respectively. For each material several tests with different amplitudes, initial densities, average mean pressures pav and average stress ratios ηav = qav /pav were performed (p = (σ1 + 2σ3 )/3, q = σ1 − σ3 ). Since the HCA model predicts the accumulation due to the regular cycles only, the irregular cycle is not discussed in the following. The direction of accumulation m = ε˙ acc /||˙εacc || (flow rule) used in the HCA model could be confirmed for all tested sands and is also not further addressed here. 3 TEST RESULTS AND DETERMINATION OF HCA MODEL CONSTANTS The increase of the intensity of
accumulation ε˙ acc = ||˙εacc || = ∂εacc /∂N (with ε = ε21 + 2ε23 ) with increasing stress or strain amplitude becomes clear from Figures 2 and 3. Figure 2 shows the accumulated permanent strain εacc as a function of the number of cycles N in the tests with different deviatoric stress amplitudes qampl . In Figure 3 the permanent strain after different numbers of cycles is plotted versus the strain amplitude. Since in the stress-controlled tests the strain amplitude  decreased slightly with N , a mean value ε¯ ampl = 1/N εampl (N )dN over N is used on the abscissa. On the ordinate the data is divided by the void ratio function of the HCA model in order to purify it from the influence of void ratio:

© 2011 by Taylor & Francis Group, LLC

with the maximum void ratio emax and the material constant Ce . The bar over f¯e in Figure 3 denotes that the void ratio function has been calculated with a mean  value e¯ = 1/N e(N )dN of void ratio. An overproportional increase of the intensity of strain accumulation with the strain amplitude can be concluded from Figure 3. The amplitude function

of the HCA model has been fitted to the data shown in Figure 3 (solid curves) delivering Campl . The Campl values given in column 3 of Table 1 are mean values over 100,000 cycles. The dependence of ε˙ acc on the grain size distribution curve is inspected in Figure 4, where the residual strain after 10,000 cycles is plotted versus d50 or Cu , respectively. In accordance withWichtmann et al. (2009) the intensity of accumulation increases with decreasing mean grain size and with increasing coefficient of uniformity. Figures 5 and 6 demonstrate the increase of the rate of strain accumulation with increasing void ratio. While Figure 5 compares the curves εacc (N ) for different initial relative densities ID0 , Figure 6 presents the residual strain after different N -values as a function of void ratio e¯ . In order to purify the data from the influence of slightly different strain amplitudes, εacc has been divided by the amplitude function of the HCA model. The bar over f¯ampl in Figure 6 denotes that the amplitude function has been calculated with a mean value ε¯ ampl of the strain amplitude. The parameter Ce (column 4 ofTable 1) was obtained from a curve-fitting of the function fe to the data in Figure 6. Since fampl is necessary to purify the data in Figure 6 and fe is used on the ordinate in Figure 3, the determination of Campl and Ce has to be done by iteration.

384

Table 1.

HCA model parameters for the 14 tested sands. C++ program

“by hand” method

Sand

emin [–]

emax [–]

Campl [–]

Ce [–]

Cp [–]

CY [–]

CN 1 [10−4 ]

CN 2 [–]

CN 3 [10−5 ]

Campl [–]

Ce [–]

Cp [–]

CY [–]

CN 1 [10−4 ]

CN 2 [–]

CN 3 [10−5 ]

L1 L2 L3 L4 L5 L6 L7 L10 L11 L12 L13 L14 L15 L16

0.634 0.596 0.591 0.571 0.580 0.591 0.626 0.541 0.495 0.474 0.414 0.394 0.387 0.356

1.127 0.994 0.931 0.891 0.879 0.877 0.817 0.864 0.856 0.829 0.791 0.749 0.719 0.673

1.60 1.43 1.76 1.92 1.77 1.70 1.46 1.53 2.03 1.40 1.68 2.06 1.76 1.36

0.60 0.64 0.59 0.55 0.52 0.56 0.51 0.53 0.50 0.47 0.40 0.32 0.33 0.31

0.40 0.29 0.69 0.53 0.29 0.12 0.11 0.36 0.42 0.39 0.39 0.66 0.55 0.23

1.84 1.94 2.72 2.52 2.77 2.57 3.49 2.21 2.41 2.70 2.44 2.67 2.15 1.99

5.61 16.8 10.5 5.07 2.77 3.01 1.41 19.3 23.3 51.4 53.6 46.6 68.6 107

0.328 0.137 0.185 0.197 0.303 0.576 0.907 0.0439 0.0257 0.0131 0.00969 0.00817 0.00732 0.00611

8.79 5.37 2.02 2.76 1.86 0 0 5.74 8.18 7.74 6.85 5.70 6.67 8.78

1.69 1.33 1.85 1.97 1.84 1.64 1.48 1.67 2.43 1.60 1.85 2.34 1.97 1.53

0.60 0.65 0.61 0.57 0.54 0.58 0.51 0.53 0.53 0.48 0.40 0.34 0.34 0.31

0.40 0.30 0.55 0.52 0.32 0.11 0.09 0.32 0.50 0.44 0.34 0.45 0.44 0.23

1.99 1.89 3.00 2.82 3.14 2.72 3.49 2.37 2.89 3.02 3.12 3.29 2.69 2.45

0.485 18.0 8.25 4.35 2.50 3.66 1.28 13.4 15.4 36.0 26.6 23.0 41.2 79.2

0.30 0.15 0.24 0.30 0.54 0.89 0.96 0.075 0.040 0.016 0.0090 0.0065 0.0070 0.0050

10.5 6.0 2.1 3.5 2.0 0.1 0 5.5 13.5 10.5 10.0 7.5 7.5 8.0

function of the normalized average stress ratio Y¯ av , where Y¯ and η are interrelated via

Figure 4. Accumulated strain εacc as a function of a) mean grain size d50 and b) coefficient of uniformity Cu .

with critical friction angle ϕc . Y¯ av is zero for isotropic stress conditions and equal to one on the critical state line. The HCA model parameter CY (column 6 of Table 1) was obtained from a curve-fitting of the function fY to the data in Figure 10:

The accumulation curves εacc (N ) in the tests with different average mean pressures pav and with a constant average stress ratio (here ηav = 0.75) coincide approximately if the tests are performed with the same amplitude-pressure ratio ζ = qampl /pav (Figure 7). The increase of the strain amplitude with increasing pressure for ζ = constant has been considered in Figure 8 where the residual strain has been divided by f¯ampl and f¯e and plotted versus pav . The decrease of the intensity of accumulation with increasing average mean pressure is obvious in Figure 8. It becomes less pronounced with increasing mean grain size. The data for some sands (e.g. L15, Figure 8) indicate almost constant accumulation rates for larger pressures. Tests with pav > 300 kPa are planned for the future. The HCA model parameter Cp (column 5 of Table 1) was obtained from a curve-fitting of the function fp to the data in Figure 8:

The shape of the curves εacc (N ) can be judged from Figure 11 where the residual strain has been divided by the functions f¯ampl , f¯e , fp and fY of the HCA model (calculated with the constants given in columns 3 to 9 of Table 1), that means it was purified from the influences of amplitude, void ratio and average stress. For uniform sands the residual strain increases almost proportional to ln(N ) up to at least N = 104 . At large numbers of cycles N > 104 , the residual strain grew faster than proportional to ln(N ) for some of the sands L1 to L7 (see e.g. L2 in Figure 11). The curves εacc (N ) for the more well-graded sands show a bending in the semi-logarithmic scale, which becomes more pronounced with increasing coefficient of uniformity of the tested material (see L11 and L15 in Figure 11). These findings agree well with the results of the earlier study documented by Wichtmann et al. (2009). The parameters CN 1 , CN 2 and CN 3 (columns 7 to 9 of Table 1) were received by fitting the data in Figure 11 with the function fN (solid curve):

For all tested materials the increase of the strain accumulation rate with increasing average stress ratio was confirmed. Figures 9 and 10 compare the accumulation curves εacc (N ) or show the residual strain as a

The loading frequency does not influence the rate of strain accumulation in non-cohesive soils (see the literature review given by Wichtmann et al. (2009)) and is thus not considered in the HCA model.

© 2011 by Taylor & Francis Group, LLC

385

Figure 5. Accumulation curves εacc (N ) in tests with different initial relative densities ID0 (all tests: pav = 200 kPa, ηav = 0.75), thick solid curves = recalculation with HCA model.

Figure 6. Accumulated strain εacc /f¯ampl as a function of void ratio e¯ (all tests: pav = 200 kPa, ηav = 0.75).

Figure 7. Accumulation curves εacc (N ) in tests with different average mean pressures pav (all tests: ηav = 0.75, ζ = qampl /pav ), thick solid curves = recalculation with HCA model.

Figure 8. Accumulated strain εacc /(f¯ampl f¯e ) as a function of average mean pressure pav (all tests: ηav = 0.75, ζ = qampl /pav ).

Figure 9. Accumulation curves εacc (N ) in tests with different average stress ratios Y¯ av (all tests: pav = 200 kPa), thick solid curves = recalculation with HCA model.

As an alternative to the “by hand” calibration outlined above, the HCA model parameters were also determined by means of a C++ program. It finds those parameters for which the sum of the squares of the

© 2011 by Taylor & Francis Group, LLC

386

differences between the experimentally obtained εacc data and the data predicted by the HCA model takes its minimum. The method may be seen as some kind of “fine tuning” of the parameters. The parameters

Figure 10. Accumulated strain εacc /(f¯ampl f¯e ) as a function of normalized average stress ratio Y¯ av (all tests: pav = 200 kPa).

Figure 11. Curves εacc (N )/(f¯ampl f¯e fp fY ), fitting of function fN .

summarized in columns 10 to 16 of Table 1 differ from those calibrated “by hand” due to simplifications of the “by hand” method (for example mean values ε¯ ampl and e¯ are used in the diagrams, parameters determined for different N -values are averaged). 4

RE-CALCULATION OF ELEMENT TESTS

The parameters given in columns 10 to 16 of Table 1 were used for recalculations of the element tests with the HCA model. The predicted curves have been added as solid lines in Figures 2, 5, 7 and 9. The parameters determined “by hand” (columns 3 to 9 of Table 1) deliver quite similar curves. In most cases the deviation between the experimental and the calculated data is small, confirming the good prediction of the HCA model. For some sands slightly too low accumulation rates are predicted for small pressures (Figure 7) which is due to deficits of the function fp . This will be inspected in more detail in future. 5

Similarly, the data for CN 1 , CN 2 and CN 3 in Figure 12j-o have been analyzed with Campl , Ce , Cp and CY calculated from Equations (7) to (10). Beside the calibration methods discussed in Section 3, the parameters Campl , Ce , Cp and CY were also estimated from the rate data (see Wichtmann et al., 2010a). CN 1 , CN 2 and CN 3 were determined both, from the data of all curves εacc (N ) and from the curves of the three tests with different amplitudes only. The poor correlation between CN 3 and d50 can possibly be improved by means of data from tests with larger numbers of cycles (N > 105 ).

SIMPLIFIED CALIBRATION PROCEDURE

In Figure 12 the HCA model parameters are plotted versus mean grain size d50 , coefficient of uniformity Cu or minimum void ratio emin , respectively. The data from the tests described by Wichtmann et al. (2009) were re-analyzed with Campl = 2.0 and are included in Figure 12. The correlations defined by Equations (7) to (13) are given in Figure 12 as solid lines and may be used for a simplified estimation of a set of parameters. The parameter Campl does not correlate with d50 or Cu (Figure 12a,b). For Ce both, a correlation with d50 and Cu (Figure 12c,d) and with minimum void ratio emin (Figure 12e) could be established. The values of Cp and CY plotted in Figure 12f–i were obtained calculating Campl and Ce from Equations (7) and (8). © 2011 by Taylor & Francis Group, LLC

387

6

SUMMARY, CONCLUSIONS AND OUTLOOK

Based on the data from approx. 350 drained cyclic triaxial tests performed on 22 quartz sands with different grain size distribution curves a simplified procedure for the determination of the parameters of the authors’ high-cycle accumulation (HCA) model has been developed. Correlations of the HCA model

Figure 12. Correlations of the HCA model parameters with d50 , Cu or emin , respectively (SF = Wichtmann et al., 2009).

parameters with mean grain size d50 , coefficient of uniformity Cu or minimum void ratio emin , respectively, have been formulated. In future the simplified calibration procedure will be extended to granular materials with fines content.

ACKNOWLEDGEMENTS The study presented in the paper has been performed in the framework of the project A8 of SFB 398 “Lifetime oriented design concepts” during the former work of the authors at Ruhr-University Bochum, Germany. The authors are grateful to DFG (German Research Council) for the financial support and to M. Skubisch who carefully performed the cyclic triaxial tests.

© 2011 by Taylor & Francis Group, LLC

REFERENCES Niemunis, A., Wichtmann, T. & Triantafyllidis, T. 2005. A high-cycle accumulation model for sand. Computers and Geotechnics, 32(4):245–263. Wichtmann, T., Niemunis, A. & Triantafyllidis, T. 2009. Validation and calibration of a high-cycle accumulation model based on cyclic triaxial tests on eight sands. Soils and Foundations, 49(5): 711–728. Wichtmann, T., Niemunis, A. & Triantafyllidis, T. 2010a. On the determination of a set of material constants for a highcycle accumulation model for non-cohesive soils. Int. J. Numer. Anal. Meth. Geomech., 34(4):409–440. Wichtmann, T., Niemunis, A. & Triantafyllidis, T. 2010b. Towards the FE prediction of permanent deformations of offshore wind power plant foundations using a highcycle accumulation model. In International Symposium: Frontiers in Offshore Geotechnics, Perth, Australia.

388

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Understanding cyclic loading behavior of soil for offshore applications J. Yang Department of Civil Engineering, The University of Hong Kong, Hong Kong, China

ABSTRACT: This paper discusses two categories of cyclic loading with special reference to offshore applications: one is the combination of static and cyclic shear stresses and the other is a continuous rotation of the principal stress axes. Particular attention is given to the fist type which is in close relation to the stability of gravity offshore structures. It is shown that the initial static shear stress, due to the installation of an offshore structure, is a key factor for the behavior of the sand beneath the structure. The impact of this factor strongly depends on the initial state of the sand in terms of its relative density and confining stress and on the magnitude of the static shear stress relative to the cyclic shear stress applied. The second type of loading is in relation to the seabed in the free field under ocean waves. It is shown for this loading condition that rotation of principal stress axes without changing the magnitude of deviatoric stress can generate pore water pressures and that the amount and rate of pore pressures is affected by the intermediate principal stress parameter.

1

INTRODUCTION

Instability of the seabed under the action of storms or earthquakes is an important consideration in the design and installation of offshore structures (breakwaters, gravity or pile-supported platforms, pipelines, anchors, etc.), since it may cause severe damage to these structures affecting their operations (Oumeraci 1994; Palmer et al. 2004). Figure 1 illustrates possible failure modes of breakwaters and pipelines in relation to the instability of the seabed. During storms that are of major concern in offshore engineering, cyclic stresses develop in the seabed soil. The stresses can result in a progressive build-up of pore water pressure in the soil and a loss of strength of the soil. Particularly, if the seabed is comprised primarily of sand, the pore water pressure within the sand may build up to the level of the effective vertical pressure resulting in liquefaction and possible instability. The amount and rate of the pore water pressure build-up depends on the properties of the seabed soil including the particle gradation, permeability and relative density; it also depends on the characteristics of the storms such as the height, length, and period of wave components. The installation of an offshore structure into the seabed presents additional considerable difficulty to the problem, because very complex interactions are involved between the structure, the seabed soil and the waves. Note that the characteristics of the structure (e.g., the geometry, weight and stiffness) can play an important role in such interactions. From a geotechnical consideration, the presence of the offshore structure will induce both normal and shear stresses on the soil elements beneath the structure, which are then superimposed by “indirect cyclic loading” due © 2011 by Taylor & Francis Group, LLC

389

Figure 1. Schematic illustration of failure modes of offshore structures: (a) breakwaters; (b) pipelines.

to rocking and translation of the structure. This loading condition is considered critical in the design of offshore structures, particularly for the gravity-type structures; it can be characterized by a sustained, static shear stress on the horizontal plane of a soil element throughout the course of cyclic loading. The term “indirect cyclic loading” is used here as opposed to the cyclic loading directly imposed by ocean waves on the seabed in the free field (i.e. without the presence of structures). In this case, the wave loading can be characterized by a continuous rotation

of the principal stress axes in the seabed soil (Madsen 1978; Ishihara & Towhata 1983). Following the pioneering work of Seed & Lee (1966), the cyclic loading behavior of soils has been extensively studied in the past decades with special reference to soil liquefaction during earthquakes. Most of the experimental studies have been focused on the cases where no static shear stresses exit. As reviewed by Yang & Sze (2010), the effect of the static shear stress is not yet well understood due to the lack of physical data and the inconsistency in the existing interpretations. With regard to the stability problems of offshore structures, a proper understanding of the soil behavior under this loading condition is important in developing better technical solutions in terms of safety, economy and reliability. This paper aims to introduce several new findings on the behavior of sand under the combination of static and cyclic shear stresses, established from a comprehensive experimental program (Yang et al. 2009; Yang & Sze 2010), to the offshore engineering community. In addition, recent advances in the study of the effect of rotation of principal stress axes on sand behavior (Yang et al. 2007) are also briefly discussed. 2 2.1

IMPACT OF STATIC SHEAR STRESS ON CYCLIC BEHAVIOR OF SAND

Figure 2. Cyclic triaxial loading conditions: (a) without initial static shear stress; (b) with initial static shear stress.

Simulation of static shear stresses

The cyclic behavior and liquefaction resistance of sand has been commonly studied by undrained cyclic triaxial tests with two-way, symmetrical loading. In such tests a sand specimen is first consolidated under the hydrostatic condition and then subjected to an undrained cyclic deviatoric stress, qcyc , which alters between positive and negative values of the same magnitude (Fig. 2(a)). This symmetrical loading aims to simulate the free-field level ground during earthquakes. To model the static shear stress in the cyclic triaxial test, the soil specimen is first consolidated
under the major and minor principal stresses, σ1c and
, to produce a static shear stress, τs , on the 45◦ plane σ3c in the specimen (Fig. 2(b)); a cyclic deviatoric stress is then superimposed under undrained conditions to produce a cyclic loading that is non-symmetrical about the hydrostatic stress state. The level of static shear stress can be measured by a parameter α defined below:


where σnc is the effective normal stress on the 45◦ plane. Obviously, with increasing α values the level of static shear increases, and the special case of α = 0 represents the symmetrical loading condition. Using the above method, a comprehensive experimental program consisting of more than 150 cyclic triaxial tests has been conducted on two standard sands (Toyoura sand and Fujian sand) to investigate the

© 2011 by Taylor & Francis Group, LLC

impact of static shear stress (Yang et al. 2009; Yang & Sze 2010). The test program covered a wide range of initial states in terms of the relative density, confining stress and the initial static shear stress. Details on the test equipment, sample preparation and test procedure can be found in Yang & Sze (2010). Three distinct failure patterns have been identified from the test program, namely flow type failure or strain softening, cyclic mobility and plastic strain accumulation. 2.2

Flow failure

Figure 3 shows the undrained cyclic response of a loose Toyoura sand specimen at the relative density of 10% and the confining stress level of 100 kPa. The initial static shear stress was applied at 40 kPa and the subsequent cyclic deviatoric stress was 7 kPa. Given this combination, no stress reversals took place during the entire process of cyclic loading. The excess pore water pressure (PWP) built up progressively until the 28th loading cycle where an abrupt rise of the PWP occurred (Fig. 3(a)). Correspondingly, a sudden, run-away deformation was observed (Fig. 3(b)). Since there was no stress reversal, cyclic loading sat entirely on the compression side and the run-away deformation took place in compression. This failure mode has been observed on all loose specimens (Drc = 10% and 20%), and is characterized by a sudden run-away deformation being similar to that observed in loose sand under monotonic loading conditions (Yang 2002). The occurrence of this failure

390

Figure 3. Typical response of loose sand (Drc = 10%,
σnc = 100 kPa, τs = 40 kPa, qcyc = 7 kPa).

Figure 4. Typical response of medium dense sand
(Drc = 50%, σnc = 100 kPa, τs = 40 kPa, qcyc = 110 kPa).

mode is regardless of stress reversal and is the result of the contractive nature of loose sand. 2.4 2.3 Cyclic mobility Shown in Figure 4 is a typical response referred to as cyclic mobility, which is similar in several aspects to that commonly observed in medium dense to dense sand subjected to cyclic loading without static shear stresses (Castro 1975). In this test, the sand specimen was consolidated at the relative density of 50% and the confining stress of 100 kPa. The static shear stress was at 40 kPa and the cyclic deviatoric stress was applied at 110 kPa, leading to stress reversals in the loading process. Note that the excess pore pressure build-up showed a different pattern with that observed in loose sand (see Fig. 4(a) and Fig. 3(a)) due to the cyclic contraction and dilation of the sand. The pore pressure kept building up but it was unable to reach the level of
. At about the 12th cycle, effective confining stress σnc transient softening and then strengthening occurred, which repeated themselves in the subsequent cycles (Fig. 4(b)). This response was more substantial in compression because stress reversals dominated on the compression side. The regain of strength and stiffness was the consequence of dilation in the sample. © 2011 by Taylor & Francis Group, LLC

391

Plastic strain accumulation

Figure 5 shows the typical response called plastic strain accumulation. In this test the Toyoura sand specimen was consolidated at the relative density of 50% and a much higher confining stress, 500 kPa. The static shear stress was as high as 200 kPa and the cyclic deviatoric stress was 440 kPa, leading to almost zero stress reversals in this case. Although the sand specimen was also in a medium dense state, it did not show the features of cyclic mobility. The axial strain accumulated continuously on the compression side (Fig. 5(b)) and, particularly, apart from a pronounced strain in the 1st loading cycle, the strain accumulation in the subsequent loading cycles was at more or less a constant rate. It is worth noting that the excess PWP at the end of the test was only about 30% of the initial effective confining
stress σnc . The modes of plastic strain accumulation and cyclic mobility have been observed in all medium dense and dense samples (Drc = 50% and 70%). It has been found that the criterion for the occurrence of cyclic mobility is that the cyclic loading should be with stress reversals. If there is no stress reversal, then plastic strain accumulation will be dominant.

Figure 6. Definitions for failure: (a) flow type failure; (b) cyclic mobility; (c) plastic strain accumulation. Figure 5. Typical response of medium dense sand
(Drc = 50%, σnc = 500 kPa, τs = 400 kPa, qcyc = 440 kPa).

2.5

Characterization of cyclic shear strength

To characterize the cyclic shear strength, the failure criterion needs to be rationally defined for each type of response. For the flow type response given in Figure 3, the onset of failure can uniquely be defined as the triggering of run-away deformations, as clearly shown in Figure 6(a). For the cyclic mobility mode and the plastic strain accumulation mode, however, there is no such definite point indicating the initiation of failure or liquefaction. For the former one, the usual way for the symmetrical loading condition is adopted to define the failure as the point where 5% double amplitude (D.A.) axial strains occur (Fig. 6(b)). For the latter, since the strain development only takes place in one direction, the occurrence of 5% peak axial strain (P.S.) in compression (Fig. 6(c)) becomes as reasonable a criterion as the 5% D.A. is for cyclic mobility. Using the failure criteria defined above, the cyclic stress ratio (CRRn ) required to cause the sample to failure at a specified number of cycles (e.g., 10, 15 or 50) can be determined as

As an example, Figure 7 presents the cyclic strength specified at 10 loading cycles for Toyoura sand at © 2011 by Taylor & Francis Group, LLC

Figure 7. Impact of static shear stress on cyclic shear strength of Toyoura sand.

various initial states. The cyclic strength is found to be highly dependent on the three initial state parameters – relative density (Drc ), effective confining
stress (σnc ) and initial static shear stress (α). Generally, CRRn always increases with increasing Drc and/or
, regardless of the level of initial shear decreasing σnc stress. It means that more dilative sand bears higher cyclic strength. One thing worth noting is the impact of the static shear stress. The impact depends on both relative density and confining stress level. For loose samples tested, the impact is found to be beneficial to cyclic

392

shear strength if the level of static shear stress (in terms of α) is low, but it tends to be detrimental when α becomes higher. In this regard,Yang & Sze (2010) have proposed a new concept called threshold α and shown that the threshold α can be estimated by using the noreversal line in the CRRn – α plane (see Fig. 7). This no-reversal line divides between loading with stress reversals (i.e. qcyc /2 > τs ) and without stress reversals (i.e. qcyc /2 < τs ). It is interesting to note that, whenever the CRRn trend line touches the no-reversal line, the CRRn is about to drop with α. In other words, when α is increased to a level such that loading required to bring the sand to failure is without stress reversal, then the cyclic resistance starts to reduce. More importantly, Yang & Sze (2010) have extended the concept to sand at high relative density within the framework of critical state soil mechanics, by introducing a state parameter that collectively accounts for relative density and confining stress (Been & Jefferies 1985). In this framework, the cyclic strength of sand having high relative density may not always increase with increasing α values; rather, it may reduce with α at larger α values when the confining stress is sufficiently high. More details can be found in Yang & Sze (2010). 3

Figure 8. State of stress in seabed due to passage of waves.

IMPACT OF ROTATION OF PRINCIPAL STRESS AXES

During the passage of waves the cyclic loading imposed on the seabed in the free field can be characterized by a continuous rotation of principal stress axes. In this case, the normal and shear stresses induced by the wave loading form a circular path in the plane of τvh and (σv − σh )/2, while the deviatoric stress remains unchanged and the direction of the principal stress (β) continuously rotates from 0◦ to 180◦ (Fig. 8). By assuming that the load induced by the propagation of waves on the surface of the seabed is in a harmonic form, the normal and shear stress components in the seabed can be determined as (Madsen 1978):

where x and z are the horizontal and vertical coordinates, respectively; σ0 is the amplitude of the harmonic wave load, L is the wave length and T is the period of waves. The undrained behavior of sand under this loading condition was investigated by Ishihara & Towhata (1983) using a hollow cylinder torsional shear apparatus. However, the apparatus could not independently control the inner and outer pressures, resulting in the ratio between the principal stresses, measured by b = (σ2 − σ3 )/(σ1 − σ3 ), and the mean total stress, © 2011 by Taylor & Francis Group, LLC

Figure 9. Response of dense sand under continuous rotation of principal stress directions with different b values: (a) pore water pressure; (b) stress path in q-p
plane.

measured by p = (σ1 + σ2 + σ3 )/3 to cyclically change during loading. Recently, a continuous rotation of the principal stress axes while maintaining conditions of constant b and p was successfully achieved in an automated hollow cylinder apparatus (Yang et al. 2007). Therefore, this loading condition relating to the passage of waves over the free-field seabed can be better replicated. Shown in Figure 9 are selected results for excess pore water pressure for three Toyoura sand specimens consolidated at the relative density of 70% and mean effective stress of 100 kPa but under different b values. In testing the three samples the deviatoric stress was maintained at 34.65 kPa. The deviatoric stress q is defined as

393

where the principal stress σ1 and σ3 are related to the stress components σv , σh , and τvh as

and σ2 is equal to the radial stress (σr ) in the hollow cylindrical configuration. The results shown in Figure 9 indicate that in all the three cases, the rotation of principal axes with constant deviatoric stress was able to generate excess pore pressures in sand specimens. Furthermore, the amount and rate of the pore pressures was strongly dependent on the parameter b. In general, the sand specimen sheared under the condition of b = 0 (i.e. σ2 = σ3 ) showed a much stronger resistance to pore pressure build-up than that under the condition of b = 1 (i.e. σ2 = σ1 ), with the case of b = 0.5 in between. It has also been found that the shear stiffness of the sand specimens degraded during the rotational shear and the degree of degradation was related to the parameter b. A detailed discussion of the tests and results can be found in Yang et al. (2007).

4

CONCLUSIONS

A proper understanding of the cyclic loading behavior of soils is essential in the development of rational technical solutions to tackle various offshore engineering problems. The main points of this paper are summarized as follows: – The cyclic loading can be classified into two major types with reference to offshore applications: (a) a combination of static and cyclic shear stresses, which is in close relation to the stability of gravity offshore structures, and (b) a continuous rotation of principal stress axes with constant deviatoric stresses, which represents the seabed in the free field during the passage of waves. – With regard to the loading type (a), the sand in the vicinity of the offshore structure may fail in three patterns: flow type, cyclic mobility and plastic strain accumulation, and the cyclic shear strength

© 2011 by Taylor & Francis Group, LLC

is closely related to three state parameters: relative
) and density (Drc ), effective confining stress (σnc initial static shear stress ratio (α). – The flow type failure is most catastrophic due to its sudden and run-away nature, whereas in the plastic strain accumulation mode the accumulation of irreversible strains during cyclic loading is more critical than the build-up of pore water pressure. – The loading type (b) can be well replicated using the hollow cylinder torsional shear apparatus that is capable of controlling the axial load, torque, inner cell pressure and outer cell pressure independently. Even dense sand can be weakened by the rotation of principal stress directions without changing the magnitude of deviatoric stress. The degree of pore pressure build-up is significantly affected by the intermediate principal stress parameter b. REFERENCES Been, K. & Jefferies, M. 1985. A state parameter for sands. Géotechnique, 35(2): 99–112. Castro, G. 1975. Liquefaction and cyclic mobility of saturated sands. J. Geotech. Engng. Div., ASCE 101(GT6): 551–569. Ishihara, K. & Towhata, I. 1983. Sand response to cyclic rotation of principal stress directions induced by wave loads. Soils Founds, 23(4): 11–26. Madsen, O.S. 1978. Wave-induced pore pressures and effective stresses in a porous bed. Géotechnique, 28(4): 377–393. Oumeraci, H. 1994. Review and analysis of vertical breakwater failures: Lessons learned. Coastal Eng., 22 (1/2): 3–29. Palmer, A. C., Teh, T. C., Bolton, M. D. & Damgaard, J. S. 2004. Stable pipelines on unstable seabed: Progress towards a rational design method. Proc., Offshore Pipeline Technology Conf., Amsterdam, The Netherlands. Seed, H.B. & Lee, K.L. 1966. Liquefaction of saturated sands during cyclic loading. J. Soil Mech. Founds. Div., ASCE, 92(SM6): 105–134. Yang, J. 2002. Non-uniqueness of flow liquefaction line for loose sand. Géotechnique, 52(10): 757–760. Yang, J. & Sze, H.Y. 2010. Cyclic behaviour and resistance of saturated sand under non-symmetric loading conditions. Géotechnique, to appear. Yang, J., Sze, H.Y. & Heung, M.K. 2009. Effect of initial static shear on cyclic behavior of sand. Proc. 17th Int. Conf. Soil Mech. Geotech Eng., Alexandra, Egypt. Yang, Z.X., Li, X.S. & Yang, J. 2007. Undrained rotational shear and anisotropy of sand. Géotechnique, 57(4): 371–384.

394

5 Shallow foundations

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Observations of shallow skirted foundations under transient and sustained uplift H.E. Acosta-Martinez AECOM Australia (formerly Centre for Offshore Foundations Systems, University of Western Australia)

S. Gourvenec & M.F. Randolph Centre for Offshore Foundations Systems, University of Western Australia, Perth, Australia

ABSTRACT: This paper summarises results from a four-year experimental programme of centrifuge tests investigating installation resistance and uplift capacity of shallow skirted circular foundations in lightly overconsolidated clay. Centric and eccentric, monotonic, transient and sustained loading were investigated and the effect of gapping along the foundation-soil interface was explored. The results from this study showed that for a skirted foundation with embedment ratio d/D = 0.3, reverse end bearing could be mobilised under transient uplift and loads of up to 50% of the undrained soil resistance could be sustained for months without significant displacement – provided nominal contact along the skirt-soil interface was maintained. The presence of gapping along the skirt-soil interface and eccentric loading reduced undrained capacity by up to 40%, and increased displacement rates under sustained uplift loading by up to a factor five.

1

INTRODUCTION

the timescale over which skirted foundations can withstand sustained uplift.

Skirted foundations, or bucket foundations, provide an economically attractive foundation solution offshore to support or anchor structures for the oil and gas industry. A key benefit of skirted foundations lies in their potential to mobilise uplift resistance due to negative excess pore pressures developed between the soil plug and the base plate. Current industry recommendations (DNV, 1992; API, 2009; ISO, 2003) are based on classical bearing capacity theory (Terzaghi, 1943; Brinch-Hansen, 1970; Vesic, 1975) and provide a conservative estimate of undrained collapse loads for skirted foundation systems, particularly due to ignoring the beneficial effect of suction beneath the foundation during transient uplift or overturning. Although industry guidelines and recommended practices acknowledge that temporary suction caused by dynamic loads may allow greater capacities to be mobilised, since the effect is temporary they advise that suction should not be accounted for in the design unless substantiated by appropriate analysis or experimentation. To date there is no formal guidance regarding the timescale over which negative excess pore pressures may be sustained. The conservative limit loads predicted by classical design methods under general loading constitute the main motivation for the study presented in this paper. A challenging issue to improve design is to investigate the response of skirted shallow foundations using appropriate experimental techniques (i.e. with correct stress scaling) to provide quantitative data regarding the response under transient uplift and about © 2011 by Taylor & Francis Group, LLC

2

OVERVIEW OF METHODOLOGY

This project involved an extensive programme of physical modelling in the UWA geotechnical beam centrifuge (Randolph et al., 1991) to investigate the transient and sustained uplift resistance of skirted foundations. 2.1

Foundation model

Foundation models with embedment depth to foundation diameter ratios d/D = 0.15 and 0.3 were used to represent skirted foundations suitable for gravity based structures, jackets and tension leg platforms. The centrifuge models were machined (in-house) with a diameter of 120 mm and skirt lengths of 18 mm and 36 mm. All centrifuge tests were carried out at an enhanced gravity of 167 g, corresponding to a prototype foundation diameter of D = 20 m and skirt depths of d = 3 m and 6 m.An internal cruciform arrangement of internal skirts (stiffeners) was provided. The four compartments formed by the internal stiffener were connected by a small hole at the cross-over to allow for drainage during installation through a single vent in the base plate. Pore pressure transducers (PPTs) and total pressure transducers (TPTs) were placed on the model for monitoring negative excess pore pressures inside the soil plug, contact and separation of the base plate with

397

Figure 2. T-bar penetrometer test results (a) undrained shear strength profile and (b) ratio of extraction to penetration resistance.

Figure 1. Schematic of foundation model (d/D = 0.3) (prototype dimensions).

the soil surface, and radial stresses on the peripheral skirt. All sensors were calibrated in-flight. Previous experience at UWA has confirmed the type of instruments used in the foundation model are reliable in the beam centrifuge (Chen, 2005; Chen & Randolph, 2007). The calibration programme was therefore limited to check the linear response of PPTs andTPTs with applied pressure in water. Calibration factors were validated periodically over the duration of the project. Figure 1 shows a schematic of the foundation model with d/D = 0.3 and location of sensors. 2.2

Soil sample

Lightly over consolidated samples of kaolin clay were used in this study. The kaolin is characterised by plastic and liquid limits of 27% and 61% respectively and a specific gravity of 2.6. The sample was initially consolidated at 1g in a press and then reconsolidated in the centrifuge to give an average (across the 8 samples) su /σv
at skirt tip level of 0.45. This soil strength condition was chosen to ensure it could support a free-standing crack along the skirt-soil interface. The consolidation history led to an in situ void ratio e0 = 1.3, wet unit weight γt = 17 kN/m3 and over consolidation ratio at skirt tip level of ∼4. A representative value of the coefficient of consolidation at the vertical stress at skirt tip level (∼40 kPa) was taken as cv(av) = 2.6 m2 /year based on results of one-dimensional consolidation tests, corresponding to a coefficient of vertical permeability of ∼10−9 m/s (Acosta-Martinez & Gourvenec, 2006). The shear strength profile was assessed inflight using a T-bar penetrometer (Stewart & Randolph, 1994). Figure 2 shows typical profiles of undrained shear strength and the ratio of extraction © 2011 by Taylor & Francis Group, LLC

Figure 3. Cyclic T-bar penetrometer test results.

to penetration resistance based on a constant T-bar factor, NT−bar = 10.5 (Stewart & Randolph, 1994). The undrained shear strength was broadly repeatable between samples, although some variations did exist across the 8 samples used during the study; for example, su at tip level for the foundation with d/D = 0.3 varies by up to 20% between samples (which could lead to a 20% difference in measured foundation resistance while indicating the same bearing capacity factor). Figure 3 shows the degradation in undrained shear strength at skirt tip level for the foundation with d/D = 0.3, from cyclic T-bar tests (both carried out in the same sample).The degradation factor gives an indication of the soil sensitivity of the kaolin clay used in this study (Yafrate et al., 2009), and may be considered indicative of the interface friction factor, α, to account for shear on the skirt-soil interface during foundation installation. Figure 3 indicates a value of α between 0.3 and 0.4 is indicated from the cyclic T-bar tests. 2.3

Experimental procedures

The foundation models were attached by a rigid arm to a one-dimensional actuator allowing load or displacement control via a ±3 kN load cell and a 25 mm

398

out in virgin sites (i.e. sites were not reused for multiple tests). 3 3.1

© 2011 by Taylor & Francis Group, LLC

Installation resistance

Figure 4 shows a typical measured profile of installation resistance in terms of average stress (load cell measurement divided by the cross-sectional plan area of the foundation). Installation resistance was repeatable for all the tests in the programme. Initially, the skirts and stiffeners penetrated until the base plate touched down on the soil surface (0 < z/D < ∼0.26). Jacking was continued until a distinct increase in the load cell reading was observed, indicating that good contact had been achieved across the base plate. Similarity of readings from diametrically opposed pairs of TPTs confirmed the verticality of the foundation during installation; also the PPT under the lid showed that no excess pore pressures were developed during installation, indicating that the vent hole was sufficient to allow egress of water during installation. The average embedment reached during installation was ∼95% of the total skirt length indicating that the skirt and stiffeners were accommodated equally by inward and outward flow of the soil (i.e. a 50:50 split). A simple theoretical model, considering installation resistance as the sum of plate bearing and skirt friction (Equation 1) was used to back-calculate the interface friction factor, α.

Figure 4. Installation resistance.

stroke linear displacement transducer (LDT). Load and displacement control was achieved via servocontrolled actuators on the centrifuge. Loads and displacements were monitored and recorded by the centrifuge’s data acquisition system at a typical frequency of 10 Hz. The load cell was zeroed once the foundation was suspended above the soil surface and submerged in the free water, such that measured loads are net of the weight of the foundation. The model foundations were installed in-flight, with the drainage valve open to allow water egress from the skirt compartments. The drainage valve was then sealed and sufficient time was allowed for re-consolidation (i.e. dissipation of the excess pore pressure generated during installation) to occur prior to a transient or sustained uplift test. In most cases, re-consolidation was carried out under load-control (at the load achieved at the “end of installation” as marked in Figure 4). Displacement-controlled reconsolidation was required for the eccentric uplift tests to prevent the foundation rotating (under the eccentrically positioned loading arm). A base-line centric load test for comparison of the eccentric load tests, i.e. e/D = 0, was carried out following re-consolidation under displacement-control for appropriate comparison of the results. Transient loading was achieved with displacementcontrol at a rate of 0.1 mm/sec to achieve undrained conditions with respect to the cross-sectional area of the foundation. Taking a representative value of coefficient of consolidation of 2.6 m2 /yr, based on 1-D consolidation tests (Acosta-Martinez & Gourvenec, 2006), the non-dimensional velocity V = vD/cv exceeds 30, the limit for an undrained soil response (Finnie & Randolph, 1994). Local drainage is likely near the skirt tip and internal stiffeners) where V ∼ 1. Sustained load tests were carried out by applying a constant average bearing pressure, q, defined as a fraction of the transient ultimate capacity per unit area under centric uplift, qu . The study described in this paper involved 8 strongbox samples and 32 foundation tests considering transient and sustained uplift, and the effect of gapping along the skirt-soil interface and load eccentricity were also considered. All tests reported were carried

OBSERVATIONS

where Nc = bearing capacity factor for tip resistance; su0 and suav are values of undrained shear strength at foundation tip level and averaged over the embedment depth; Ab = total end bearing area; and As = internal and external surface area of the skirt and stiffeners. Values of interface friction factor 0.3 < α < 0.4 were back calculated from the profiles of installation resistance across the 8 samples and 32 tests. The backcalculated values of α are in good agreement with the degradation factor predicted from cyclic T-bar tests (shown in Figure 3). Detailed interpretation of installation of skirted foundations is presented in Gourvenec et al. (2009) and Acosta-Martinez and Gourvenec (2010). 3.2 Transient (undrained) uplift Figure 5 shows net transient uplift resistance normalised by the initial undrained shear strength at skirt tip level, su0 (i.e. measured by the T-bar), plotted against normalised uplift displacement, w/D. Uplift resistance is taken as load cell measurement divided by the cross-sectional plan area of the foundation. The load cell measurements have been corrected for the increasing difference in overburden pressure outside the foundation and the weight of the soil plug during uplift. Load displacement responses are shown

399

Figure 5. Transient uplift response.

for cases of (i) centric loading with nominal contact along the skirt-soil interface for foundation embedment ratios d/D = 0.15 and 0.3; (ii) centric loading with nominal contact and with a gap along the skirt-soil interface for d/D = 0.3; and (iii) centric and eccentric loading with nominal contact along the skirt-soil interface for d/D = 0.3. Considering firstly the effect of foundation embedment ratio (final points on curves labelled “d/D = 0.15 no gap” and “d/D = 0.3 no gap” in Figure 5), a marked difference – nearly 300% – is observed in the measured ultimate uplift capacity between the foundations with d/D = 0.15 and 0.3, while plasticity solutions of bearing capacity for shallow foundations failing in general shear (e.g. Martin & Randolph, 2001) indicate a 20–25% difference over the same change in embedment ratio. It is not strictly appropriate to compare the measured normalised uplift resistance with theoretical bearing capacity factors since an increase in operative shear strength occurs during re-consolidation following installation and prior to uplift. The increase in operative shear strength, compared to the in situ shear strength measured by the T-bar and used to normalise the observed uplift resistance, will imply an artificially high ‘apparent’ bearing capacity factor. Theoretical bearing capacity factors are also only available for linearly increasing shear strength profiles as opposed to the parabolic profile of the lightly overconsolidated soil sample in these tests. Nonetheless, a bearing capacity factor Nc of no less than 6.5 would be expected for a foundation with d/D = 0.15 based on theoretical solutions for reverse end bearing (e.g. Martin & Randolph, 2001). An upper limit to pullout resistance under local shear failure of the foundation, i.e. simple pull-out, can be estimated with Equation 1 using α = 1 (after consolidation). This corresponds to qu /su0 = 2.2, three fold less than the lower limit of reverse end bearing capacity, and indicates that local shear governed failure of the foundation with d/D = 0.15. Bearing capacity factors in the range ∼7 < Nc < ∼12 are predicted by lower and upper bounds for d/D = 0.3 across the full range of skirt-soil

© 2011 by Taylor & Francis Group, LLC

interface roughness, comparing well with the measured apparent bearing capacity factor of 9. The load cell data and the PPT measurements under the base plate, in conjunction with visual observations during testing and inspection of the sites after testing, indicated a local shear mechanism governed failure of the skirted foundation with d/D = 0.15 while general shear reverse end bearing governed failure of the foundation with embedment ratio d/D = 0.3. Current industry guidelines assume local shear governs uplift resistance, which can significantly under-estimate the capacity of a skirted foundation if reverse end bearing is mobilised. For example, for the upper limit of uplift resistance under local shear, the normalised uplift resistance of the foundation with d/D = 0.3 in this study is qu /su0 = 5.4 or 40% less than actually mobilised with reverse end bearing. Since reverse end bearing was not mobilised with the lower embedment ratio, further testing to investigate the effect of gapping and load eccentricity under transient and sustained loading concentrated on the deeper skirted foundation model with embedment ratio d/D = 0.3. The effect of gapping on the transient uplift capacity was investigated by creating a gap (in-flight under horizontal displacement control) along the skirt-soil interface. Two conditions were investigated (i) uplift immediately after gap formation and (ii) uplift after an extended period of re-consolidation following gapping. No difference in uplift resistance, qu /su0 , or failure mechanism was observed under transient uplift immediately after formation of a gap. In contrast, a 40% reduction in capacity was observed if a period of re-consolidation was allowed following crack formation prior to transient uplift. Reduced undrained uplift capacity following gapping and re-consolidation are likely to be related with softening of the surrounding soil and vertical and lateral propagation of the gap, creating a hydraulic connection between the soil plug contained within the skirts and the free water surface, thus preventing development of reverse end bearing. A complete description of the tests involving gapping, including the apparatus development, is presented by Acosta-Martinez et al. (2010a). The effect of load eccentricity was evaluated for normalised eccentricity, e/D = 0.12 and 0.25 compared with a base-line case e/D = 0. The difference in ultimate capacity between the centric load tests with nominal contact for the same embedment ratio (labelled “d/D = 0.3 no gap” and “e/D = 0” in Figure 5) arises due to the load- and displacement-controlled consolidation (discussed in Section 2.3). The consolidation stress continually diminished throughout displacement-controlled consolidation contributing to the lower uplift capacity. As would be expected, load eccentricity has a detrimental effect on the transient uplift capacity, with the effect increasing with the degree of eccentricity. Ultimate undrained uplift capacity was reduced by 30% due to a loading at an eccentricity of one quarter of the distance from the mid-line of the foundation and by

400

Figure 6. Displacement time histories during sustained uplift.

Figure 7. Effect of eccentric loading on time–dependent displacements during sustained uplift.

40% due to a loading eccentrically half way between the centre and the edge of the foundation. A kinematic mechanism involving pure rotation followed by simultaneous rotation and vertical displacement was observed during the tests. Gapping along the skirt-soil interface or loss of negative excess pore pressure inside the soil plug were not observed during eccentric uplift study in these tests. Details of the equipment, test procedures and interpretation of the suite of tests investigating eccentric loading are presented by Acosta-Martinez et al. (2010b). Figure 8. Degradation of uplift resistance under sustained load.

3.3 Sustained uplift Figure 6 shows time histories of displacements normalised by the foundation diameter measured during sustained uplift for the cases of nominal skirt-soil contact and for gapping along the skirt-soil interface. Figure 6 shows the time-dependent ‘consolidation’displacements, wc , after deducting the immediate ‘elastic’ displacement, wi . The component of immediate displacement depended on the magnitude of uplift load but typically gave wi /D < 1% (as indicated in Figure 7). Sustained loading tests were stopped at a maximum displacement of wc /D ≥ 2%, or a prototype time tp = 1 year, whichever occurred first. As would be expected, the rate of displacement increased with increasing load. If contact was maintained along the skirt-soil interface, loads of 20% of the uplift capacity were sustained for up to a year with displacements of <2% of the diameter, while the same displacement occurred within a month for a load of 70% of the uplift capacity. Gapping had a detrimental effect on foundation performance, increasing the rate of displacement up to five-fold, even at low loads. Nonetheless, the displacement rates remained relatively constant during the duration of the tests with gapping. Figure 7 shows the effect of eccentric loading on the rate of displacement under sustained uplift. The magnitude of immediate elastic displacement and the rate

© 2011 by Taylor & Francis Group, LLC

of consolidation displacement increased with increasing load eccentricity. Rates of displacement increased 2.5 fold and 15 fold for eccentricities e/D = 0.12, 0.25 respectively. Initial rotation followed by rotation and vertical displacement of the foundation as observed under transient uplift was also observed under sustained uplift. Figure 8 shows the degradation of uplift capacity, as a fraction of the ultimate undrained capacity under centric uplift, q/qu . Curves are presented in terms of normalised total displacements (i.e. including immediate elastic displacements) w/D of 0.5%, 1% and 2%. For the soil conditions used in this study and foundation embedment ratio d/D = 0.3, when nominal contact was maintained along the skirt-soil interface, uplift loads q/qu = 0.2 were sustained for two years without total normalised displacement, w/D, exceeding 2%. For the same displacement criterion, loads in the region of half the ultimate vertical capacity were sustained for a month or more. Gapping along the skirt-soil interface or load eccentricity led to a higher rate of displacement (as shown in Figures 6 and 7), and a consequent increased rate of degradation of uplift capacity with time.

401

4

SUMMARY OF OBSERVATIONS

For the foundation and soil conditions considered in this study, full reverse end bearing could be mobilised with a skirted foundation with embedment ratio d/D = 0.3 provided nominal contact was maintained along the skirt-soil interface. A centric vertical load of 20% of the ultimate undrained uplift capacity was sustained for two years without significant displacement provided nominal contact was maintained along the skirt-soil interface. Gapping along the skirt-soil interface and load eccentricity reduced the undrained uplift capacity by up to 40%. Gapping and load eccentricity also led to an increase in the rate of displacements under sustained loading; similar displacements observed after one year under centric load when nominal contact was maintained were observed within a couple of weeks in the presence of a gap along the skirt-soil interface or under eccentric load. The presence of gapping along the skirt soil interface short circuits the drainage path, which leads to accelerated dissipation of negative excess pore pressures in the soil plug. The detrimental effect of gapping could be addressed with provision for mitigation of gapping in the form of gap arrestors surrounding the foundation. A practical design of gap arrestor is currently under investigation at COFS. From a practical perspective, the results summarised in this paper constitute a promising finding for future applications of shallow skirted foundations to resist uplift for more than transient conditions without excessive deformations. Explicit consideration of negative excess pore pressures during uplift can translate into higher efficiencies for shallow skirted foundation systems compared with traditional methods of design, provided a nominal seal along the skirt-soil interface can be maintained. Among the spectrum of structures that may benefit are gravity based structures, jacket (or template) structures, tension leg platforms and the novel concept of offshore gravity based storage tanks that could be buoyant for a period of weeks following offloading.

REFERENCES Acosta-Martinez, H.E. & Gourvenec, S. 2006. Onedimensional consolidation tests on kaolin clay. Research Report GEO: 06385, Centre for Offshore Foundation Systems, University of Western Australia.

© 2011 by Taylor & Francis Group, LLC

Acosta-Martinez, H.E. & Gourvenec, S. 2010. Installation resistance and bearing capacity of a shallow skirted foundation in clay. Proc. 7th Int. Conf. on Physical Modelling in Geotechnics. Zurich, 2:1007:1012. Acosta-Martinez, H.E., Gourvenec, S.M. & Randolph, M.F. 2010a. Effect of gapping on the transient and sustained uplift capacity of a shallow skirted foundation in clay. Soils and Foundations (accepted). Acosta-Martinez, H.E., Gourvenec, S. & Randolph, M.F. 2010b. Centrifuge study of capacity of a skirted foundation under eccentric transient and sustained uplift. Géotechnique (accepted). API 2009. RP2A: Recommended practice for planning, designing and constructing fixed offshore platforms., Washington, DC. Brinch-Hansen, J. 1970. A revised and extended formula for bearing capacity. Danish Geotechnical Institute Bulletin No.28, 5–11. Chen, W. 2005. Uniaxial behaviour of suction caissons in soft deposits in deepwater. PhD thesis, The University of Western Australia. Chen, W. & Randolph, M.F. 2007. External radial stress changes and axial capacity for suction caissons in soft clay. Géotechnique, 57(6): 499–511. DNV 1992. Classification notes No. 30.4, Foundations. Finnie, I.M.S. & Randolph, M.F. 1994. Punch-through and liquefaction induced failure of shallow foundations on calcareous sediments. Proc. International Conference on Behaviour of Offshore Structures – BOSS ’94, 217–230. Gourvenec, S.M., Acosta-Martinez, H.E. & Randolph, M.F. 2009. Experimental study of uplift resistance of shallow skirted foundations in clay under transient and sustained concentric loading. Géotechnique, 59(6): 525–537. ISO 2003. 19901-4 Petroleum and natural gas industries: Offshore structures: Geotechnical and Foundation Design. Martin, C.M. & Randolph, M.F. 2001. Applications of the lower and upper bounds theorems of plasticity to collapse of circular foundations. In Proc. Int. Conf. on Computer Methods and Advanced Geomechanics. Tucson, 2, pp. 1417–1428. Randolph, M.F., Jewell, R.J., Stone, K.J.L. & Brown, T.A. 1991. Establishing a new centrifuge facility. Proc. Int. Conf. Centrifuge 91. Boulder, Colorado. pp. 3–9. Stewart, D.P. & Randolph, M.F. 1994. T-bar penetration testing in soft clay. J. of Geotechnical Engineering, 120(12): 2230–2235. Terzaghi, K. 1943. Theoretical Soil Mechanics. Wiley, New York. Vesic, A.S. 1975. Bearing capacity of shallow foundations. Foundation Engineering Handbook. Van Nostrand, New York. pp. 121–147. Yafrate, N., DeJong, J., DeGroot, D. & Randolph, M. 2009. Evaluation of remolded shear strength and sensitivity of soft clay using full-flow penetrometers. J. Geotech. and Geoenv. Engng, 135(9): 1179–1189.

402

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Numerical study of grillage foundations on sand under combined VHM loading M. Banimahd, A. Maconochie & J. Oliphant Offshore Engineering Division, Technip, Aberdeen, UK

ABSTRACT: Grillage foundations have been frequently used to support offshore subsea structures such as manifolds, PLETs and PLEMs. They offer lighter options as compared to solid foundations of the same dimensions and this could be beneficial when the crane lift capacity on the installation vessel is limited. Despite wide spread application of these foundations and their potential advantages, no rigorous approach is available to evaluate their responses under combined loading and consequently to design these kinds of foundations. In the paper, the bearing capacity of grillage foundations on sandy soils is investigated under combined vertical, horizontal and moment loading employing the finite element method. Before carrying out grillage simulations, the model configuration is validated against the problems of single foundations and two interfering strip foundations on sand with available analytical solutions. Two dimensional and three dimensional analyses are then conducted where the effects of spacing and foundation roughness on the response of the grillage foundation are explored. The failure mechanisms of the grillage foundations under combined loading are also discussed in detail based on the results of the simulations. The paper proposes a 3D failure surface for the design of grillage foundations subjected to horizontal, vertical and moment loading. 1

INTRODUCTION

A grillage foundation consists of perpendicular sets of steel beams, as shown in Figure 1. These foundations are lighter and also experience lower hydrodynamic resistances during installation, in comparison to an equivalent solid mudmat. These make them preferable options as far as installation of subsea structures is concerned. The problem of bearing capacity of a grillage foundation is a three dimensional problem. To simplify the problem, these types of foundations can be however idealized as two perpendicular sets of parallel strip footings which interact together to resist the applied load. These closely spaced strip footings cannot be treated as isolated footings, due to interactions between them. Among others Stuart (1962) and Javadi & Spoor (2004) indicated that depending on the spacing between two closely spaced foundations they can undergo different failure mechanisms and consequently have different vertical bearing capacities. At a wide spacing, no interference takes place and the total load on the pair of footings will be simply twice the bearing capacity of a single footing. As the spacing reduces, passive bearing zones of footings interfere and the failure surface of each footing becomes asymmetrical. Eventually as the footings get closer, a state will be reached in which the pair will act as a single foundation in this state; the soil between the individual units forms an inverted arch, which moves down with the footing as the load is applied. A higher degree of interaction is expected for sands with larger friction © 2011 by Taylor & Francis Group, LLC

Figure 1. A grillage foundation supporting a pipeline valve.

angles when the spacing is small (see e.g. Kumar and Ghosh, 2007). Offshore foundations are often exposed to combined vertical (V), horizontal (H) and moment (M) loading. The response of a rigid footing on sand subjected to combined loading can be represented by projections of a unique 3D interaction surface, on the V-H, V-M/B and H-M/B planes (B is width of foundation). It has been shown that the failure envelopes in V-M/B and H-M/B planes would take symmetrical parabolic forms whereas the failure envelope on H-M plane could be expressed by inclined ellipses for a single foundation on sand (see e.g. Gottardi & Butterfield, 1995). The inclination of the failure surface in

403

Figure 2. 2D representation of a grillage foundation (positive direction of loads and displacements are shown).

the plane implies that the horizontal/moment capacity of a shallow foundation on sand is a function of the direction of the applied moment/horizontal load for a single foundation. Employing the V-H-M framework, this paper reports the results of a finite element study on grillage foundations under combined loading. The effects of spacing, foundation roughness and 3D nature of the problem have been taken into account.

2 2.1

Figure 3. A 3D finite element model of a grillage foundation.

FINITE ELEMENT MODELLING General consideration

In this study both a 2D representation of a grillage foundation (Figure 2) and a 3D grillage foundation model (Figure 3) are considered. Utilising the symmetry under vertical loading, only a quarter of the foundation is modeled in the 3D simulations as shown in Figure 3. To investigate the interaction of a foundation under vertical load, a global efficiency factor is defined as the ratio of bearing capacity of a grillage foundation (a set of parallel strip footings) to the sum of the capacity of the individual footings when acting individually. Employing ABAQUS 6.7-1, small strain simulations have been conducted to investigate the response of grillage foundations under combined horizontal, vertical and moment loading as schematically shown in Figure 2. Different spacing (S) ranging from 1b to 5b (b is footing width and B is overall foundation width) are investigated (Fig. 2). A loose sand condition (φ
= 30◦ ) is considered in this section unless otherwise stated. The MohrCoulomb constitutive model has been employed for loose sand (φ
= 30◦ ). Different interface conditions, namely fully rough (δ = φ
= 30◦ ), relatively rough (δ = φ
− 5◦ = 25◦ ) and smooth (δ = 0.5◦ ) cases have been considered where required. Analogous to the experimental investigation methods, the failure envelopes can be numerically obtained using two different approaches, namely a loaddisplacement controlled probe and swipe tests. In the load-displacement controlled probe test, several load tests are conducted to obtain the points of failure of the foundation under combined loading. In the swipe test (side swipe test, Tan, 1990), in order to establish the failure surface in the ij loading plane, a displacement in the i-direction is imposed until the ultimate load in the i-direction is reached. Keeping the displacement in the i-direction fixed, the foundation is displaced in the j direction until the ultimate load in the © 2011 by Taylor & Francis Group, LLC

Figure 4. The effect of meshing on vertical response of a single rough foundation (B = 2m).

j direction is reached. It is postulated that the load path obtained would trace the failure surface in the respective plane (e.g. V-H and V-M planes). In H-M plane, fixed displacement ratio (i.e. the ratio of horizontal displacement to rotation is kept constant) simulations are conducted to ensure that the failure surface in the H-M plane has been identified correctly. 2.2

Model configuration

To select an appropriate meshing configuration, the well known problem of a shallow foundation is revisited. A strip foundation (B = 2 m) on loose sand (φ
= 30◦ , γ = 20 kN/m3 ) is modeled. Different mesh configurations are considered with different meshing densities around the footing. The soil-foundation interface is assumed to be rough (δ = φ
). Figure 4 depicts the footing vertical responses for two different meshing configurations as compared with Terzaghi’s solution (Nγ = 19.7). The finer mesh provides the closest response to Terzaghi’s solution and is therefore adopted for this study. Assuming sand behaves as an elastic-perfectly plastic material with constant friction angle, the bearing capacity factor (Nγ ) is constant, irrespective of foundation size. However, elastic-perfectly plastic finite element

404

Figure 7. Efficiency factors for a 2D grillage foundation on loose sand with different interface conditions.

Figure 5. Numerical Swipe tests against experimental failure envelopes in V-M planes for a single rough strip footing (B = 2 m).

(see Kumar & Ghosh, 2007). The calculated efficiency factor matches well with the values associated with failure mechanism 2. A considerable rise in bearing capacity of interfering footings is simulated when the spacing ratio is less than or equal to one. 3 VERTICAL LOADING Figure 7 presents the efficiency factor for 2D grillage foundations with different spacings on loose sand. For spacings higher than 1b, the interactions between strip footings are limited and they effectively resist the load separately. As expected, the smooth grillage foundation results in lower efficiency for smaller spacing in comparison to rough and relatively rough grillage foundations.

Figure 6. Comparison of numerical efficiency factors and the theoretical solutions (Kumar and Ghosh, 2007) for two interfering rough strip footings on medium dense sand.

4

COMBINED LOADING

4.1 Vertical-Horizontal loading simulations can produce a spurious reduction of bearing capacity factor (Nγ ) with increasing foundation size, depending on the meshing arrangements (Woodward & Griffiths, 1998 and Banimahd & Woodward, 2006). To minimise this effect, as recommended by Woodward & Griffiths (1998), the mesh is scaled based on the foundation size, i.e. the number of elements underneath the foundation and meshing configuration are kept unchanged irrespective of the foundation size. Swipe tests in V-H and V-M planes are also conducted on the single strip footing for validation purposes. The numerical swipe tests trace the experimental failure envelopes (e.g. Gottardi & Butterfield, 1995) in V-H and V-M planes reasonably well (only V-M failure plane is shown in Figure 5). The case of two interfering rough strip footings on medium dense sand (φ
= 35◦ ) has also been considered as part of the validation study. Different spacings ranging from S = 0.4b to 4b are considered. Figure 6 compares efficiency factors computed from the finite element simulations compared to the analytical solutions for two different assumed failure mechanisms © 2011 by Taylor & Francis Group, LLC

The horizontal capacity of a grillage foundation is expected to be significantly affected by soilfoundation interface friction. In case of a smooth foundation, sliding takes place along the soil/foundation interface irrespective of the applied vertical load (i.e. Hult always equals V.tanδ). This is not, however, the case as the interface becomes rougher. Figure 8 presents the V-H failure point against proposed failure surfaces for rough and relatively rough cases. It is seen that the horizontal capacity of a relatively rough footing is slightly less than that of a rough one at a given vertical load. These failure surfaces are very similar to the failure surface proposed by Gottardi and Butterfield (1995) for a rough single foundation, indicating that the spacing between the foundations does not have an important effect on the interaction between normalized vertical and horizontal loads.

405

4.2 Vertical loading-moment Several load control numerical simulations have been conducted to define failure points at different vertical load levels in V-M plane as presented in Figure 9

Figure 8. Proposed failure envelope against failure points in V-H plane for different spacing.

Figure 11. Rotational response of a 2D grillage foundation under positive and negative moment at a given vertical and horizontal load ((a) S = 1b, (b) S = 3b).

Figure 9. Proposed failure envelope against failure points in V-M plane for different spacing.

This is a different rotational resistance mechanism when compared to a single continuous foundation, as no rotation failure mechanism is seen around the strip footings especially if the spacing is considerably large (S > 1b).

Figure 10. Plastic strain distribution underneath grillage foundation with different spacing under V-M loading.

along with proposed failure envelopes. According to this figure, the normalised resisting moment slightly increases as the spacing increases from 1b to 5b. Comparing the failure envelopes for grillage foundations and solid foundation (Fig. 5 versus Fig. 9), it can be said that grillage foundations resist considerably higher normalised moments. This can be advantageous in cases where the design is mainly governed by moment capacity rather than vertical bearing capacity. Figure 10 depicts the failure mechanisms for 2D grillage systems with different spacing under combined V-M loading. While there are some interactions between the components of the grillage foundation with S = 1b, no interaction exists in the case of S = 4b. In addition, especially for the foundation with higher spacing, the local rotations of grillage components are limited and they resist the moment through their vertical bearing resistance as shown in Figure 10. © 2011 by Taylor & Francis Group, LLC

406

4.3

Horizontal load-moment

Figures 11a and b compare the rotational response of 2D grillage foundations with different spacing under positive and negative moments. It is clearly seen in Figure 11a that the rotational capacity of a foundation under negative moment is higher than that of a foundation under positive moment for S = 1b (positive horizontal loads were applied; see Figure 2 for definition of positive direction). Such a visible difference is not indicated when the rotation responses of grillage foundations with S = 3b are simulated under positive and negative moments (Fig. 11b). Insensitivity of the rotational capacity to the direction of the horizontal load for grillage foundations with large spacing can be linked to the rotational resistance mechanism of these foundations being different from that of single foundations (see section 4.2). This also suggests that as the spacing increases, the skewness of failure envelope in the H-M plane would reduce. According to Figure 12a, the failure envelope for S = 1b is slightly skewed to reflect the dependency of

Figure 12. Proposed failure envelope in H-M plane for S = 1b (a) and S.1b (b).

Figure 13. 3D simulation against 2D simulation of grillage foundation under vertical loading (a) S = 1b (B = 0.9 m), (b) S = 4b (B = 2.5 m).

rotational-horizontal bearing capacity on the direction of the applied load. However, the failure envelope for S > 1b is considered to be circular (Fig. 12b). As seen in the Figures 12a and 12b the fixed ratio displacement tests trace the proposed failure surfaces reasonably well.

5 THREE DIMENSIONALITY Figures 13a and 13b present the vertical responses of 3D grillage foundations against those of 2D grillage models. To aid comparison the vertical bearing capacities of square foundations (with the same net areas as the respective grillage foundations) have also been calculated using the classical bearing capacity theory. In the 2D grillage model, an obvious failure point (where response gradient changes significantly) is noted for S = 1b, but no apparent failure point is seen in the 3D grillage foundation response up to the simulated level of displacement (Fig. 13a). Considering the foundation force at the last step as a conservative approximation of the failure (near failure) point, it can be judged that the bearing capacity of the 3D model is close to the bearing capacity of the equivalent square foundation. This is, however, not the case for larger spacings as shown in Figure 13b; the bearing capacities of 3D grillage foundations become significantly lower than equivalent square ones. For a spacing of 4b, the failure loads from both 2D and 3D grillage © 2011 by Taylor & Francis Group, LLC

407

Figure 14. Plastic strain distribution under beam crossing points for grillage foundations with different spacings (vertical displacement: 0.01 m).

foundations are remarkably close while the bearing capacity of an equivalent square foundation is significantly higher than the bearing capacities of both 2D and 3D grillage foundations. The distribution of the failure zones on the surface and in depth in 3D simulations are shown in Figure 14 for S = 2b and S = 4b cases for a given displacement of 0.01 m. A difference in the failure mechanisms of

the grillage foundations with S = 2b and S = 4b is distinguished. For S=4b, the failure mechanism is fully developed under the grillage beams both away from and at crossing points (where perpendicular beams cross). For S = 2b, the full failure surface is only about to form under the beam in a section away from crossing points. Full failure surfaces are not seen to develop at crossing points for S = 2b (Fig.14). This implies that the grillage foundation (S = 2b) can accommodate more load due to additional soils resistance under the crossing points. The significant increase in soil resistance under crossing points, as compared to those in mid sections, can be attributed to local increases in the confining pressure in soil due to 3D constraints. 6

CONCLUSIONS

In this paper, the bearing capacity of grillage foundations on sand under horizontal, vertical and rotational combined loading has been investigated numerically. The effects of three dimensions, the soilfoundation interface and foundation spacing have all been addressed. The main findings are summarized below: – The vertical capacity of a grillage foundation on sand is a function of the spacing between the individual beams. – The normalized V-H failure envelope of a relatively rough grillage foundation is very similar to that of a single foundation and is not affected by the spacing. – The normalized resisting moment slightly increases with spacing. Grillage foundations can resist considerably higher normalized moments than solid foundations.

© 2011 by Taylor & Francis Group, LLC

– The rotational capacity of a grillage foundation with higher spacing (S > 1b) is insensitive to the direction of applied horizontal load. – 3D constraints significantly affect the vertical bearing capacity of grillage foundations with low spacing. Based on the results of this study, a 3D failure envelope for these foundations has been proposed through defining the failure surfaces in V-H, V-M and H-M planes. REFERENCES Banimahd M. and Woodward P. K. 2006. Load-displacement and bearing capacity of foundations on granular soils using a multi-surface kinematic constitutive soil model, International Journal for Numerical and Analytical Methods in Geomechanics, 30(9): 865–886. Gottardi G. and Butterfield R. 1995. The displacement of a model rigid surface footing on dense sand under general planar loading, Soils and Foundations, 35 (3):71–82. Javadi A. and Spoor G. 2004. Soil failure patterns and loadsinkage relationships under interacting shallow footing and wheel arrangement, 88(3): 383–393. Kumar J. and Ghosh, P. 2007. Ultimate bearing capacity of two interfering rough strip footings, International Journal of Geomechanics, 7(1):53–62. Stuart, J.C. 1962. Interference between foundations, with special reference to surface footing in sand, Geotechnique, 12 (1): 15–22. Tan K. 1990. Centrifuge and theoretical modeling of footing on sand, PhD thesis, University of Cambridge, UK. Woodward P.K. and Griffiths D.V. 1998. Observation on the computation of the bearing capacity factor Nγ , Geotechnique, 48(1): 137–141.

408

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

The vertical bearing capacity of grillage foundations in sand M.F. Bransby, P. Hudacsek, J.A. Knappett & M.J. Brown The University of Dundee, Dundee, UK

N. Morgan & D.N. Cathie Lloyd’s Register, Aberdeen, UK & Cathie Associates, Brussels, Belgium

R. Egborge Subsea7, Aberdeen, UK

A. Maconochie & G.J. Yun Technip, Aberdeen, UK

N. Brown & A. Ripley Acergy Norway AS, Norway & Acergy, Aberdeen, UK

ABSTRACT: Grillage foundations have been suggested as an alternative to flat plate foundations for seabed infrastructure such as pipeline end manifolds and pipeline end terminations.These foundations have the advantage that they can be installed in an offshore environment more quickly because of their hydrodynamic properties. Seabed infrastructure is typically subjected to a combination of vertical, self-weight loading and horizontal loading from pipelines, snag loads or hydrodynamics and so the combined vertical-horizontal capacities of the foundations are critical. However, little is known about the capacity of grillage foundations either to purely vertical loading or to combinations of load. This paper reports a series of physical model tests designed to address this knowledge gap. The experimental methods are presented first followed by typical results showing drained foundation capacity for pure vertical loading. 1

INTRODUCTION

Shallow foundations may be used to support offshore infrastructure such as pipeline end manifolds (PLEM), pipeline end terminations (PLET) or temporary anchors. Shallow foundations can directly rest on the seabed surface (for example they may consist of a flat steel plate – a ‘mudmat’), or may be skirted if large loadings are expected or soil conditions are poor. Manifold foundation structures are installed by lowering them from a vessel with a crane to the seabed. If the structure is large, good sea conditions are required because of splash zone wave forces. This means that installation may be delayed whilst waiting for appropriate weather conditions with cost consequences. During operation, a manifold is subjected to the vertical dead weight structural loading, W plus any additional loading during operation. These additional loads are likely to be horizontal and caused by (i) snag loads (from trawling gear or anchors), (ii) pipeline expansion or jumper loads, or (iii) hydrodynamic loads (in the case of shallow water). In many cases these will be applied relatively close to the level of the seabed because of the low height of the structures. This means that moment loads will be small and so combinations © 2011 by Taylor & Francis Group, LLC

409

of vertical and horizontal load (albeit potentially with some torsional loading) will govern the foundation design. The capacity of foundations under combinations of vertical and horizontal (i.e.V-H) loading has been studied for a range of soil conditions. There are two main design approaches: (i) the Terzaghi bearing capacity method, and (ii) the failure envelope approach. In the bearing capacity method, modification factors for load inclination are applied to the vertical bearing capacity equations as first suggested by Terzaghi (1943). More recently, use of the failure envelope method has become more common. In this method, the load components (V, H and M) applied to the foundation are plotted on an interaction diagram and an envelope of failure states is ascertained (e.g. Zaharescu, 1961; Georgiadis, 1985; Nova & Montrassio, 1991; Gottardi & Butterfield, 1993). Equations describing failure envelopes have been presented by a range of authors for different soil conditions (e.g. sand, clay.) and foundation types (e.g. plan shapes, embedment ratios etc.). It should be noted that the majority of equations for drained conditions (e.g. Nova & Montrassio, 1991; Gottardi & Butterfield, 1993) have been for flat plate foundations resting on the soil surface.

Figure 1. Schematic of grillage geometry (Note: Not to scale. Many more grilles would be used in an offshore foundation).

Recently, grillage foundations have been used in several offshore projects to replace conventional flat plate foundation. Grillages (Figure 1) consist of many shallow, thin vertical grilles connected rigidly together. Typically, each grille is of thickness, t = 5–10 mm, length, d = 50 mm and have a centre-to-centre spacing, s varying from 20 mm to 80 mm. Grillages have the advantage over conventional foundations because water may flow between the grilles during installation through the splash zone. This will allow installation in poorer weather conditions and so have financial advantages to the contractor as they are likely to reduce vessel utilisation time during offshore installation. In addition, there is a possibility that the foundations may require less steel than conventional flat steel plate foundations and have a larger horizontal sliding capacity. Because of their novelty there is currently no generally accepted method to calculate the bearing capacity of grillages under either pure vertical or combined vertical – horizontal loading. In addition, it is not clear how bearing capacity is affected by the spacing, s, of the grilles and their thickness, t (or spacing ratio, s/t) for different soil conditions. For example, what soil condition and spacing ratio combinations are grillage foundations sufficient so that they can be used instead of flat plate foundations? A joint industry project was conducted with the aim of filling this knowledge gap. An extensive series of laboratory physical model tests were conducted by the University of Dundee which investigated first the vertical bearing capacity of grillage foundations and then the combined V-H capacity in drained, sandy soil. This paper considers the capacity of grillage foundations under pure vertical loading for plane strain grillage conditions. It reports first the experimental methodology used in this study followed by presentation of some preliminary results for grillages of different spacing (s/t = 2.5, 4, 6 and 8) in loose siliceous sand. 2

EXPERIMENTAL METHODS

2.1 Introduction In the series of tests reported in this paper, the capacity of grillage foundations under purely vertical loading was investigated. The foundation capacities © 2011 by Taylor & Francis Group, LLC

Figure 2. Direct shear test results (Dr = 8.7%).

were compared to standard design methods for both conventional shallow foundations and piles. 2.2

Foundation properties

Tests were carried out under plane strain conditions in a box of 300 mm width and length 1000 mm with a soil sample of approximately 360 mm depth. Grillage foundations were constructed from steel plates each of which were of thickness, t = 5 mm or 10 mm, width, L = 300 mm and height, d = 150 mm. They were connected together with 6 bars and spacer blocks to ensure that the whole foundation was rigid and each grille was parallel (Fig. 1 and 3). Figure 3b shows a grillage foundation with eight 5 mm grilles with a centre-to-centre spacing, s = 40 mm. The connection system allowed the spacing, s between the grilles and the number of grilles, N to be varied between tests. The individual grilles and their spacing were similar to those used offshore. 2.3 The soil sample A uniform, dry fine silica sand was used in the tests. The sand had d50 = 0.18 mm, d10 = 0.12 mm, ρmax = 1760 kg/m3 and ρmin = 1460 kg/m3 . To prepare samples of known relative density a mesh of similar dimensions to the box was placed at the base of the box. Sand was placed in the box and then the base mesh was slowly extracted from the soil. This sheared the soil to critical state giving a loose condition. The density of the soil was measured and found to be uniform with ρ = 1487 kg/m3 .This gave a relative density, Dr = 8.7%. Direct shear tests conducted on samples prepared with Dr = 8.7% found that φ = 30.8◦ with ψ ≈ 0◦ . These test results are shown in Figure 2. 2.4 Apparatus The apparatus for vertical loading is shown schematically in Figure 3a and a photograph is shown in Figure 3b. The grillage foundation was attached to a hydraulic actuator via an S-shaped tensioncompression load cell to measure the vertical load on the foundation. The actuator had a stroke length of 300 mm and was actuated using a hydraulic pump. A linearly variable differential transformer (LVDT) was positioned to measure the vertical position of the

410

Table 1. Test programme.

Figure 3. Vertical grillage testing apparatus.

Test identifier

s (mm)

t (mm)

s/t

N

V16 V5 V18 V11

12.5 20 30 40

5 5 5 5

2.5 4 6 8

5 5 5 5

Figure 4. Load-displacement results for test V5: t = 5 mm; s = 20 mm (s/t = 4); N = 5 (B = 85 mm); L = 300 mm; Loose sand.

foundation. Image analysis was also used to measure foundation and soil movements. This required the front face of the box containing the sand sample to be constructed from Perspex to allow the soil and foundation to be photographed repeatedly with a digital camera during each test. Digital image analysis was later performed with the GeoPIV program (White et al., 2003).

are loose and the grille thickness, t = 5 mm in all tests. The grillage spacing ratio (s/t) is varied from 2.5–8 across the different tests (see Table 1).

2.5 Methodology

3.1

The grillage foundation was initially held several millimetres above the soil surface with the vertical actuator. The hydraulic pump was started and the actuator displaced the foundation vertically downwards at a velocity of approximately 0.4 mm/sec. Digital photographs were captured of the front face of the container throughout the test at the same time as the load cell and LVDT readings were recorded. Each test was stopped after approximately 60 mm of penetration of the foundation, just before the connector bars between the grilles came into contact with the soil. This amount of penetration ensured that data was available for z = 50 mm, the typical maximum length. In many tests, the foundation was then extracted from the soil to estimate the interface friction on the grilles.

The load-displacement results from a test with 5 grilles with thickness, t = 5 mm and spacing, s = 20 mm (so s/t = 4) in loose sand is shown in Figure 4. There is a clear increase in foundation capacity as the penetration, z of the grilles increases and no capacity before the grilles penetrate significantly. Also shown on Figure 4 are the results from separate calculations of foundation capacity assuming that either the foundation behaves as a solid, embedded plate (‘fully plugged solution’) or that each grille penetrates the soil and generates soil resistance as it was a plane strain pile (‘multiple pile solution’). For the solid foundation assumption it is assumed that the foundation is of breadth B (defined by the external dimensions of the grilles; Figure 1) and has embedment depth, z (i.e. the soil between the grilles move as if part of the foundation). The standard Terzaghi equation (1943) is used to calculate the bearing pressure at failure, qf :

2.6 Test programme In this paper four vertical tests are reported to investigate the effect of spacing ratio, s/t and penetration, z on bearing capacity of the grillage. Sand conditions © 2011 by Taylor & Francis Group, LLC

3

411

RESULTS Result for s/t = 4

where qf = Vo /A where the base area, A = BL, γ is the unit weight of the soil, c the apparent cohesion and Nγ , Nc and Nq are the bearing capacity factors which vary with the angle of friction, φ of the soil. Reissner (1924) found the closed form solution for Nq :

Hansen (1970) suggested empirically that:

The solid line with triangular markers (‘fully plugged solution’) on Figure 4 was produced combining Equations 1 to 3 and then also allowing for lateral friction on the outside of the external grilles (using the equations presented for the pile solution later). Parameters used were γ = 14.59 kN/m3 (from the measured ρ = 1487 kg/m3 ), c = 0 kPa and φ = 30.8◦ (from the direct shear test results, Figure 2). Figure 4 shows that the measured bearing capacity of the grillage foundation reaches that of the assumed solid, embedded foundation when z = 50 mm but then seems to slightly exceed this solution. Alternatively, a penetration, z = 23.7 mm is required to obtain the calculated flat plate bearing capacity for no embedment (Vo = 0.27 kN) which is likely to be the design value for a conventional mudmat. Thus, significant penetration of the foundation is required to generate the flat plate capacity, but there is additional capacity should the structure withstand additional foundation settlement (which may occur during installation when better tolerated). The dotted line on Figure 4 is the solution assuming that each grille acts as a single, plane strain pile. Therefore the capacity of each grille is obtained by summing the base resistance and the skin friction as done for axially loaded piles and the capacity of the whole foundation is obtained by multiplying the individual grille capacity by the number of grilles, N . The base capacity, Qb of each grille is given by a slightly modified version of the standard pile endbearing equation to allow for the base area (Ab = tL):

Figure 5. Load-displacement results for test V11: t = 5 mm; s = 40 mm (s/t = 8); N = 5 (B = 165 mm); L = 300 mm; Loose sand.

The dotted line on Figure 4 was calculated using the same parameters as for the shallow foundation assumption, but also using K = 1 (i.e. approximately twice K0 as appropriate for driven piles in sand, Kulhawy, 1984) and δ = φ − 5◦ = 26.4◦ . Figure 4 suggests that the pile equations are appropriate for very small penetrations, but that as the grille penetration increases, the foundation capacity increases more rapidly than given by the multiple pile solution with the parameters used. This increase might be because of enhanced silo pressures between the grilles which raise the normal stress on the interface (or effectively increase K) as in the case of plugged piles (e.g. Randolph et al., 1991). This is discussed in more detail by Bransby et al. (2009) who suggest an analytical solution for the bearing pressure, q before plugging for a large number of grilles:

where a = 2K tanδ/(s − t). The results using Equation 7 are shown on Figure 4 and give a better approximation to the experimental results, but require some improvement. 3.2

For a uniform soil layer, the skin friction on each grille of length, L and penetration depth, z becomes:

where K is the lateral earth pressure coefficient and δ is the angle of interface friction between the soil and the steel grille. Note that each grille has two sides and the sum of these two sides is included in Eq. 5. If it is assumed that each grille is unaffected by the others (i.e. assuming no ‘pile’group effect), the overall capacity of the foundation is:

© 2011 by Taylor & Francis Group, LLC

412

Result for s/t = 8

Figure 5 shows the equivalent results for a grillage foundation with a wider grille spacing (s/t = 8), but otherwise identical foundation and soil properties to the previous test. The simple analytical results for the rigid foundation and piled solutions are also shown on Figure 5 for comparison with the test results. For this wider grille spacing the measured load-penetration curve follows the pile solutions more closely. However, the foundation capacity is larger than predicted using the pile solution and this disparity increases with embedment. Again, the reason for the difference between the predicted pile solution and the measured foundation response could be due to grille-to-grille group effects or inter-grille arching/plugging effects

Figure 6. Load-displacement results in loose sand normalised by flat plate capacity (t = 5 mm; N = 5; L = 300 mm).

Figure 7. Penetration required to achieve equivalent flat plate capacity as a function of s/t; Loose sand, N = 5, t = 5 mm.

that increase the normal pressure acting on the grille interfaces and the grille tips. A large spacing ratio, s/t = 8 at first suggests that grille-to-grille interaction is unlikely, although it may be the ratio between the depth of soil between grilles, z, and the space between them (i.e. s − t) which is important for silo effects. At the end of the test, z/(s − t) = 1.7 and so the soil contained between each grille is much deeper than its width. Very significant foundation penetrations are likely to be required to mobilise the calculated equivalent flat plate capacity for zero embedment (i.e. the standard mudmat solution) for the spacing s/t = 8. Figure 5 shows that this displacement was not reached during the test at a penetration of 60 mm, but tentative extrapolation of the loading curve suggests that approximately 120 mm of penetration would be required. This may be a larger displacement than tolerated by the structure supported by the foundation, although working loads should not approach the vertical bearing capacity. 3.3 Result for different grillage spacings Figure 6 shows the load-displacement results from the two vertical tests already reported, along with two additional tests with grille spacings, s/t = 2.5 and s/t = 6. The load in Figure 6 is normalised by the capacity of an equivalent flat plate foundation of the same overall width at zero embedment (the standard ‘mudmat’ design capacity, Vo ). Figure 6 highlights the difference in foundation capacity with spacing ratio. When the plates are very closely spaced (e.g. s/t = 2.5 or 4) the grillage can provide the full flat plate capacity of an equivalent width conventional foundation, though some settlement is necessary to achieve this: 12 mm for s/t = 2.5 and 25 mm for s/t = 4. As the spacing increases still further (s/t = 6) the grillage is not quite able to mobilise the full equivalent flat plate capacity at the maximum penetration achieved in the test (60 mm). For s/t = 8 less than 50% of the equivalent flat plate capacity © 2011 by Taylor & Francis Group, LLC

413

has been achieved when the tests were terminated at a penetration = 60 mm. The penetration required to achieve the flat plate capacity has been plotted against s/t in Figure 7. For the case of s/t = 8, which does not achieve this capacity, an estimate of the penetration required has been made by extrapolating the load-penetration curve as discussed earlier. Figure 7 suggests that full flat plate capacity may be achieved by grillages with s/t ≤ 4 in loose sand if only 25 mm of settlement is tolerated. The maximum spacing ratio will be a function of the amount of structural settlement which is permitted. For example, if instead 60 mm is tolerated (dotted line on Fig. 7) a spacing of s/t = 6 may just be acceptable. Note, however, that Figure 7 is based on results for tests with only five grilles (N = 5) and so should not be relied upon for design until the effect of number of grilles has been studied in detail. Soil displacements in the foregoing tests were obtained by comparing the sequential digital images captured during each test.This was done using GeoPIV (White et al., 2003). Figure 8 shows vectors of accumulated soil displacements for the complete installation process (i.e. penetration of the grillage by 60 mm) for the tests discussed previously. The vectors in Figure 8 are all plotted at the same scale factor to facilitate direct comparison across the different tests, and the grillage has been added to scale for each test. Due to a malfunction with camera, there were no images available for test V5 (s/t = 4). There is clearly significantly more soil displacement occurring in Figure 8a (where s/t = 2.5) as a block of soil between the grilles is displaced vertically downwards and the soil displaces with a general shear failure mechanism somewhat similar to that of an embedded foundation. This explains why the bearing capacity approaches that calculated for an embedded foundation when s/t = 2.5; the inter-grille soil has ‘plugged’. In contrast, Figure 8c shows the displacements for the most widely spaced foundation investigated (s/t = 8). Much smaller soil displacements are

foundations under vertical loading only. Such foundations may be used for offshore infrastructure such as pipeline end manifolds, pipeline end terminations or temporary anchors to replace conventional foundations under appropriate load combinations. The preliminary results presented here have demonstrated that the bearing capacity of the grillages is strongly affected by the spacing of the individual grilles in loose sand. Closely spaced grilles will plug after moderately small penetrations and give the capacity of an equivalent flat plate (a conventional ‘mudmat’). In contrast, widely spaced grilles will not plug until excessive displacements which may not be tolerable to the structure. Selection of grille spacing therefore will depend on the levels of tolerable structure settlement as well as the soil characteristics. Results for only five grilles in loose sand were presented here. Consequently, the results of this paper should not yet be used for design without further work. Further investigation is required to quantify performance in different soil types or densities and for more realistic numbers of grilles. In addition, more work is required to investigate their capacity under combined (V-H-M) loading. ACKNOWLEDGEMENTS The work conducted has been funded by Acergy UK, Subsea 7 and Technip. The authors wish to thank their respective companies for permission to publish this paper. The views expressed are those of the authors alone and do not necessarily represent the views of their respective companies. REFERENCES

Figure 8. Vectors of soil displacement at failure for vertical tests in loose sand.

observed for the same grille penetration as the grilles penetrate the soil and just push away the volume of the grilles themselves (as during pile jacking); the grilles have not plugged. In summary, the soil displacement fields confirm the earlier findings. Plugging occurred for the most closely spaced foundation (s/t = 2.5), but not for the more widely spaced ones (s/t = 6, 8). Figure 4 suggests that if images had been available for the test with s/t = 4, the onset of plugging may have been observed in the final images of the sequence. 4

CONCLUSIONS

The paper reports a series of laboratory model tests investigating the bearing capacity of grillage © 2011 by Taylor & Francis Group, LLC

Bransby, M.F., Knappett, J.A., Hudacsek, P. and Brown, M.J. (2009). The vertical capacity of grillage foundations. Submitted to Geotechnique, October 2009. Georgiadis, M. (1985). Load-path dependent stability of shallow footings. Soils and Foundations, 25(1), 84–88. Gottardi, G. and Butterfield, R. (1993). On the bearing capacity of surface footings on sand under general planar loads. Soils and Foundations, 33(3), 68–79. Kulhawy, F.H. (1984). Limiting tip and side resistance, fact or fallacy. Symposium on Analysis and Design of Piled Foundations, ASCE, San Francisco, 80–98. Nova, R. and Montrassio, L. (1991). Settlements of shallow foundations on sand. Géotechnique 41(2), 243–256. Randolph, M.F., Leong, E.C. and Houlsby, G.T. (1991). One-dimensional analysis of soil plugs in pipe piles. Géotechnique 41(2), pp.587–598. Reissner, H. (1924). Zum Erddruckproblem. Proc. 1st Int. Conf. Appl. Mech., Delft, 295–311. Terzaghi, K. (1943). Theoretical soil mechanics. New York: Wiley. White D.J., Take W.A. and Bolton M.D. (2003). Soil deformation measurement using particle image velocimetry and photogrammetry. Géotechnique, 53(7), 619–631. Zaharescu, E. (1961). Sur la stabilité des fondations rigides. Proc. 5th Int Conf. Soil Mech., Paris, 1, 867–871.

414

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Behaviour of skirted footings on sand overlying clay C.T. Gan, K.L. Teh, C.F. Leung, Y.K. Chow & S. Swee Centre for Offshore Research and Engineering, National University of Singapore

ABSTRACT: On sand overlying soft clay, the bearing resistance developed in the upper soil layer could be substantially larger than that in the lower layer. Installing an offshore foundation on this kind of soil condition could possibly bring about punch-through risk when the imposed load exceeds the upper soil bearing resistance. Intuitively, the punch-through risk could be reduced by minimizing the bearing resistance variance between the upper and lower layers by either reducing the bearing resistance of the upper soil or increasing bearing resistance of the lower soil. The former case is explored in this paper by adopting a footing with extended skirt denoted as skirted footing. This paper reports the findings obtained from a series of centrifuge model tests to investigate the effect of skirt height on the development of bearing resistance of skirted footings in sand overlying clay. The findings suggest that skirted footings produce a lower peak bearing resistance and punch-through depth as compared to spudcan foundation. The footing with skirt height equaling the sand thickness appears to greatly reduce the risk of punch-through failure. 1

INTRODUCTION

The foundation of jack-ups is not custom-designed for a specific site. Therefore, they must be designed to remain stable, regardless of the soil conditions (Poulos, 1988). When a spudcan is installed in strong soil layer overlying weak soil layer such as sand overlying clay where there is a potential of reduction in bearing resistance as a spudcan penetrates, punch-through risk exists. Punch-through happens when the spudcan penetrates uncontrollably through the strong layer into the underlying weak layer, resulting in excessive tilting of the structure. The hull tilting condition can be worsened by additional distributed weight due to shifting of the centre of gravity of the hull towards the punchedthrough leg. A punch-through may subject the leg(s) to very large stresses and cause significant damages to the structure and equipment onboard. This is likely to cause heavy losses in terms of time and cost. A point to note is that punch-through has been occurring at an alarming rate of once per year (Osborne & Paisley 2002). This indicates the needs for improved procedures in identifying and managing such risk. While with detailed site surveys of the area and good quality soil data the punch-through risk can be identified and a safe rig installation requires the risk to be adequately mitigated. Intuitively, this can be achieved by sufficiently reducing the bearing resistance of the upper layer or increasing the bearing resistance of the lower layer such that the difference of bearing resistance between the layers is minimum or reduced to a level where the anticipated spudcan punch-through behaviour is deemed manageable. A limited number of approaches are available to mitigate punch-through risk. Perforation drilling

© 2011 by Taylor & Francis Group, LLC

is a ground preparation method conducted prior to rig installation which involves (partially) removal of strong layer and hence reduces the bearing resistance of the stronger layer. This method has been implemented at some punch-through identified sites with varying degree of success. However, this method is rather time-consuming. Chan et al. (2008) reported a perforation drilling operation conducted in offshore Southeast Asia which took up to 171 hours to drill 106 perforations of 0.91 m diameter and 28 m deep, and only two-thirds of the perforations were successfully drilled. To implement the method, it also requires a sufficient operation time in advanced of the rig installation. There is another form of mitigation which involves no modification to the ground condition but incorporating some cautious procedures during rig installation. According to Rapoport and Young (1987), the procedures include: • Using a small air gap so that the vertical displace-

ment during punch-through failure can be reduced as the leg load is partially countered by the hull buoyancy as it penetrates the water line. • Preload in water so that the leg loads are reduced during penetration. • Preloading one leg at a time. While these procedures are generally found effective in mitigating the punch-through risk, incorporating these procedures inevitably increase the installation time. Also, there are certain conditions where these procedures may not be applicable or suitable. For instance, preloading one leg at a time may lead to intolerance RPDs and the correction procedure will have to perform to ensure that the limit is not exceeded.

415

The feasibility of adopting skirted footings, which are flat footings/spudcans equipped with skirts along their circumferences, to mitigate the punch-through effects was conceptualized by Svano & Tjelta (1993). As compared to a spudcan, a skirted footing penetrating in single layer soil would develop a different resistance profile with relatively low resistance during penetration of skirt and followed by sudden a sudden increase in resistance when the flat base is making contact with the ground surface. This trend of resistance profile has been confirmed by test results reported by Byrne & Cassidy (2002) in clay and Teh et al. (2006) in sand. Following the idea of Svano & Tjelta (1993), if the skirt can be designed such that the penetration resistance of the skirt alone is less than the available force during jacking, that is before the hull is lifted free from the sea, the punch-through risk could be greatly reduced. Some preliminary centrifuge test results were reported by Teh et al. (2008) in which the penetration responses of a model spudcan and a skirted footing in medium dense sand overlying soft clay were compared. It should be noted that the skirted footing had a skirt height that was shorter than the upper sand layer thickness. The test results revealed that the skirted footing developed a lower magnitude of peak bearing resistance in the upper layer and the measured punch-through distance, dP-T (as defined in Fig. 2) was shorter, implying a more controllable penetration response as compared to that of the spudcan. Assuming that to a large extent, the skirted footing penetration response in strong soil overlying soft soil is governed by its skirt height or more precisely the ratio of the skirt height to the thickness of the upper layer. Following this, the effectiveness of using skirted footing to mitigate punch-through risk will therefore be affected by its skirt height. This aspect deserves investigation. A series of centrifuge tests was conducted by modeling the penetration of skirted footings with different skirt heights into sand overlying clay. In this paper, the set-up and procedures of the tests are described in the following before the test results are presented. Finally, the practical issues concerning the use of skirted footing particularly the selection of skirt height are discussed. 2

EXPERIMENTAL SET-UP AND MATERIAL PROPERTIES

The centrifuge model tests presented in this paper were conducted on the beam centrifuge at National University of Singapore (NUS).This is a 2 m radius centrifuge that is designed for a payload capacity of 40g-tonnes and can spin-up to a maximum acceleration of 200 g. A stack of 100 tracks silver-graphite slip rings is mounted on top of rotor shaft for power and signal transmission between the centrifuge machine and the control room. Detailed information about NUS centrifuge can be found in Lee et al. (1991) and Lee (1992). © 2011 by Taylor & Francis Group, LLC

Table 1. 2003).

Properties of Malaysia kaolin clay (after Goh

Parameter

Value

Liquid limit Plastic limit Specific gravity Coefficient of consolidation (at 100 kPa) Coefficient of permeability (at 100 kPa)

80% 35% 2.60 40 m2 /yr 2.0 × 10−8 m/s

Modified Cam Clay parameters: Critical state frictional constant Slope for normal consolidation line Slope for swelling line

0.9 0.244 0.053

Table 2.

Properties of Toyoura Sand (after Teh 2007).

Parameter

Value

Specific gravity Uniformity coefficient Average particle size Dmax Dmin D50 D10 Range of density 1 Critical state friction angle, φcv

2.65 1.3 0.2 mm 0.3 mm 0.115 mm 0.2 mm 0.163 mm 1335–1645 kg/m3 32◦

1

Jamiolkowski et al. (2003)

Malaysia kaolin clay was used in this study. The engineering properties of the clay are shown in Table 1. Firstly, the dry kaolin powder was mixed with water at a water content of 1.5 times its liquid limit which is 120% of the clay powder weight. They were mixed thoroughly in a large air-tight mixer with a constant suction applied for 4 hours to form fully saturated clay slurry. Prior to the clay slurry pouring, a 30 mm thick sand layer that serves as drainage layer was placed at the base of a 550 mm diameter model container. A layer of silicon grease was applied on the inner wall of the model container to reduce the friction between the sidewall and the soil. The trapping of air pockets were minimized as the clay slurry was placed under water. To form a relatively firm clay surface for sand placement, the clay slurry was then subjected to a gradually increasing pre-consolidation pressure from 0.5 kPa to 15 kPa at 1 g. The overall pre-consolidation process took 7 days to complete. Next, the surface water was removed before the placement of sand. The type of sand used for the experiment is Toyoura sand. The engineering properties of the sand are shown in Table 2. The sand was prepared following the method adopted by Teh (2007) which is a spot-type sand raining process with proper control of sand drop height. The relationship between the relative density of the sand layer formed and the sand drop height has been established by researchers from the same research team using the same apparatus (Eio 2003 and Teh 2007). In the present tests, the relationship was re-calibrated by conducting a number of trial

416

sand raining tests. It is established that raining at a drop height of 600 mm consistently produces a dense sand layer with relative density of 95%. This drop height was adopted in the preparation of the upper sand layer. A 54 mm thick sand layer was formed on top of the earlier prepared clay sample. The sand was then saturated by creating a controlled differential hydraulic head condition. The soil sample was kept fully saturated with a layer of free water layer maintained throughout the consolidation and testing processes. The soil sample was then moved onto the platform of the centrifuge and subjected to self-weight consolidation at 100 g for at least 7 hours. At the end of this consolidation process, the underlying clay sample is believed to have achieved at least 90% degree of consolidation determined based on the surface settlement of similar soil sample monitored previously by Teh (2007). The spudcan/skirted footing was jacked into the model ground at a penetration rate of 0.5 mm/s that results in drained condition in sand and undrained in clay following the velocity group parameter proposed by Finnie (1993). 3

RESULTS AND DISCUSSIONS

Unless otherwise stated, all experiment results presented in this paper are presented in either normalised form or prototype dimension scaled according to the scaling laws published by Garnier et al. (2007). A total of four tests were performed using different model foundations which include a model spudcan and three model skirted footings with different skirt heights, L, of 0.5, 1.0 and 1.5 times the sand layer thickness, Hs , respectively All the model foundations have the same prototype diameter of 8 m. Two tests were performed in a sample. For each model skirted footing, four 1 mm (model scale) diameter holes were drilled at the flat base of the footing to facilitate the water draining. Figure 1 shows the dimensions and load reference points, L.R.P, of the model foundations. The summary of test details is presented in Table 3. The underlying clay layer was normally consolidated under the sand overburden as well as soil self-weight. Hence, the undrained shear strength, su , profile is expected to increase linearly with depth. Although miniature ball penetrometer tests were conducted with an aim to characterize the su profile of the underlying clay, the ball penetration response appears to be questionable with inconsistent development of resistance with depth. Similar anomaly has been observed by Teh (2007) and Lee (2009) which is generally attributed to the unstable state of the soils trapped underneath the ball as the ball penetrate continuously from the upper sand layer into the lower clay. The su profile is therefore approximated using the modified cam clay framework proposed by Roscoe & Burland (1968) and it is expressed as follows:

where z is the depth beneath the sand-clay interface, in meters and su in kPa. Equation 1 generally agrees © 2011 by Taylor & Francis Group, LLC

Figure 1. Schematic diagram showing dimensions and load reference points (L.R.P.) of spudcan and skirted footings (all dimensions are in mm, model scale). Table 3.

Summary of test details.

Test

Skirted height, L m

1

Spudcan Skirted footing 1 Skirted footing 2 Skirted footing 3

0 2.7 5.4 8.1

0 0.5 1.0 1.5

L/Hs

Note: 1 Hs = 5.4 m (prototype scale)

with the strength profiles reported by Teh (2007) who used similar materials in preparing the test samples. 3.1

Spudcan versus skirted footing

The penetration resistance profiles of the model spudcan and skirted footing 1 (L/Hs = 0.5) are compared in Figure 2. The penetration resistance is calculated by dividing the measured load by the widest crosssectional area of the footing and penetration depth, d, is measured from the model ground surface to the L.R.P of the footing. For the spudcan test, the penetration resistance increases with depth and reaches a peak resistance, qpeak , of 338 kPa at d of 0.65 m, followed by a sudden loss in resistance before resistance of similar magnitude as qpeak is developed at 5.6 m. In this paper, the difference in elevation between qpeak and the ‘regained qpeak ’ is defined as punch-through distance, dP-T , as indicated in Figure 2. For profile

417

the footing resistance may be still subjected to softening effect from the underlying clay. This explains the slight drop in penetration resistance of skirted footing 1 from 3.0 m to 3.9 m (Fig. 2). While the mechanisms discussed above seem to explain the test results adequately, they are postulated to certain extent. Further investigations on the failure mechanism shall be conducted in the future to verify the postulated mechanisms. 3.2

Figure 2. Penetration behaviour of spudcan and skirted footing 1 in sand overlying clay.

which shows dP-T > 0, which is observed in the spudcan penetration resistance profile, and the qpeak falls between the range of the applied loads i.e. qpeak is greater than the lightship weight but smaller than the preload, punch-through risk should be identified. On the other hand, the penetration resistance of skirted footing 1 increases gradually as the skirt penetrates into the soil. At d = 2.95 m i.e. some distance after the footing base made full contact with the ground, the qpeak is developed but with a smaller magnitude of 250 kPa as compared to that of spudcan. The corresponding dP-T is also shorter. An explanation is offered in the following. By drawing analogy to an open tube pile, the q before the footing base is fully in contact with the sand, is the sum of inner and outer skirt wall frictions, and end bearing on skirt annulus (Houlsby & Byrne 2005a, b). When the flat base of the skirted footing is fully in contact with the ground, the skirt chamber is likely to be filled by sand. After a short distance of penetration which allows the trapped sand to undergo some degree of compaction, the skirted footing and trapped sand may move like a rigid block as the footing advances. At this point onwards, the skirted footing may be analogous to a cylindrical footing with foundation base defined at the level of skirt tip. The q is therefore contributed by the outer skirt wall friction and end bearing mobilised across the footing widest cross-section area at the foundation base level. The thickness of the sand layer where the ‘cylindrical footing’ is founded should be viewed in term of effective value, H
s , which may be approximated as (Hs − L). For instance, for skirted footing 1, the H
s is 2.7 m (see Fig. 1b). When H
s > 0, the end bearing component of © 2011 by Taylor & Francis Group, LLC

Effect of skirt height

Figure 3 shows the penetration resistance profiles of three skirted footing of different skirt heights (refer to Table 3 for test details). Fundamentally, the bearing capacity of a foundation in sand overlying clay is contributed by resistances in both the upper and lower layers in which the correct proportion of these resistance components is partly determined by the relative distance of the foundation base to the underlying clay. For a skirted footing behaving like a ‘cylindrical footing’i.e. composite block of skirted footing and trapped sand, such distance is essentially H
s . For skirted footing 1, H
s is greater than zero. The ‘cylindrical footing’ is viewed as founded on sand overlying clay system and the footing resistance ia expected to be affected by the softening effect due to the underlying clay. Therefore, the penetration profile registers a qpeak and followed by post-peak reduction. The degree of post-peak reduction should be dependent on H
s in which intuitively a higher H
s (i.e. shorter skirt) leads to a higher post-peak reduction. For skirted footing 2 in which the skirt height is equal to the sand thickness (Hs = L), the H
s is essentially zero. During the formation of ‘cylindrical footing’ mechanism, the base of the ‘cylindrical footing’ reaches the sand-clay interface. The footing is likely to be loaded directly on the underlying clay with the upper sand layer serving as overburden. Following this argument, one would not expect the peak and postpeak responses to occur. However, it is worth noting that even when H
s = 0, there could be a possibility of developing peak and post-peak responses during the transition from the ‘open tube pile’to ‘cylindrical footing’ mechanisms if the total resistance of the former mechanism is larger than that of the latter mechanism. The resistance components mobilised by each failure mechanism have been discussed in Section 3.1. When the skirted footing continues to penetrate into the clay layer, there ought to be some progressive reductions in side friction along the outer skirt wall as the shear interface changes from sand-wall to clay-wall. However, this is countered by the increasing overburden pressure and the mobilized shear strength of the underlying clay that yields a higher end bearing capacity with depth. The test results of skirted footing 1 and 2 imply that the latter gain outweighs the former reduction, resulting in a smoothly increasing penetration resistance. Skirted footing 3 represents a case with H
s < 0 in which the skirt height is longer than the sand layer

418

two skirted footings, it still performs better than the spudcan.

3.3

Figure 3. Penetration behaviour of skirted footings of different skirt heights in sand overlying clay.

thickness. Interestingly, the measured profile indicates a post-peak reduction. Further, the post-peak reduction takes place before the flat base of the footing is in contact with the ground surface but the skirt tip has penetrated into the clay layer. The mechanism associated to the ‘cylindrical footing’ as described earlier is not yet formed. So the possible explanations for this interesting penetration response should concern only the ‘open tube pile’ mechanism. As mentioned earlier, the total resistance developed under this failure mechanism is the sum of inner and outer skirt wall frictions, and end bearing on skirt annulus. As the skirt advances, there is an increase in side frictions due to a greater skirt wall surface in contact with soils (solely sand before the skirt tip penetrates beyond the sand-clay interface). It is noted that the friction increment is much greater in sand than in clay. However, the shear strength of the sand particularly dense sand is susceptible to large strain reduction, suggesting that the actual increase in friction would not be proportional to the increase in wall surface area. On the other hand, there is a relatively significant reduction in end bearing as the skirt annulus is shifting from in uniform sand to sand-over-clay and later to soft clay. The results shown in Figure 3 imply that the loss in the end bearing outweighs the gain in the side friction. The skirted footing 3 regains the penetration resistance shortly after the flat base fully touches the model ground at around d of 8.8 m (see Fig. 3). It may be worth mentioning that the qpeak and dP-T recorded by skirted footing 3 are smaller than that of spudcan. This suggests that although the performance of skirted footing 3 is less satisfactory as compared to the other © 2011 by Taylor & Francis Group, LLC

Practical issues

The findings obtained from the present study suggest that when a footing equipped with skirt penetrates in sand overlying clay, it mobilises a lower qpeak and experience a smaller dP-T The findings also show that the height of the skirt affects the performance of the skirted footing in which the ‘optimum’ skirt height appears to be equal to the sand layer thickness. Ironically, as pointed out in Section 1 that the foundation of jack-up rigs is not custom-designed for a specific site and the skirt height is normally designed to be less than a quarter of the foundation diameter to ensure smooth rig towing (Purwana 2010, pers. comm.). As the upper sand layer thickness varies with region and area, the optimum performance of skirted footings in mitigating punch-through may not be achieved. Another practical issue is the installation of skirted footings in multi-layered soil particularly a soil condition where a sand layer is sandwiched in between two soft clay layers. The inner space of the skirt may be partially/fully filled up with the upper soft clay before penetrating through the sand layer, and hence altering the soil failure mechanism. In other words, the effectiveness of the skirt may be reduced. This topic is currently being investigated by the NUS research team. Owing to limitations in model fabrication, the skirt thickness modeled in the current study in prototype scale is 0.3 m (note that it is only 3 mm in model scale). Such dimension exceeds the typical dimension adopted in the field. Furthermore, stiffeners that are employed to increase the skirt integrity were not modeled in the present study. These differences may affect the performance of the skirted footing and hence the issues deserve further investigation.

4

CONCLUSIONS

This paper presents a series of centrifuge model tests on spudcan and skirted footings penetration on sand overlying clay. The results suggest that the skirted footings provide a more steady penetration response with reduced peak bearing resistance and punch-through distance when the peak bearing resistance is exceeded. The former enables the footing to penetrate through the strong layer more easily and the latter shortens the leg uncontrollable penetration distance. The post-peak reduction is essentially eliminated in a test featuring a skirted footing with skirt height equal to the sand thickness. However, the optimum use of skirted footing for jack-up rig installation in sand overlying clay depends on various factors such as the sand thickness in relation to the skirt height, the ratio of maximum footing preload to peak bearing resistance, clay shear strength and soil stratification.

419

ACKNOWLEDGEMENT The authors wish to acknowledge the officer of the Geotechnical Centrifuge Laboratory, National University of Singapore, Mr. Wong Chew Yuen, for his valuable assistance in conducting the centrifuge model tests.

REFERENCES Byrne, B.W. & Cassidy, M.S. (2002). Investigating the response of offshore foundations in soft clay soils. Proc. 21st Int. Conf. Offshore Mech. And Artic Eng., Oslo, OMAE2002-28057. Chan, N.H.C., Paisley, J.M. & Holloway, G.L. (2008). Characterization of soils affected by rig emplacement and swiss-cheese operations – Natuna Sea, Indonesia, a case study. 2nd Jack-up Asia Conference, Singapore. Eio, T. L. 2003, Sand preparation for geotechnical model test. BEng thesis, National University of Singapore, Singapore. Finnie, I.W.S. (1993). Performance of shallow foundations in calcareous soils. PhD thesis, University of Western Australia. Garnier, J., Gaudin, C., Springman, S.M., Culligan, P.J., Goodings, D., Konig, D., Kutter, B., Phillips, R., Randolph, M.R. & Thorel, L (2007). Catalogue of scaling laws and similitude questions in geotechnical centrifuge modelling. International Journal of Physical Modelling in Geotechnics. 8(3), 1–23. Goh, T.L. (2003). Stabilisation of an excavation by an embedded improved soil layer. PhD thesis, National University of Singapore. Houlsby, G.T. & Byrne, B.W. (2005a). Design procedures for installation of suction caissons in clay and other materials. Proc. of the Institution of Civil Engineers-Geotechnical Engineering, 158, No. 2, 75–82. Houlsby, G.T. & Byrne, B.W. (2005b). Design procedures for installation of suction caissons in sand. Proc. of the Institution of Civil Engineers-Geotechnical Engineering, 158, No. 3, 135–144.

© 2011 by Taylor & Francis Group, LLC

Lee, F.H. (1992). The University of Singapore Geotechnical Centrifuge User Manual. Research Report No. CE001. July 1992. Lee, F.H., Tan, T.S.,Yong, K.Y., Karunaratne, G.P. & Lee, S.L. (1991). Development of geotechnical centrifuge facility at the National University of Singapore. Proc. Int. Conf. Centrifuge 1991, pp. 11–17. Jamiolkowski, M.B., Lo Presti, D.C.F. & Manassero, M. Evaluation of relative density and shear strength of sand from cone penetration test (CPT) and flat dilatometer (DMT). In Soil behaviour and soft ground construction, ed by J.T Germaine, T.C. Sheahan and R.V. Whitman, ASCE, GSP 119, pp. 201–238. Lee, K.K. (2009). Investigation of potential spudcan punchthrough failure on sand overlying clay soils. PhD thesis, University of Western Australia. Osborne, J. & Paisley, J. 2002. South East Asia Jackup Punch-throughs: the way forward? Proceedings of the Conference on Offshore Site Investigation and Geotechnics – Sustainability and Diversity, London. Poulos, H.G. (1988). Marine geotechnics. London: Unwin Hyman. Rapoport, V. & Young, A.G. (1987). Foundation Performance of Jack-Up Drilling Units,Analysis of Case Histories. International Conference on Mobile Offshore Structures. Roscoe, K.H. & Burland, J.B. (1968). On the generalized stress-strain behaviour of wet clay. Engineering Plasticity, Eds. J. Heyman, F.A. Leckie, Cambridge Univ. Press. 535– 609. Svano, G. & Tjelta T. I. (1996). Skirted spud-cans – Extending operational depth and improving performance. Marine Structures 9(1 Spec. Iss.): 129–148. Teh, K.L. (2007). Punch-through of spudcan foundation in sand overlying clay. PhD thesis, National University of Singapore. Teh, K.L., Byrne, B.W. & Houlsby, G.T. (2006). Effects of seabed irregularities on loads developed in legs of jack-up unit. 1st Jack-up Asia Conference, Singapore. Teh, K.L., Gan, C.T., Hu, E.H.J., Leung, C.F. & Chow, Y.K. (2008). A comparison of penetration behaviour of spudcan and skirted footing in sand overlying clay. 2nd Jack-up Asia Conference, Singapore.

420

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Numerical study of piping limits for suction installation of offshore skirted foundations and anchors in layered sand L.B. Ibsen Aalborg University, Denmark

C.L. Thilsted Dong Energy Power, Denmark

ABSTRACT: Skirted foundations and anchors have proved to be competitive solutions for various types of fixed offshore platforms, subsea systems and an attractive foundation alternative for offshore wind turbines. One main design challenge for skirted structures in sand is to penetrate the skirted deep enough to obtain the required capacity. In order to overcome the high penetration resistance in sand suction assisted penetration is needed. Suction installation may cause the formation of piping channels, which break down the hydraulic seal and prevent further installation. This paper presents a numerical study of failure limits during suction installation in respect to both homogenous and layered soil profile. A numerical flow analysis is performed to determine the hydraulic gradients developing in response to the suction applied, and the results are presented as simple closed form solutions useful for evaluation of suction thresholds against piping. These closed form solutions are compared with large scale tests, performed in a natural seabed at a test site in Frederikshavn, Denmark. These solutions are also valid for penetration studies of other offshore skirted foundations and anchors using suction assisted penetration in homogeneous or layered sand. Due to the complexity of the domain and the governing differential equation, the problem is solved numerically. A numerical solution can be obtained using either finite difference or finite element methods. In this paper, the problem is solved using the commercial finite difference program FLAC3D (Itasca, 2005).

1

INTRODUCTION

More than 485 suction anchors, had been installed for anchoring floaters at more than 50 different sites by the year 2004 (Andersen et al. 2005). Most of these anchors are in clay, but some are also in sand or layered soils. Examples of skirted foundations in sand are the offshore steel platforms at Draupner E and Sleipner T sites in the North Sea (Tjelta 1995). Skirted foundations in sand can also be used to increase the moment fixity and can be an attractive foundation alternative for offshore wind turbine as the bucket foundation installed in Frederikshavn has shown. (Ibsen 2008). In order to overcome the high penetration resistance in sand, suction assisted penetration is needed. The suction creates a pressure differential across the caisson lid, effectively increasing the downward force on the caisson while reducing the skirt tip resistance. This study has been a part of a research project whose aim is to develop a skirted foundation often referred to as the “bucket foundation” as a foundation for offshore wind turbines. At the time of writing, two bucket foundation have been installed, one at Horns Rev II and the other located in Frederikshavn, Denmark, (Ibsen 2008). Figure 1 shows an installation test of a 4 × 4 m bucket at the test site in Frederikshavn. © 2011 by Taylor & Francis Group, LLC

Figure 1. Installation tests on 4 × 4 m buckets in a natural seabed at the test site in Frederikshavn, Denmark.

The installation of bucket foundation for offshore wind turbines differs for several reasons. Compared to oil and gas jackets, the bucket foundation offers less self-weight to assist penetration and the offshore parks are predominantly located at shallow waters, <30 m, which reduces the maximum available suction capacity.

421

Figure 2. Definition of dimensions. Figure 3. The critical suction has been achieved and soil failure by piping has occurred. The test was performed with a 4 × 4 m bucket.

The bucket foundation is a large diameter moment resistant structure and its cost efficiency is significantly improved by increasing the ratio of skirt length L over diameter D to approximately L/D ≈ 1 while the wall thickness t is kept at a minimum. The geometric definitions are shown in Figure 2. This paper presents a numerical study of the installation of large diameter thin-walled suction caissons in sand. The objective is to evaluate suction failure limits during installation in respect to piping in both homogeneous and layered sand. Steady-state flow analyses were performed to determine the flow and the hydraulic gradients developed in response to applied suction beneath the caisson lid. The results are presented as simple closed form solutions, valid for a wide range of boundary conditions, and useful for evaluation of suction thresholds against piping in homogeneous or layered sand. These closed form solutions are compared with a large scale field test, installed in a natural seabed at the test site in Frederikshavn, Denmark.

2

Figure 4. CPT test performed prior to the installation of the buckets.

FIELD TEST DATA

Since installation data from field installation of suction caissons in sand are limited this project has conducted a substantial amount of installation tests on 2 × 2 m and 4 × 4 m buckets which have been performed at the offshore test site in Frederikshavn, Denmark, Ibsen (2008). One of the focus points for these installation tests has been to study the critical suction causing piping. Failure during suction assisted installation occurs when certain thresholds are exceeded. The failure may be caused by either formation of piping channels or cavitation of pore water. The formation of piping channels occurs when the applied suction increases and causes an upward flow, reducing the effective stresses within the caisson, and eventually liquefying parts of the internal soil matrix. Local piping channels break down the hydraulic seal and prevent further installation, as shown in Figure 3. © 2011 by Taylor & Francis Group, LLC

Three installation tests are studied in this paper. They are all installed within a test site of 13 × 14 m. The results of the CPTs performed prior to the installation are shown in Figure 4. The applied suction p needed to install the 2 × 2 m buckets can be seen in Figure 5. In the figure, the normalized suction p/γ
is plotted against the normalized penetration depth h/D where γ
is the submerged unit weight of the soil and D is the diameter of the bucket. In the figure 5 it is also seen that installation failure by piping did occur during the installation of bucket 4, at a depth 1.56 m. The piping channels were filled with sand and the outer soil surface leveled, in order to restart the installation. A new failure occurred at a depth of 1.7 m and the test was stopped. Figure 5 shows that the suction needed to install bucket 5 is higher than the suction resulting in piping

422

Figure 6. Schematic illustration of the axisymmetric flow do-main during suction installation.

Figure 5. The applied suction p needed to install the three buckets. The diameter and skirt length is 2 m × 2 m.

during the installation of bucket 4. The only difference between the installation tests is the presence of silt layers, see Figure 4. • Bucket 2 is installed in a homogeneous sand layer. • Bucket 4 is installed where one thin silt layer is

present at a depth of 2.7 m. • Bucket 5 is installed in a layered soil profile with

thin silt layers at depth of 1.2, 2.4 and 3.5 m. It is assumed that these thin silt layers act as impermeable flow boundaries and change the steady-state flow field around the skirt tip as it approaches the layer. The theory is that the presence of these impermeable flow boundaries will increase the suction thresholds against piping as it was observed from the installation test with Bucket 5. The influence of the flow boundary is modeled and studied in the following sections.

3

The term (1/r2 )∂2 u/∂φ2 vanishes due to the axissymmetry of the caisson. The differential equation must be solved with appropriate boundary conditions to determine the hydraulic gradient field which arises from the pressure difference, between the ambient seabed water pressure, γw hw + pa and the pore pressure beneath the lid, γw hw + pa + p. pa is the atmospheric pressure. Due to the complexity of the domain and the governing differential equation, the problem is solved numerically. A numerical solution can be obtained using either finite difference or finite element methods. In this paper, the problem is solved using the commercial finite difference program FLAC3D An axisymmetric model was created with a grid consisting of a total of 5,904 zones and an outer boundary located, in the distance, 20R the caisson, as shown in Figure 6. The case where L → ∞ is simulated as L = 20R. The boundary conditions along the caisson skirt, the bottom boundary and the axisymmetric axis are Neumann’s conditions, preventing a flow orthogonal to the boundary. The boundary conditions of the soil surface in the caisson, the free surface and the outer boundary are Dirichlet conditions with prescribed pore pressures. The steady-state flow model computes the exit hydraulic gradient i next to the skirt and that gradient is used to calculate the seepage length s in terms of the applied suction p as:

NUMERICAL MODEL

The pumping action results in the suction p inside the bucket, which then causes a steady-state flow field to evolve in the soil, as shown in Figure 6. This yields a constant influx of water, which must be pumped out to maintain a constant level of suction. Assuming isotropy the seepage problem reduces to the well-known Laplace’s differential equation, ∇ 2 h = 0. It may conveniently be expressed in terms of pore pressure, u = γw h and cylindrical coordinates (r, z, φ) due to the circular geometry of a suction caisson:

© 2011 by Taylor & Francis Group, LLC

The normalized seepage length s/h is a unique function of the relative penetration length h/D.

4

NUMERICAL RESULTS

The steady-state flow simulations were conducted for two different cases at various embedment depths 0.1D > h > 1.2D. In the first case, simulations were conducted to investigate bucket installation in homogeneous soil, the results are shown in Figure 7a. The second case simulates a bucket installed in sand over a impermeable flow boundary, located in the depth L . The results are shown in Figure 7b.

423

Figure 7. The results of the FLAC calculation are plotted as normalized seepage length for exit gradient versus relative penetration. a) Installation in homogenous sand. b) Installation in sand over a flow boundary.

4.1

Installation in homogeneous sand

The following empirical expression is given to approximate the numerical data for the installation in homogeneous sand.

Equation (3) is fitted to two boundaries. For a very small h/D ratio, equation (3) approaches 2.86, a theoretical solution for a sheet-pile wall, suggested by Hansen (1978). For an infinitely long bucket, all the hydraulic head loss occurs inside the bucket with evenly spaced horizontal equipotential lines. Therefore, the normalized length tends to unity. For installation in homogenous sand the internal hydraulic gradients have been investigated by several researchers using finite element programs as Plaxis and SEEP. Senders & Randolph (2009) performed calculations with the finite element programme Plaxis and proposed a similar expression for the exit gradient:

Figure 8. Seepage length for exit gradient versus relative pene-tration predicted by equation (4), (5) and (6).

Figure 8 show that these three different formulations predict similar seepage length for penetrations of practical interest 0.1 ≤ h/D ≤ 1. 4.2

For very small h/D ratio equation (4) approaches π, which is a theoretical solution for a sheet-pile wall, suggested by Scott (1963). Feld (2001) performed calculations with the finite element program SEEP and proposed that the seepage length could be estimated as:

Installation in sand over a flow boundary

The following empirical expression is given to approximate the numerical data for the installation in layered sand:

where (s/h)ref is calculated from equation (3). It is seen that equation (6) approaches equation (3) if © 2011 by Taylor & Francis Group, LLC

424

the distance to the flow boundary L is large in comparison to the diameter of the bucket D. 5

CRITICAL SUCTION

The formation of local piping channels occurs when the exit hydraulic gradient, next to the caisson wall, exceeds the gravitational force, and thereby reduces the effective stresses to zero. The critical gradient is:

The exit hydraulic gradient i can also be expressed in terms of the applied suction p and the seepage length s as: Figure 9. Normalized critical suction versus relative penetration. The critical suction is calculated with different ratios L /D.

where γw is the unit weight of water and γ’ is the submerged unit weight. The critical suction resulting in formation of local piping channels are therefore

By combining equation (6) with equation (9) the critical suction can be expressed as:

bucket 5 the flow boundary was at a depth of 1.2 m. This increases the suction capacity and the bucket was penetrated with the highest applied suction without any failure occurring. It is shown that these thin silt layers act as flow boundaries and increase the suction thresholds against piping.

7 Figure 9 shows the critical suction calculated by equation (10) with different ratios L /D. If L /D is large then the critical suction approaches the threshold for penetration in homogeneous sand. It is also seen that the presence of a flow boundary will increase the threshold where critical suction will occur. 6

PREDICTION OF FIELD TEST DATA

In Figure 10, the suction needed to install the bucket is plotted against equation (3) and (10). The figure shows that suction close to or higher than critical, predicted by equation (3), can be applied without significant consequences. This is particularly seen in the installation test with bucket 5. It is seen that the suction needed to overcome the resistance during the installation of the bucket 2 never violated the critical suction predicted by equation (10) with the flow boundary at 2.7 m. This was not the case in the installation test with bucket 4. At a depth of 1.56 m the applied suction violated the failure criterion predicted by equation (10) and piping channels were formed and observed during the test. At the test with © 2011 by Taylor & Francis Group, LLC

425

CONCLUSION

By comparing the numerical studies with the installation tests it is shown that it is the exit gradient next to the skirt which controls when piping will occur. For installation in homogeneous sand, the internal hydraulic gradients have been investigated by several researchers using computer programmes such as Plaxis, SEEP and FLAC.These studies have resulted in different formulations, but the empirical expressions predict similar critical suctions for skirt penetrations of practical interest. However, experience from installation of prototype foundations have shown that gradients close to critical, predicted by the expressions for homogenous sand, can be applied without significant consequences. The same was observed in the field test reported in this paper. It is stated that the presence of thin silt layers will act as flow boundaries and increase the suction thresholds against piping. The influence of the flow boundary was studied in this paper. The results are presented as simple closed form solutions and shown to predict thresholds against piping in homogeneous or layered sand. Future studies have to be performed in order to establish the thresholds against piping when the skirt penetrates through a flow boundary.

Figure 10. Installation tests analyzed using equation 10 with the flow boundaries interpret from the CPT tests in Figure 4.

REFERENCES Andersen, K.H., Murff, J.D., Randolph, M.F., Clukey, E., Erbrich, C., Jostad, H.P., Hansen, B., Aubeny, C., Sharma, P., and Supachawarote, C. (2005). “Suction anchors for deepwater applications.” Proc., INT Symp. On Frontiers in offshore Geotechniques (ISFOG). Keynote Lecture, Perth, Western Australlia, p. 3–30. Feld, T (2001). “Suction bucket, a new innovative foundation concept applied to offshore wind turbines.” Aalborg university, Aalborg. Hansen, B. (1978). Geoteknik og fundering del II. Laboratoriet for fundering. DTH. (In Danish). Ibsen, L.B. (2008). Implementation of a new foundations concept for Offshore Wind farms. Proc. Nordisk Geoteknikermøte nr. 15 NGM 2008, 3–6 September 2008 Sandefjord, Norge, 1–15.

© 2011 by Taylor & Francis Group, LLC

Itasca (2005). “FLAC3D – Fast lagrangian analysis of continua: Fluid-Mechanical Interaction”, Itasca Consulting Group Inc., Minneapolis, USA. Scott, R.S. (1963). Principles of soil mechanics. AddisonWesly Publiching Company, Inc. Senders. M. and Randolph M. F. (2009) “CPT-Based Method for the Installation of Suction Caissons in Sand” Jour. of Geotechnical and Geoenvironmental Enginnering. Senepere, D. and Auvergne, G. A. (1982) “Suction anchor piles – a proven alternative to driving or drilling.” Proc., 14th Offshore Technology Conf., Houston, Texas, 483–493. Tjelta T. I. (1995) “Geotechnical experience from the installtion of the Europipe jacket with bucket foundations” Proc., Offshore Technology Conf., Houston, Texas, Paper No. 7795.

426

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Shallow foundation performance in a calcareous sand B.M. Lehane The University of Western Australia

ABSTRACT: The paper presents results from eight small scale load tests performed on shallow footings founded on an unsaturated calcareous sand at a site located about 100 km North of Perth, Western Australia. These tests highlight inadequacies of traditional bearing capacity approaches and show the potential for more reliable estimates of bearing capacity through direct correlations with CPT data. Both the shear stiffness and creep characteristics of this particular calcareous sand deposit are shown to be consistent with those inferred from footing tests on a siliceous sand.

1

INTRODUCTION

Field research to improve predictions of the performance of shallow foundations in sand has largely concerned siliceous sands because of their predominance in urban areas around the world. However, virtually no field load test data have been reported for shallow foundations in offshore calcareous sands, which are abundant in oil and gas development regions. This shortage of data coupled with the well known high compressibility of calcareous sands lead to relatively conservative shallow foundations designs for offshore facilities. The field testing programme reported in this paper provides some redress for the data shortage and adds insights into a foundation’s stiffness and capacity in a uniformly graded calcareous sand. Figure 1. View of Ledge Point test site.

2 TEST PROGRAMME AND GROUND CONDITIONS AT TEST SITE Eight footings load tests, designated F1 to F8, were conducted at a coastal aeolian calcareous sand site in Ledge Point (see Figure 1), located about 100 km north of Perth, Australia. The tests, details of which are summarised in Table 1, involved loading of three footing types: (i) 300 mm diameter (D), 50 mm thick steel plates, (ii) 300 mm high reinforced concrete cylinders with D = 580 mm and (iii) 600 mm square, 110 mm thick reinforced concrete bases. A CPT truck was employed to provide the reaction for the load tests (see Figure 2). The sand around each footing was removed to a depth E (provided in Table 1); loose sand in its vicinity was then placed to give an overburden height above footing level with the height (H) provided in Table 1. Footings were ‘bedded-in’ and load tested incrementally to the maximum available reaction load of 200 kN or to a maximum footing settlement of at least 20 mm. Results for F3 were only recorded to a © 2011 by Taylor & Francis Group, LLC

maximum settlement of 7 mm due to a data-logger malfunction. The Ledge Point sand is uniformly graded with a mean effective particle size (D50 ) of 0.25 mm, a uniformity coefficient of about 2 and a calcium carbonate content of 90%. An electronic microscope image of the sand is reproduced in Figure 3, where the hollow structure and high aspect ratio of grains is apparent. Maximum and minimum void ratios of 1.21 and 0.90 were recorded by Sharma (2004), who measured a critical state friction angle for the sand of 39◦ . Thirteen, closely spaced, Cone Penetration Tests (CPTs) were conducted to a depth of about 4.5 m at the locations, relative to footing tests, indicated on Figure 4. Each CPT revealed a highly stratified profile comprising 0.5 m–1.5 m thick layers with high CPT end resistances (qc ) of ∼15 MPa alternating with layers of comparable thickness but relatively low qc values of ∼2 MPa. No systematic spatial variation of these

427

Table 1.

Footing load test details.

Footing number

D or B (m)

E (m)

H (m)

Zone of influence (m)

qc,avg (MPa)

E at s/D = 1% (MPa)

q at s/D = 20%, qf (kPa)

q at s/D = 10%, q0.1 (kPa)

F1 F2 F3 F4 F5 F6 F7 F8

D = 0.3 B = 0.6 D = 0.3 B = 0.6 D = 0.58 D = 0.58 D = 0.3 D = 0.3

0.5 0.3 0.3 0.3 0.4 0.35 0.2 0.4

0.15 0.11 0.15 0.11 0.2 0.17 0.05 0.15

0.5–0.9 0.3–1.0 0.3–0.7 0.3–1.0 0.4–1.1 0.35–1.0 0.2–0.65 0.4–0.8

6.1 8 7.1 7.2 11 7.2 4.3 8.3

32 15 21 21 15 17 16 41

1100 1100 1150 1420 ∼1500 1100 1250 1350

1000 880 1075 1175 1300 930 930 1120

Figure 2. Load testing a square footing.

Figure 4. Relative locations of footing tests and CPTs.

Figure 3. Ledge Point calcareous sand.

layers was evident, and it would appear that the profile arose due to the aeolian deposition of loose sand within very weakly cemented or much denser, strongly undulating calcareous terrain. Pits excavated for the footings were easily excavated (i.e. cementation levels appeared absent or very low) and their sides remained stable below a depth of 0.3 m due to the presence of © 2011 by Taylor & Francis Group, LLC

some suction. The CPTs indicated that the depth to the water table was in excess of 4.5 m. Figure 5 presents the CPT qc values recorded within the depth of influence of each footing test. CPTs were not performed immediately adjacent to F1 and F2 and, for these cases, the profile plotted on Figure 5 is an average of CPTs conducted close by (e.g. CPT1 and 4 were averaged to produce the F1 profile). Figure 5 indicates that, in the upper horizons (of most relevance to the footings), qc generally increases with depth to an average value of about 10 MPa at 0.8 m depth. The footings’ zone of influence, estimated using the recommendations of Burland & Burbidge (1985), is listed in Table 1 along with the average qc value within this zone (qc,avg ). 3

FOOTING BEARING CAPACITY

The variations of the applied bearing pressure (q) with the settlement (s) normalised by the footing width or

428

stress mobilised at settlement ratios of 10% and 20% were estimated by hyperbolic extrapolation (if necessary) of the curves on Figure 6; bearing pressures at a nominal (large) settlement ratio of 20% are referred to here as the footing bearing capacity (qf ). Equivalent footing elastic stiffness values, derived using the equation for a rigid punch, are provided in Table 1 at a typical working stress settlement ratio of 1%. The estimated qf values are about 1300 ± 200 kPa for all footings, irrespective of the footing width, shape or embedment. This trend is unexpected in a granular material but can be explained relatively well using the standard bearing capacity equation (e.g. as provided in API 2007) by a employing a non-zero cementation component (c
) of approximately 6 kPa and a critical state friction angle of 39◦ ; this cementation component contributes 75% of the total bearing capacity. The coefficient of variation for the ratio of predicted to measured capacity (Qp /Qm ) increases from 0.14 when c
= 6 kPa to 0.27 if c
is assumed zero. For the latter case (c
= 0), a mean Qp /Qm ratio of unity is achieved for a friction angle of 46◦ . This angle is far in excess of the angle of tan−1 (0.66 tan φ
) = 28◦ recommended by Finnie & Randolph (1994) for footings on un-cemented calcareous soil. The mis-match with standard bearing capacity theory, if c
= 0, may also be due to the size/stress level effect on the Nγ factor identified by De Beer (1965). Footing capacities at Ledge Point did not increase with embedment in the linear manner predicted by the standard bearing capacity equation. This characteristic has also been observed by Lehane (2010) in a review of available footing bearing capacity data and suggests that the mechanism of collapse is more akin to hemispherical expansion than to the classical Prandtl shear mechanism. Given this trend and the difficulties in assessing in-situ levels of cementation (noting that low levels of c
can have an enormous influence on capacity), it is suggested here that a more reliable estimate of capacity may be obtained using a simple correlation with the CPT qc value. The ratios of bearing stresses at a settlement to width ratio of 10% (q0.1 ) to the average qc value in the footings’ zone of influence (qc,avg ) are plotted against the qc,avg on Figure 7. Lehane & Randolph (2002) show that a similar trend is indicated by q0.1 values for bored piles and Lehane (2009) established the following relationship from measured q0.1 values at the base of piles in siliceous sand (noting that all terms in this relationship have the same units of stress):

Figure 5. CPT qc values within zone of influence of the test footings.

diameter (s/B or s/D) measured in each of the eight footing tests are presented in Figure 6. Loads were applied in increments with each increment applied for 5 minutes; typically ten increments were employed to achieve the ultimate or maximum load. The bearing © 2011 by Taylor & Francis Group, LLC

with f1 = 2.4 ± 0.7 where σv
is the average in-situ vertical effective stress within the zone of influence and pa is atmospheric pressure (=100 kPa). This same relationship is plotted on Figure 7, employing f1 = 2.4 and for a typical range of σv
values. It is evident that equation (1) also provides a very good representation of the variation observed in the calcareous sand at Ledge Point.

429

Figure 6. Bearing pressure vs. normalised settlement measured in eight footing tests.

Figure 8. Footing settlements during load increments (Footing F6). Figure 7. Dependence of q0.1 on qc and stress level.

4

CREEP SETTLEMENT

Calcareous sands are generally assumed to creep at a higher rate than siliceous sands because of the more compressible nature of their sand grains. The footing experiments showed that, after the settlement that occurred during application of the load increment, © 2011 by Taylor & Francis Group, LLC

creep settlements (sc ) varied approximately with the logarithm of time (t). A typical variation, as indicated by F6, is shown on Figure 8. The creep rate varied approximately with the square of the mobilised strength (q/qf ) or inverse of the factor of safety. Defining a creep coefficient, m, as the slope of relative settlement (s/D) or (s/B) versus natural logarithm of time (s/D/lnt), values for m were determined for each load increment in each footing test. Representative m

430

(ii) There is no correlation between E0.1 and qc,avg e.g. the E0.1 /qc,avg ratio varies from 1.4 for F4 to 5.2 for F1. (iii) This range of E0.1 /qc,avg ratios is comparable to that employed in the design of footings on siliceous sand. The higher compressibility of the Ledge Point sand evidently had a low influence on the stiffness operational in these tests. The equivalent stiffnesses of the footings are approximately twice those measured in shallow footing tests on siliceous sand by Lehane et. al. (2009) at the Shenton Park test site in Perth. The average CPT qc value within the footings’ zone of influence at Shenton Park was about 3.5 MPa and hence less than half of the Ledge Point averages (see Table 1). It is clear therefore that the Ledge Point calcareous sand does not exhibit stiffness characteristics which are dissimilar to equivalent siliceous sand deposits.

6

Figure 9. Variation of creep parameter (m) with mobilised strength.

values are plotted on Figure 9 which also plots the following equation:

It is seen that Equation (2) provides an approximate upper-bound to the Ledge Point m data. It is noteworthy that Lehane et al. (2009) showed that the identical equation provided a best-fit to creep data from footing tests on siliceous sands in Perth. It would therefore appear that the Ledge Point sand does not creep more than the siliceous dune sand in Perth. Both sands are above the water table and recent triaxial tests at UWA suggest that the degree of saturation over a certain range can have a marked effect on the shear creep characteristics. The same research suggested that equation (2) provides an upper-bound to the likely level of creep when the sand is fully saturated.

CONCLUSIONS

The footing tests performed at Ledge Point showed that the high calcium carbonate content of this sand (and hence high grain compressibility) did not lead to footing responses that differed significantly from those expected in a siliceous sand. It is shown that footing bearing capacities are best assessed using a correlation with the CPT qc value (or pressuremeter limit pressure) rather than with standard bearing capacity formulae. The operational stiffness and creep characteristics of the near surface (partially saturated) Ledge Point sand are comparable to those of siliceous sands under similar conditions.

ACKNOWLEDGEMENTS The author would like to acknowledge the contribution of four final year students at UWA namely Scott Doncon, Ben Hall, Nicola Reeves and Yuli Yao, all of whom assisted in the execution and interpretation of the tests described here. The technician team at UWA, led by Jim Waters, and the CPT contractor, Probedrill Pty Ltd, who provided free use of their CPT truck, are also gratefully acknowledged. REFERENCES

5

FOOTING STIFFNESS

A detailed analysis of the stiffness of the Ledge Point footings is outside the scope of this paper. Some general trends can, however, be drawn from Figure 6 and Table 1: (i) The equivalent linear stiffness at a settlement to width ratio of 1% (E0.1 ) is approximately 18 MPa for all footings except F1 and F8, which are almost twice as stiff. © 2011 by Taylor & Francis Group, LLC

American Petroleum Institute (API) 2007. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design, RP2A, Revisions/Edition: 21, Washington. Burland, J.B. and Burbidge, M.C. (1985). “Settlement of foundations on sand and gravel.” Proc. Design and Construction, Institution of Civil Engineers, Vol. 78, December, 1325–1381. De Beer, E. (1965). The scale effect on the phenomenon of progressive rupture in cohesionless soils. Proc. 6th Int. Conf. Soil Mech Found. Eng., 2, Montreal, Balkema, 13–17.

431

Doncon S. (2009). Settlement of shallow foundations on calcareous sand. Final Year Honours thesis, University of Western Australia. Finnie, I.M. and Randolph, M.F. (1994). Bearing response of shallow foundations in uncemented calcareous soil. Proc. Centrifuge ’94, 1, Singapore, Balkema, Rotterdam, 535–540. Lehane B.M., Doherty J.P. and Schneider J.A. (2009). Settlement prediction for footings on sand. Keynote Lecture, Proc. 4th International Symposium on deformation characteristics of geomaterials, Atlanta, 1, 133–152, IOS press, The Netherlands. Lehane B.M. and Randolph M.F. (2002). Evaluation of a minimum base resistance for pipe piles in sand.

© 2011 by Taylor & Francis Group, LLC

432

J. Geotechnical & Geoenvironmental Engrg. ASCE, 128 (3), 198–205. Lehane B.M. (2009). Relationships between axial capacity and CPT qc for bored piles in sand. Keynote Lecture, Proc. 5th International Symposium on deep foundations on bored and auger piles, Ghent, 1, 61–76, Taylor Francis Group, UK. Lehane B.M. (2010). Assessing the bearing capacity of footings on sand using in-situ test data. (in preparation) Sharma, S.S. (2004). Characterisation of cyclic behaviour of calcite cemented calcareous soils. PhD Thesis, The University of Western Australia.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

A numerical study of the vertical bearing capacity of skirted foundations D.S.K. Mana, S. Gourvenec & M.F. Randolph Centre for Offshore Foundation Systems, UWA, Crawley, Australia

ABSTRACT: Finite element analysis is used to investigate the vertical bearing capacity of circular skirted foundations considering the effect of embedment ratio, foundation-soil interface roughness and soil strength heterogeneity. The effect of idealising a skirted foundation as a solid rigid plug and idealising geometry to conditions of plane strain are also addressed through comparison of bearing capacity factors and kinematic mechanisms accompanying failure. A closed-form expression is presented that enables prediction of bearing capacity factors for circular skirted foundations over a practical range of embedment ratio, skirt-soil interface roughness and soil strength heterogeneity, to within ±2.5% of the finite element calculations.

1

INTRODUCTION

Skirted foundations are used to support or moor a variety of offshore structures, such as gravity based structures, tension leg platforms, jacket structures, storage tanks and various sub-sea infrastructure (such as manifolds and pipeline end terminations – PLETs). Investigation of bearing capacity of shallow foundations has typically considered rigid plugs or buried plates (e.g. Martin 2001, Martin & Randolph 2001, Bransby & Randolph 1999, Salgado et al. 2004, Edwards et al. 2005, Gourvenec 2008). The effect of a deformable soil plug (as confined by a skirted foundation) has received less attention, and analysis has been limited to plane strain conditions and a fully rough foundation-soil interface (Yun & Bransby 2007, Gourvenec & Barnett 2010). A particular interest with skirted foundations is whether or not an internal mechanism will develop within the soil plug, hence reducing bearing capacity (unless provision is made for internal skirts or soil improvement within the soil plug). An internal mechanism would not be expected to develop in soils with uniform shear strength with depth, i.e. the soil plug would be expected to displace as a rigid body. However, in soils with increasing shear strength with depth, a failure mechanism may develop within the soil plug reducing bearing capacity. From existing theory, it would be anticipated that an internal failure mechanism would only develop when a Hill-type mechanism governs failure, as opposed to a Prandtl-type mechanism (illustrated in Figure 1). It is well established that a Hill-type mechanism governs failure of smooth-based surface foundations and becomes the critical failure mode for rough-based surface foundations when soil strength heterogeneity is significant (Kusakabe et al. 1986). Martin & Randolph (2001) noted that a Hill-type mechanism is always critical for smooth-based shallowly embedded © 2011 by Taylor & Francis Group, LLC

433

Figure 1. Kinematic failure mechanisms of shallow foundations under vertical load (modified from Kusakabe et al. 1986).

foundations and is occasionally critical for roughbased shallowly embedded foundations, notably with increased soil strength heterogeneity – as for surface foundations. Skirted foundations are essentially rough-based, due to the soil-soil interface at foundation level and it would be of use to investigate the embedment ratio, as a function of soil strength heterogeneity, at which an internal mechanism will or will not develop. This paper presents results from an investigation of bearing capacity of circular and strip shallow foundations. Initially, results of skirted foundations, with a deformable soil plug and embedment modelled by a rigid plug are compared for a uniform shear strength and a highly heterogeneous shear strength profile. Subsequently, a parametric study considering the effect on bearing capacity of a range of soil strength heterogeneity and skirt-soil interface roughness is presented for circular skirted foundations. 2

FINITE ELEMENT MODELS

Small strain finite element analyses were carried out using the software Abaqus (Dassault Systèmes 2009). Shallow foundations with embedment provided either by a peripheral skirt or a rigid plug were considered in axi-symmetry and plane strain. Embedment ratios

Figure 3. Bearing capacity factor for rough-sided circular skirted foundations and rigid solid plugs – kD/sum = 0. Figure 2. Finite element mesh, d/D = 0.5.

(d/D or d/B) of 0 (surface), 0.1, 0.2, 0.3 0.5 and 1 and skirt wall thickness (where relevant), t/D = 0.008 were considered, where d is the skirt embedment depth, D is the diameter of the circular foundation and B is the breadth of foundation in plane strain. The soil was modelled as a linearly elastic perfectly plastic Tresca material with a submerged unit weight γ
= 6 kN/m3 . Within the elastic regime, the soil had a stiffness ratio E/su = 500, where E is the Young’s modulus of the soil. Poisson’s ratio ν was taken as 0.499. Within the plastic regime, uniform undrained shear strength and linearly increasing strength with depth were considered. Linearly increasing shear strength with depth is described by su = sum + kz, where sum is the undrained shear strength at the mudline and k is the gradient of increase in shear strength with depth z and the degree of heterogeneity is described by the dimensionless group kD/sum . A range of soil strength heterogeneity 0 (uniform with depth) ≤ kD/sum ≤ 20 were considered. The foundation-soil interface was modelled with a contact surface, rough in shear and perfectly bonded preventing slip or separation. For modelling interface roughness, a narrow band of soil elements, the same width as the skirt thickness, was considered adjacent to the outer face of the skirt. This narrow band of soil elements was given shear strength equal to α times the intact shear strength in the rest of the soil mass at that depth. (This approach is necessary as a contact surface with constant interface friction factor cannot be defined in Abaqus.) A typical axi-symmetric finite element mesh is shown in Figure 2. Similar mesh discretisation was maintained for different embedment ratios and for the plane strain meshes. The width of the modelled soil zone was more than three times the diameter from the centre of the skirt and depth was more than six times the diameter of the skirt, to prevent boundary effects on the bearing capacity. Roller supports were provided along the vertical boundaries and along the base of the mesh. Second-order bi-quadratic continuum elements with reduced integration were used to model the soil © 2011 by Taylor & Francis Group, LLC

Figure 4. Bearing capacity factor for rough-sided strip skirted foundations and rigid solid plugs – kD/sum = 0.

(CAX8R and CPE8R).The foundations were modelled as rigid bodies, pre-embedded (i.e. installation was not modelled). Each foundation was brought to failure by a displacement-controlled load path. A reference point (RP) for applying and recovering loads and displacements to the foundation was specified along the centre line of the foundation at skirt tip level. In order to eliminate the effect of foundation weight on the net end bearing capacity of the foundation, the part of the foundations below the mudline were prescribed a unit weight equal to that of soil and the part above was modelled as weightless. 3 3.1

RESULTS AND DISCUSSION Skirted foundations and rigid solid plugs – kD/sum = 0

Figures 3 and 4 show the calculated vertical bearing capacity factors, Nc = qult /su , for rough-sided circular and strip skirted foundations and rigid solid plugs in soil with uniform undrained shear strength, i.e. kD/sum = 0. A vertical bearing capacity factor Nc = 5.95 was calculated for the surface circular foundation, under predicting the exact solution of 6.05 by less than 2% (Cox et al. 1961). This slight under prediction is attributed to the rounding of vertices of the

434

Figure 5. Displacement contours for strip (left) and circular (right) rough skirted foundations with kD/sum = 0 (Contour interval δ/D = 0.025).

hexagonal Tresca yield surface in Abaqus (Taiebat & Carter, 2008), possibly exacerbated by the larger deformation of soil elements close to the foundation in axi-symmetry compared with plane strain (Gourvenec et al. 2006). An ultimate vertical bearing capacity Nc = 5.22 was calculated for the surface strip foundation which over predicts the exact solution of 5.14 by less than 2%. The calculated bearing capacity factors for the embedded foundations are validated by comparison with available upper and lower bounds for rigid solid plugs shown in Figures 3 and 4 (Martin 2001, Bransby & Randolph 1999). The bearing capacity factor increases with increasing embedment ratio, as would be expected, and the bearing capacity factors for skirted foundations are coincident with those for a rigid solid plug of equivalent embedment for both the circular and strip foundations (as also noted by Yun & Bransby (2007) for strip foundations). The non-linearity of the relationship between bearing capacity and embedment ratio for the circular foundation geometry (as seen in Figure 3) indicates the transition of failure mechanisms as illustrated in Figure 5; from a traditional Prandtl-type surface failure mechanism (a & b), to an annular flow mechanism where the angle of exit at the soil surface tends to 90◦ (c) to a confined mechanism (d & e). The almost linear relationship between bearing capacity and embedment ratio for the plane strain case (Figure 4) is reflected in the similarity of the failure mechanisms of a Prandtl-type surface failure irrespective of embedment ratio (Figure 5). The more extensive failure mechanisms, both vertically and laterally, under plane strain conditions compared to axi-symmetry are clear from Figure 5. Prandtl-type mechanisms governed failure of both the solid rigid foundations and the skirted foundations i.e. the soil plug displaced as a rigid body, consistent with the coincidence of the bearing capacity factors presented in Figures 3 and 4. 3.2 Skirted foundations and rigid solid plugs – kD/sum = 20 Figures 6 and 7 show the calculated vertical bearing capacity factors, Nc = qult /su0 , where su0 defines the undrained shear strength at foundation level, for © 2011 by Taylor & Francis Group, LLC

435

Figure 6. Bearing capacity factor for rough-sided circular skirted foundations and rigid solid plugs – kD/sum = 20.

Figure 7. Bearing capacity factor for rough-sided strip skirted foundations and rigid solid plugs – kD/sum = 20.

skirted foundations and rigid solid plugs in a soil with linearly increasing undrained shear strength with depth, kD/sum = 20 (considered to represent the high end of soil heterogeneity commonly encountered). The calculated bearing capacity factors are validated by comparison with upper and lower bounds for rigid circular plugs (Martin 2001). For the strip foundation, no theoretical solution is available for kD/sum = 20, but an upper bound solution for rigid strips for kD/su0 → ∞ is shown (Bransby & Randolph 1999). The vertical bearing capacity factors for the skirted foundations are coincident with those for rigid solid plugs of equivalent embedment for d/D > 0.23 and 0.28 for circular and strip geometry respectively. For smaller embedment ratios, lower bearing capacity factors are calculated for the skirted foundations; more so for the strip foundation (−8% for d/D = 0.1) compared with the circular foundation (−5% for d/D = 0.1). Examination of the kinematic mechanisms accompanying failure, shown in Figure 8 for the circular case, reveals an internal mechanism inside the skirt for d/D < 0.3, consistent with the diminishing vertical bearing capacity relative to that for the solid plug foundation. It is interesting to consider shape factor, sc = Nc,circle /Nc,strip , as a function of soil strength heterogeneity and embedment ratio. Salençon & Matar (1982) and Houlsby & Wroth (1983) present shape

Figure 8. Displacement contours for rough circular solid plugs (left) and skirted foundations (right) respectively – kD/sum = 20 (Contour interval δ/D = 0.025).

Figure 10. Shape factor as a function of embedment ratio and soil strength heterogeneity.

Figure 9. Shape factor as a function of soil strength heterogeneity for surface foundations.

Figure 11. Bearing capacity factor for circular skirted foundations with varying interface roughness – kD/sum = 0.

factors for surface foundations as a function of heterogeneity parameter kD/sum based on plasticity solutions. Both show diminishing shape factor with increasing shear strength heterogeneity and report shape factors of less than unity for kD/sum > 2. The results from this study indicate sc = 0.83 for d/D = 0 and kD/sum = 20, falling in the line of extrapolation of Houlsby and Wroth’s data (Figure 9). FEA of surface circular foundations with varying kD/sum reported by Gourvenec & Randolph (2003) also agreed well with the lower bound solutions of Houlsby and Wroth. Thus, as reported by Martin & Randolph (2001), the characteristic solutions given by Houlsby & Wroth (1983) are exact, within the accuracy of the numerical implementation, rather than just a lower bound (LB). Figure 10 shows shape factor as a function of embedment ratio for kD/sum = 0 and 20, showing increasing shape factor with increasing embedment ratio and reducing shape factor with increasing soil strength heterogeneity. For kD/sum = 20, the shape factor is less than unity only for d/D < 0.2. Shape factor clearly vary with both embedment ratio and soil strength heterogeneity in a complex manner (resulting from the different kinematic mechanism governing failure), thus highlighting the benefit of considering foundation shape explicitly rather than relying on © 2011 by Taylor & Francis Group, LLC

shape factors to adjust bearing capacity calculations based on plane strain conditions.

3.3

Circular skirted foundations – Effect of foundation-soil interface roughness and shear strength heterogeneity

In the foregoing, a single heterogeneous soil profile was considered and the foundation-soil interface was assumed to be fully rough, with no slip or separation of soil at the interface permitted. The purpose of the analyses was to illustrate different governing modes of failure and the effect of idealising a skirted foundation as a solid rigid plug and idealising geometry to conditions of plane strain. A parametric study was subsequently undertaken to investigate the effect of interface roughness and degree of soil strength heterogeneity on bearing capacity of circular skirted foundations. A range of interface friction αsu , with α between 0 (frictionless) and 1 (fully rough) for shear strength profiles with kD/sum between 0 (uniform) and 20 was considered. Figure 11 shows bearing capacity factors, Nc = qult /su0 , against embedment ratio as a function of interface roughness for the case of uniform shear strength profile (kD/sum = 0). Lower bearing capacity is mobilised with reduced skirt-soil roughness

436

Figure 12. Variation of bearing capacity factor with skirt-soil interface roughness as a function of embedment ratio kD/sum = 0.

as would be expected, and the non-linearity between bearing capacity factor and embedment ratio becomes more marked with reducing skirt-soil roughness. Figure 12 shows the data from Figure 11 presented in an alternative form as bearing capacity factor against friction factor as a function of embedment ratio, giving approximately linear relationships, as shown by the dashed lines of best fit. Equation 1 defines the best fit lines giving the bearing capacity factor for a given interface roughness, Nc,α as the sum of the bearing capacity factor for a smooth footing (representing the intercept on the vertical axis in Figure 12) and an additional component due to skirt roughness for varying embedment; clearly the effect of skirt roughness becomes increasingly significant with increasing embedment. Equation 1 predicts bearing capacity factors to within 2.5% of the finite element calculations.

Bearing capacity of shallow foundations has been alternatively presented as the sum of base resistance and shaft resistance (e.g. Byrne & Cassidy 2002,Yun & Bransby 2007), analogous to the calculation of capacity of a deep pile foundation. This approach is only valid if the base and shaft resistance can be decoupled, which is only be the case if a single prevalent mechanism governs failure independent of skirt-soil interface roughness.Yun & Bransby (2007) present upper bound analyses that indicate that this is not the case and that base bearing may be 10% lower for a smoothsided foundation than a rough-sided foundation with d/D = 1. Observations from the finite element analyses in this study also showed the governing failure mechanism is dependent on interface roughness. Bearing capacity factors for circular skirted foundations with varying interface roughness were additionally calculated for conditions of soil strength heterogeneity given by kD/sum = 2, 5, 10, 15 and 20. Although the relationship between bearing capacity factor and embedment ratio differ in form for cases of linearly increasing shear strength with depth compared with the case of uniform shear strength (as shown by © 2011 by Taylor & Francis Group, LLC

Figure 13. Bearing capacity factors for smooth-sided circular skirted foundations, Nc,α=0 , in soils with varying soil strength heterogeneity.

Figure 14. Coefficient C for Equation 2.

comparing Figures 3 and 6), a linear relationship is observed when the data is presented as bearing capacity factor against interface roughness as a function of embedment ratio (as illustrated in Figure 12 for kD/sum = 0). Over the range of kD/sum considered, the relationship between bearing capacity factor and skirt-soil interface roughness can be expressed by a generalised form of Equation 1

where the value of Nc,α=0 and the constant C depend on the value of kD/sum . The bearing capacity factors for smooth-sided foundations, Nc,α=0 , are plotted against embedment ratio for 0 ≤ kD/sum ≤ 20 in Figure 13. (These have also been validated against upper and lower bounds for circular rigid soil plugs (Martin, 2001)). The relationships are not conducive to description by a closed-form expression and must be read from (or interpolated between) the plotted data. The values of the constant C, required for Equation 2, are plotted in Figure 14 as a function of kD/sum and can be described by an exponential expression

437

Thus, with Equations 2 and 3 and Figure 13, bearing capacity factors can be predicted for circular skirted foundations over a range of embedment ratios 0 ≤ d/D ≤ 1, with skirt-soil interface roughness 0 ≤ α ≤ 1 in soils with shear strength heterogeneity 0 ≤ kD/sum ≤ 20 to within ±2.5% of the finite element calculations. 4

CONCLUSIONS

Finite element analysis has been used to investigate the vertical bearing capacity of circular skirted foundations as a function of embedment ratio, foundation-soil interface roughness and soil strength heterogeneity and to assess the effect of idealising foundation geometry with a rigid soil plug and to conditions of plane strain. Three-dimensional effects are shown to have a significant influence on bearing capacity of shallow foundations with shape factors varying in a complex manner with foundation embedment ratio and soil strength heterogeneity such that explicit consideration of foundation shape is recommended. The effect of the type of embedment, i.e. solid or skirted, is shown to be less significant than three-dimensional effects with bearing capacity factors for circular skirted foundations a maximum of 5% less than that of an equivalent solid rigid plug. Internal mechanisms are shown to be most prone for lower embedment ratios and high soil strength heterogeneity. In practice, internal mechanisms are likely to be more significant under horizontal loading and overturning than under pure vertical load. A closed-form expression is presented for the calculation of bearing capacity factors for circular skirted foundations over a practical range of embedment ratio, skirt-soil interface roughness and soil strength heterogeneity, to within ±2.5% of the finite element calculations. ACKNOWLEDGEMENT The Centre for Offshore Foundation Systems was established under the Australian Research Council’s Research Centres Programme and is supported by the State Government of Western Australia through the Centres of Excellence in Science and Innovation Program. The work presented in this paper was funded by an ARC grant DP0988904. This support is gratefully acknowledged.

Byrne, B. & Cassidy, M.J. 2002. Investigating the response of offshore foundations in soft clay soils. 21st Int. Conf. Offshore Mech. and Arctic Engng., New York, USA, OMAE2002-28057. Cox,A.D., Eason, G. & Hopkins, H.G. 1961.Axially symmetric plastic deformation in soils. PhilosophicalTransactions of the Royal Society of London (Series A), 254: 1–45. Dassault Systèmes 2009. Abaqus analysis users’ manual, Simula Corp, Providence, RI, USA. Davis, E.H. & Booker, J.R. 1973. The effect of increasing shear strength with depth on the bearing capacity of clays. Géotechnique 23(4): 551–563. Edwards, D.H., Zdravkovic & Potts, D.M. 2005. Depth factors for undrained bearing capacity. Géotechnique 55(10): 755–758. Gourvenec, S. 2008. Effect of embedment on the undrained capacity of shallow foundations under general loading. Géotechnique 58(3): 177–185. Gourvenec, S. & Barnett, S. (2010 accepted, Géotechnique). Undrained failure envelope for skirted foundations under general loading. Gourvenec, S., Randolph, M.F. & Kingsnorth, O. 2006. Undrained bearing capacity of square and rectangular footings. International Journal of Geomechanics 6(3): 147–157. Gourvenec, S. & Randolph, M.F. 2003. Effect of strength non-homogeneity on the shape and failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique, 53(6): 575–586. Houlsby, G.T & Wroth, C.P. 1983. Calculation of stresses on shallow penetrometers and footings. Proc. IUTAM/IUGG Symp. On Seabed Mech., Newcastle upon Tyne, 107–112. Kusakabe, O., Suzuke, H. & Nakase, A. 1986. An upper bound calculation on bearing capacity of a circular footing on a non-homogeneous clay. Soils & Foundations 26(3): 143–148. Martin, C.M. 2001. Vertical bearing capacity of skirted circular foundations on Tresca soil. Proc. Int. Conf. Soils Mech. and Geotech. Engng (ICSMGE), Istanbul Martin, C.M. & Randolph, M.F. 2001. Applications of the lower and upper bound theorems of plasticity to collapse of circular foundations. In (ed.), Proc. 10th Int. Conf. Int. Assoc. of Computer Methods and Advances in Geomech (IACMAG), Tucson, 1417–1428. Salençon, J. & Matar, M. 1982. Capacité portante des foundations superficielles circulaires. Journal de Mécanique théorique et appliquée 1(2): 237–267. Salgado, R., Lyamin, A.V., Sloan, S.W. & Yu, H.S. 2004. Two and three-dimensional bearing capacity of foundations in clay. Géotechnique 54(5): 297–306. Taiebat, H.A. & Carter, J.P. 2008. Flow rule effects in the Tresca model. Computers & Geotechnics 35(3): 500–503. Yun, G. & Bransby, M.F. 2007. The undrained vertical bearing capacity of skirted foundations. Soils & Foundations 47(3): 493–505.

REFERENCES Bransby, M.F. & Randolph, M. F. 1999. The effect of embedment depth on the undrained response of skirted foundations to combined loading. Soils & Foundations 39(4): 19–33.

© 2011 by Taylor & Francis Group, LLC

438

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

The effect of torsion on the sliding resistance of rectangular foundations J.D. Murff & C.P. Aubeny Texas A&M University, Austin, Texas, USA

M. Yang Aker Solutions (formerly Texas A&M University, Houston, Texas, USA)

ABSTRACT: The undrained sliding capacity of a shallow foundation on the sea bed under lateral load is one of the key considerations in its design. Lateral loads may arise from combinations of wind, wave, current, earthquake and/or ice loading and from operational loads. A number of offshore structural systems have recently been developed that are founded on rectangular mats. The existence of torsion will reduce the lateral sliding capacity of the foundation. This paper describes a kinematically admissible model, based on plastic limit analysis, that accounts for the effect of torsion on lateral sliding resistance including the interaction among the passive, active, and side shear components of resistance. The relative effects predicted by the model are supported by finite element analyses. It is demonstrated that the torsion effect can play a very significant role in the foundation design.

1

INTRODUCTION

2

Classical methods for analyzing shallow foundations generally do not include the effects of torsion loading, as such loading is relatively uncommon; however, analyses of a few special cases have been reported. Torsion effects on the sliding resistance of a group of surface footings were addressed by Murff and Miller (1977) using the upper bound method. More recently Finnie and Morgan (2004) used limit equilibrium methods to study torsion-sliding response of rectangular surface footings andYun et al. (2009) have used the finite element method to study the interaction among sliding, torsion and vertical load on surface footings. Recent offshore applications such as LNG facilities in shallow water subjected to asymmetric environmental loads and deepwater pipeline terminals and manifolds subjected to offset pull-in and thermal loads have underscored the importance of torsion effects on rectangular foundations. These foundations normally are equipped with shear skirts so that the foundation is effectively embedded. In some cases a weak horizontal layer below the foundation may exist which is a preferential failure plane. In such cases the lateral resistance includes not only the shear strength of the failure plane but passive and active resistance on the skirts or vertical soil interface and side shear on the skirts or soil interface. To circumvent the requirement for always having to conduct finite element analyses on such complex problems, a relatively simple solution is proposed which employs the upper bound method of plasticity. © 2011 by Taylor & Francis Group, LLC

MECHANISM

The solution approach taken here is an application of the upper bound method of plasticity (Drucker and Prager, 1952). This requires a kinematically admissible mechanism for which the work rate of external forces is equated to the rate of internal dissipation of energy. The unknown force is evaluated and subsequently minimized with respect to parameters defining the mechanism geometry. The failure mechanism employed in this analysis is shown in Figures 1(a), 1(b), and 1(c). The soil defined by DEFG down to a depth d is a rigid block rotating about a vertical axis through point xo , yo . The long wedges shown on each side of the block translate and deform to accommodate the block motion. A brief explanation of the details of the mechanism follows. Figure 1(a) is a plan view of the block DEFG. The zones of the triangular wedges along the sides undergo passive or active failure as indicated by letters “p” and “a”. Figure 1(b) is an elevation view of a cross section HH’ through the rigid block and wedges on sides 1 and 3. Figure 1(c) shows a typical velocity field for a wedge cross section for the velocity component normal to and compatible with the block’s vertical face. The wedge cross section in the example slips along B1-C1 without deforming. The wedge does deform in the direction parallel to the block sides (x direction in this example) to accommodate the variation in the block lateral velocity. The block also has a velocity parallel to its sides but since

439

Figure 1. Schematic of sliding-torsion mechanism.

the wedge cannot displace in this direction there must be slip at the interface (A1-B1 in this case). It should be noted that owing to the inclusion of both active and passive zones in the wedge the work rate done by the unit weight of the soil is zero. The sources of dissipation are thus as follows:

where ε1 is the maximum principal tensor strain. The velocities in the wedge are

• Shear deformation in the wedges due to varying

The shear strain rates are thus

velocity along the sides. In the example shown in Figures 1(b) and 1(c) this gives rise to shear strains εxy and εxz while εyz = 0. The dissipation rate for a Tresca yield condition is

© 2011 by Taylor & Francis Group, LLC

440

The maximum principal strain rate is then

The dissipation is then integrated over the volume of the wedge in question. It should be noted here that su is taken as a function of z and since the integrand is then a function of z only the dissipation expression can be reduced to a single integral as follows

The subscript RE1 indicates the right end of the wedge on side 1. A similar expression is developed for the other seven wedge ends. • Slip at the base of the rigid block DEFG relative to the rigid soil below the block. The dissipation on the slip surface below the footing is determined by integrating the relative slip velocity over the area below the footing as follows

Note that the integral is independent of whether the wedge is yielding actively or passively. The dissipation in the remaining wedges can be determined in a similar manner. • Slip between the rigid rotation block and the vertical faces of the wedges. The wedges move up relative to the block and the block moves parallel to its edges relative to the wedges. The non-zero slip velocities on the vertical face are then

Taking the resultant and integrating over the vertical interface along the wedge gives

where su is evaluated at depth d. The external work rate of the resultant load F is then

The upper bound estimate of the capacity F is then found by setting the external work rate equal to the ˙ canceling the sum of all dissipation rate terms, D, ˙ and solving for F which gives β’s,

This expression is then minimized with respect to the optimization variables xo , yo , w1 , w2 , w3 , and w4 . Where the depth of the failure plane is obvious, d can be taken as given but it can also be included as an optimization variable. where α is a factor between 0 and 1 indicating the fraction of shear strength mobilized on the vertical interface. The subscript RS1 indicates the resultant velocity on side 1. The dissipation on the other sides is determined in a similar manner. • Slip along wedge ramps (B1–C1 in the example). The resultant velocity up the wedge ramp is then

and the dissipation is thus

The subscript RR1 indicates the resultant velocity on ramp 1. • Slip at the two ends of each wedge relative to the rigid soil outside the wedges. The resultant relative velocity on the ends of the wedge is the same as the slip on the ramps except that x is equal to L/2 or W/2 . The dissipation on the right end of side 1 is then

© 2011 by Taylor & Francis Group, LLC

3

PARAMETRIC STUDY

A brief parametric study was carried out to demonstrate the effect of torsion on sliding capacity for varying aspect ratio (W/L), varying failure surface depth, and varying load angle. Figure 2 summarizes the results of the study for W/L = 0.5 and 1.0 and for d/L = 0.0 and 0.05 for a load angle of 90 degrees. The results are normalized by the sliding and torsion resistance of the respective footings for d = 0, that is by the sliding and torsion results of a simple plate resting on the surface. For the case of d/L = 0.0, the curves for varying W/L have the same limiting values and differ only slightly in shape. For the case where the failure surface is at a depth of 0.05 L the limiting normalized values are, of course, greater than 1.0 due to contributions from the passive and active wedges and the side shear. The shapes of the interaction surfaces however are somewhat similar. For the case of a surface footing the interaction curves are not affected by the load angle. Figure 3 shows the effect of load angle for the embedded footing with a W/L ratio of 1/2. As expected the pure torsion values are unaffected by the load direction. The sliding resistance, however, is significantly affected. The end on loading (0 degrees) gives the smallest resistance due to the smaller end area. The resistance increases with load direction to approximately 45 degrees due to

441

Figure 2. Normalized interaction curves for varying w/l and d/l for a load angle of 90 degrees (broadside).

Figure 4. Comparison of upper bound solution with finite element results.

where

A simpler empirical expression has been determined by curve fitting the analytical solution as follows:

Figure 3. Normalized interaction curves for varying load angle for d/l = 0.5 and W/L = 1/2.

the larger projected areas and then decreases slightly for broadside loading. 4 VERIFICATION One check on the model is to compare model results with available analytical results for limiting cases. The sliding capacity of the surface plate (d = 0.0) is simply Fmax = LWsu which is identical to the model. For the embedded footing a solution can be obtained for purely broadside loading or purely end on loading by adding the plate sliding resistance to that of the active and passive wedges, the wedge ends, and the side shear on faces parallel to the load directions. These too, check exactly. Note that the triangular wedges are only exact solutions for the full plane strain solution with α = 0. For α = 1 the error in the passive/active component for the triangular wedge solution is approximately 10 percent. The overall error is generally much less. It is also possible to obtain a closed form solution for pure torsion, i.e. rotation about the centroid, as follows:

© 2011 by Taylor & Francis Group, LLC

Except for very small errors due to numerical integration, the model results are identical to the closed form solution. Additional checks were made for the maximum torsional capacity with the solutions presented by Yun, et al (2009) and Finnie and Morgan (2004). These were found to check essentially exactly. Figure 4 shows interaction diagrams for the case of a surface footing subjected to combined sliding and torsional loading compared to the solution presented by Yang, et al. (2010). The solution by Yang et al. was obtained for a deeply embedded, infinitely thin plate but the plots normalized by their maximum values are believed to be appropriate for this comparison. As shown in the Figure the differences in the interaction diagrams between the upper bound solutions and the FEM solutions are quite small. It is also noted that the solutions for the square and rectangular footings are in the correct relative order.

442

5

SUMMARY AND CONCLUSIONS

In this paper we have presented a solution to the problem of an embedded rectangular footing subjected to sliding and torsion loading. This solution is relevant to recent offshore problems ranging from shallow water LNG facilities to deepwater pipeline terminals and manifolds. The solution presented is capable of considering such complexities as combined torsion and sliding, footing embedment, and non-homogeneous soil strength profiles but is relatively straightforward

to implement on a simple spreadsheet. Example calculations for a range of parameters are presented and various comparisons are made to verify the solution is consistent with other known solutions. REFERENCES Drucker, D. C. and Prager, W. 1952. Soil Mechanics and plastic analysis or limit design. Q. Applied Math., 10: 157–165. Finnie, I.M.S. and Morgan, N. 2004. Torsional Loading of Subsea Structures, Proceedings Fourteenth International Offshore and Polar Engineering Conference, Toulon, France.

© 2011 by Taylor & Francis Group, LLC

Murff, J. D. and Miller, T. W. 1977. Stability of Offshore Gravity Structure Foundations Using the Upper Bound Method, Proceedings, Offshore Technology Conference, Houston, TX., May 1977. Yang, M., Murff, J. D. and Aubeny, C. P. 2010. Undrained Capacity of Plate Anchors Under General Loading, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Accepted for publication. Yun, G. J., Maconochie, A., Oliphant, J. and Bransby, F. 2009. Undrained Capacity of Surface Footings Subjected to Combined V-H-T Loading, Proceedings Nineteenth International Offshore and Polar Engineering Conference, Osaka, Japan.

443

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Foundation design challenges of the MCR-A skirted gravity platform L. Tapper & C. Humpheson Arup

B.M. Lehane The University of Western Australia

ABSTRACT: The MCR-A platform is a skirted gravity base platform to be installed in the Caspian Sea offshore Turkmenistan in ground conditions composed of stiff clay. Advanced 3D numerical analysis was required to adequately model the annular geometry and high environmental loading on the substructure. High suction pressures are needed to install the skirts to the depth necessary for foundation stability. The high suction pressures required a complex skirt stiffener arrangement and this posed challenges in quantifying the penetration resistance to be overcome in order to ensure the required skirt penetration could be achieved. In addition, the presence of spudcan footprint ‘craters’ in the direct vicinity of the platform resulting from previous jack-up operations required special design consideration. The approaches carried out to overcome these geotechnical design challenges are described in this paper. 1

INTRODUCTION

The Magtymguly Collector Riser platform (MCR-A) is a skirted gravity base foundation to be installed in the Magtymguly field located in the Caspian Sea, offshore Turkmenistan (Fig. 1). The substructure design for the platform has been undertaken by Arup. This paper provides a high level overview of a selection of the more challenging aspects of the foundation design, including: (i) Assessment of the bearing capacity of the annular foundation geometry under 3D loading. (ii) Quantifying the penetration resistance of skirts with a complex stiffener arrangement. (iii) Evaluation of the impact on foundation installation and in-place performance of spudcan induced ‘craters’ in the vicinity of the proposed platform. 1.1 Platform details and installation approach The MCR-A substructure comprises a square annular shaped, steel plated gravity base foundation, of outer dimension 54 m and inner dimension of 34 m. 5 m deep skirts extend below the base plate to form 12 compartments. Four legs made up of steel plated box girders beneath a four-chord steel tubular lattice are used to support the topsides (Fig. 2). This unusual arrangement allows the platform to be fabricated in Malaysia in components that can be shipped through the Volga-Don canal for final assembly in Turkmenistan. As limited offshore installation kit is available in the Caspian Sea, this configuration also provides the platform with sufficient buoyancy to © 2011 by Taylor & Francis Group, LLC

445

Figure 1. Site location.

float, and to be self-installing using controlled ballasting. The skirts are then penetrated into the seabed using a combination of self weight and under-base suction. Further details of the MCR-A platform are given in Pinna et al. (2009). 1.2

Ground conditions

Geotechnical site investigations were undertaken at the MCR-A location which included piezocone penetrometer testing (PCPT), seismic PCPT and sampling boreholes. The investigation was supported by offshore and onshore laboratory testing involving a suite of characterisation and advanced tests. Ground conditions at the site comprise firm to stiff over-consolidated clay to approximately 11 m below seabed, beyond which interbedded clay and sand layers were present to depth. The design monotonic

(2007). However, accurately assessing the MCR-A foundation capacity using these guidelines was challenging because: a) Whilst shape factors are given to account for rectangular or square shaped foundations, these cannot be readily applied to the annular geometry of the MCR-A foundation. b) Procedures are mostly geared towards in-plane loading rather than the 3D and multi-directional loading that MCR-A is subjected to. c) Solutions are only available for a range of idealised soil profiles.

Figure 2. Sketch of the MCR-A platform.

Figure 3. Three dimensional foundation loading regime.

undrained simple shear strength (su-ss ) profile with depth (z) within the zone of influence of the foundation can be approximated by su-ss = 35 + 4 z (kPa), although a soft layer of roughly 0.5 m thickness was present at 2.5 m. Cyclic strength profiles were also derived to account for the cyclic nature of the foundation loading. 2

FOUNDATION CAPACITY ASSESSMENT

The MCR-A base will be subjected to significant environmental loading (see Fig. 3). A typical design load case due to self weight and a 100 year return period storm event comprises a vertical load of about 240 MN, a horizontal load of up to 40 MN and moments in excess of 1000 MNm. MCR-A is also subjected to high seismic loading, but this aspect is not addressed in this paper.

2.1 Analytical assessment of bearing capacity The ultimate bearing capacity of shallow offshore foundations are commonly assessed using methods based on classical bearing capacity theory provided in design codes such as DNV (1992), ISO (2003) & API © 2011 by Taylor & Francis Group, LLC

As a result of these factors, only a simplistic analytical assessment of the ultimate capacity could be undertaken. This was based on the DNV (1992) method, supplemented with procedures outlined in API (2007) for the determination of effective area for moment loading in two directions. The skirts were included in the assessment, which enabled the weak layer at 2.5 m to be avoided by transferring the foundation loads to more competent deeper soils. The resulting soil plug trapped in the compartments could also be mobilised to assist in overcoming the large overturning moment. The approach adopted was to use the overall foundation area to assess the vertical and lateral stresses applied to separate sections of the base, and check the stability of these in isolation. This approach was considered to be conservative, and doesn’t account for the kinematic constraint provided by the annular shape, which affords additional capacity compared to the individual elements acting in isolation (see for example Fisher & Cathie 2003, Gourvenec & Jensen 2009). After allowing for DNV (1992) partial load and material factors the analytical analyses indicated a minimum reserve capacity of about 5% for the worst load case. 2.2

Finite element analysis

Given the constraints of the analytical analysis, finite element (FE) analysis was used to justify the foundation capacity as it allowed the complex interaction between the annular foundation and non-linear soil to be modeled with greater confidence. FE analysis was undertaken using both 2D and 3D finite element models.Analyses for all design loading cases were performed in 3D using the LS-DYNA FE program (LSTC 2006) whilst the Oasys 2D SAFE FE program (Oasys 2006) was employed for parametric studies (Fig. 4). Details in 3D of the base pontoon, skirts and stiffeners were modeled to provide a realistic structural representation of the foundation and its stiffness (Fig. 4). Vertical loads and moments, depending on their origin, were applied as either a pressure over the foundation base, or on top of the pontoon at the platform leg areas. The total lateral load was distributed evenly over the foundation at seabed level. The 2D model considered a plane strain section through the centre of the foundation in an east-west/ north-south direction. Skirts are modeled as plates

446

Figure 5. SAFE and LS-DYNA results: a) LS-DYNA orthogonal load vertical displacement contours at load factor = 1.6 b) vertical displacement results with increasing load factor.

Figure 4. FE models: a) 3D LS-DYNA b) 2D SAFE.

with equivalent bending stiffness to the stiffened skirts at the centre of a compartment. The base was represented by a continuous beam which was assigned an equivalent stiffness using results of the 3D analyses. The vertical pressure due to the vertical load and overturning moment was applied as a linear distribution across the two pontoons by assuming a rigid annular base. The lateral loads were applied as a horizontal shear stress across the pontoon width assuming the lateral load uniformly distributed over the annular base. In both the 3D and 2D models, (undrained) stability analyses were carried out using a non-linear shear stress-strain model that incorporated the small strain shear modulus (Go ) measured in seismic cone tests and a variation with shear strain inferred from the cyclic analysis for the design storm. The stability runs were undertaken using factored shear stress-strain curves and factored loads.

Figure 6. Indicative details of the MCR-A skirt stiffeners.

in excess of 1.5 are required before excessive displacements occur. As a partial load factor of 1.3 had to be achieved to satisfy code requirements, the FE analysis for this case provides confidence to the finding of the analytical analysis that adequate foundation capacity exists.

2.3 Results of FE modeling An example of the stability results obtained using the 2D and 3D models for a preliminary orthogonal load case is given in Figure 5. The vertical displacements measured at points A and B with the application of increasing environmental load factor are plotted. Similar results are achieved between the 2D and 3D analysis. They show that the foundation on the side of Point B experiences larger vertical displacement and is approaching failure in bearing. Little vertical displacement is occurring at point A on the trailing side of the foundation, which instead experiences more lateral movement such that a sliding type failure is being approached. Figure 5b shows that load factors © 2011 by Taylor & Francis Group, LLC

3

SKIRT INSTALLATION ASSESSMENT

The estimated soil resistance to achieve skirt penetration dictates the maximum suction pressure that needs to be designed for. High installation pressures are expected at the MCR-A site, which proved to be the most onerous design case for the skirt panels inducing high in-plane and out-of-plane loading. This required both horizontal and vertical stiffeners that were arranged to minimise the penetration resistance whilst maintaining structural integrity (Fig. 6).

447

Figure 7. Schematic of penetration resistance with: a) case of full flow-round, b) case of no flow-round, c) soil plug heave.

3.1

Penetration resistance

Due to the complex stiffener arrangement, a key challenge faced was assessing the side friction and end bearing that may be attracted by each of the different skirt panel elements as they penetrate into the clay. Uncertainty was associated with the soil mechanisms that might be occurring during the skirt installation, particularly in relation to the manner in which the soil behaves after coming in contact with the first horizontal stiffener. Above this stiffener, the internal soil may remain self-supporting (Andersen & Jostad 2004), or alternatively, the soil may partially or completely flow around the stiffener, which has also been observed for stiff clay (House & Randolph 2001). These two possible scenarios of full flow-round and non flow-round are illustrated in Figure 7. Two design methods were used to predict the skirt penetration resistance. These were DNV (1992), which uses empirically based coefficients applied to measured cone resistance, and DNV (2005) which utilises the simple shear undrained shear strength. To account for the uncertainties involved, lower-bound, best estimate, and upper bound predictions were made. The resulting maximum suction pressures from these estimates ranged from approximately 50 kPa for a no flowround assumption, to approximately 250 kPa assuming full flow-round. 3.2

Figure 8. Measured vs. predicted penetration resistance calculated for the case of no soil flow-round.

Plug heave

A second challenge associated with the installation was quantifying the soil plug heave within the compartments due to the soil being displaced by the skirt and stiffeners (Fig. 7). Excessive heave could prevent the target penetration depth being reached by the soil surface prematurely coming into contact with the base plate, or causing pumping inlets to clog up. For installation without suction, typically half the skirt wall volume may heave inside the compartment, whilst all the volume displaced by the stiffeners remains inside. For installation using suction, it can be the case that the total structural volume occurs as heave inside the compartment (Andersen & Jostad 2004; Chen & Randolph 2007). The magnitude of heave may © 2011 by Taylor & Francis Group, LLC

be further increased if the soil does not flow around the horizontal stiffeners and instead is pushed upwards, leaving the gaps to be filled with water. For MCR-A, the plug heave assessment was based on a corner compartment as this was most critical having stiffeners on all four walls. The calculation was based on 100% of the displaced soil volume heaving inside the compartment, and allowed for partial water gaps occurring between the horizontal stiffeners. The component of heave resulting from the suction pressure was also estimated using 2D FE analysis. The final heave volume height was predicted to be approximately 300 mm based on a uniform displacement across the compartment. 3.3

Centrifuge testing

Given the challenges posed by the installation, a physical model testing programme was undertaken to examine the effects of the stiffener geometry on the penetration resistance and plug heave in overconsolidated clay. This involved a series of centrifuge installation experiments carried out at the University of Western Australia (UWA) as is described in Westgate et al. (2009). Three separate scale models were fabricated based on a corner compartment of MCR-A. To help investigate the impact of the stiffeners, the Type 1 model had no stiffeners, the Type 2 model had only horizontal stiffeners, whilst the Type 3 model contained the complete arrangement of both horizontal and vertical stiffeners. The results of the penetration resistance measured during the installation of each model type are given in Figure 8. Included in this figure is the predicted resistance calculated based on the assumption that end bearing occurs only on the first (lowermost) horizontal stiffener and does not flow inside the spacing of the stiffeners. It can be seen there is good agreement when this assumption is adopted. This provided evidence that a zero flow-round condition occurs, an assumption which was also supported by visual observations. The plug heave profile was also measured after the installation of each of the model types. As was expected, the magnitude of heave increased with

448

Figure 9. LS-DYNA vertical displacement from spudcan.

increasing model structural volume. For Type 3, representative of the MCR-A, the heave at prototype scale was less than the tolerable value of 500 mm, providing confidence that the degree of plug heave would not pose significant risk to the MCR-A installation.

These findings were in-line with those of Hossain et al. (2004).

4

4.2

Figure 10. Comparison of strengths inferred from T-bar tests.

IMPACT OF SPUDCAN CRATERS

Subsequent to the design of the MCR-A substructure, a jack-up rig was unexpectedly installed for a period in the direct vicinity of the proposed location of the platform (see Fig. 11a). The jack-up spudcans were 18.2 m in diameter and are reported to have penetrated to a depth of approximately 4 m. The disturbed ground conditions remaining after spudcan removal will be to the detriment of the foundation installation and in-place performance. As MCR-A must be linked to adjacent infrastructure, limited scope exists to move the platform location to avoid the disturbance. To support initial planning to resolve this problem, a preliminary investigation was undertaken to quantify the impact of the spudcan craters on the foundation design. In lieu of actual field data after removal of the spudcans, it was necessary to estimate the resulting soil conditions in regard to the extent and nature of the disturbed zone of soil beneath the crater and of the seabed heave. A final assessment is to be made when the results of a future site investigation are available. 4.1 Soil deformation due to the spudcan Soil displacement mechanisms during spudcan penetration in overconsolidated clay were studied experimentally by Hossain et al. (2004). A significant width of soil was observed to move downward with the spudcan whilst soil at the spudcan edge moved laterally and then upward creating surface heave. To enable the actual spudcan shape and soil conditions at the MCR-A site to be considered, 3D FE analysis of spudcan penetration was undertaken using LS-DYNA. The results of the zone of disturbance and heave profiles are shown in Figure 9 for a depth d to diameter D penetration ratio of d/D = 0.2. The zone of soil disturbance was found to have maximum width and depth dimensions of 1.1D (20 m) and 1.2D (22 m) respectively, and a heave profile of maximum width and height of 1.5d (5 m) and 0.45d (1.5 m) respectively. © 2011 by Taylor & Francis Group, LLC

Spudcan impact on soil strength

Experimental studies (e.g. Leung et al. 2007) have shown that spudcan installation and removal can lead to a (short term) reduction by a factor of 2 in the undrained shear strength of normally consolidated clays situated beneath the spud can footprint. To provide an indication of the likely impact of soil strength in higher OCR clays at MCR-A, a centrifuge testing programme was undertaken at UWA. This involved performing 2 T-bar penetration tests (TB1 and TB2) outside the installation area of a square foundation penetrated to a depth d/B = 0.2, and 1 T-bar (TB3) at the centre of the crater after removal of the spudcan. Project constraints meant that the sample clay strength was lower than that of the in-situ material, and that the foundation was not an exact replica of the prototype spudcan. The undrained strengths measured (derived using the T-bar factors given in Lehane et al. (2008) and corrected for shallow penetration effects) are compared in Figure 10 along with estimations using the standard relationship between vertical effective stress (σv
) and OCR. Lower strength is seen below the crater, arising partly because of the lower σv
values in the clay and partly because of a permanent reduction in strength due to the soil disturbance. The latter component leads to a strength that is 88% of the undisturbed strength, whilst allowing for the reduced σv
results in an overall strength in the clay beneath the craters that is approximately 60% of the undisturbed strength. 4.3

Spudcan impact on foundation design

The preliminary assessment of foundation stability accounting for the presence of spudcans was undertaken using similar soil models and procedures as described in Section 2. The spudcan craters were included in both models as is shown in Figure 11. The soil strengths were adjusted in the estimated zone of disturbance based on the interpretation previously

449

5

CONCLUSIONS

The geotechnical design of the MCR-A skirted gravity based foundation involved a number of challenges, several key ones of which have been presented in this paper. In describing the approaches taken to address them, this paper has highlighted how various methods of assessment, such as finite element analysis and physical modeling, can be used to overcome geotechnical design uncertainties. ACKNOWLEDGEMENTS The authors would like to acknowledge colleagues at Arup who assisted with work that has been discussed. Particular mention is made of Anton Pillai who ran the LS-DYNA analyses, and Zack Westgate (formerly of Arup) who among other things undertook skirt penetration assessment.

Figure 11. FE models with craters a) LS-DYNA b) SAFE.

REFERENCES

Figure 12. SAFE and LS-DYNA results with spudcan craters.

described. The 2D model is conservative since the craters are modeled as infinite parallel trenches. An example of the stability results obtained using the 2D and 3D models for an orthogonal load case is given in Figure 12. The vertical displacement due to increasing load factor is provided for the cases of i) no spudcan present, ii) spudcan crater included, and iii) spudcan crater included with reduced soil strength. It is seen that as these cases are imposed, increasing displacement is experienced for a given load factor. For the load case considered, the results qualitatively indicate that a 30% reduction in capacity could be expected based on the 2D SAFE analysis, or a 10% reduction from 3D LS-DYNA analysis. The estimated installation pressures were also reviewed given crater intersection of the foundation footprint, which may inhibit generation of the required suction in the compromised compartments. The heave mounds surrounding the craters have to be compressed by the base plate to penetrate the skirts, further increasing the penetration resistance. These factors meant that the maximum required suction pressure could increase by as much as 50% and complicate achieving a level foundation installation. A detailed assessment of these problems is to be undertaken following the proposed site investigation. © 2011 by Taylor & Francis Group, LLC

Andersen, K.H. & Jostad, H. P. 2004. Shear strength along the inside of suction anchor skirt wall in clay. Proc. Offshore Technology Conf., Houston, Texas, OTC 16844. API 2007. RP2A Recommended practice for planning, designing and construction of fixed offshore platforms. Chen, W. & Randolph, M. F. 2007. External radial stress changes and axial capacity for suction caissons in soft clay, Geotechnique. 57( 6), pp 499–511. DNV 1992. Classification notes No. 30.4, Foundations. DNV 2005. Recommended practice DNV-RP-E303: Geotechnical design and installation of suction anchors in clay. Fisher, R. & Cathie, D. 2003. Optimisation of gravity based design for subsea applications. Proc. Int. Conf. on Foundations, ICOF, Dundee, Scotland. Gourvenec, S. & Jensen, K. 2009. Effect of embedment and spacing of conjoined foundation systems on undrained limit states under general loading. Int. J. Geomech., 9(6) 267–279. Hossain, M.S., Randolph, M. F. & Hu, Y. 2004. Bearing capacity of spudcan foundation on uniform clay during deep penetration. Proc. 23rd Int. Conf. on Offshore Mechanics and Arctic Engineering, Vancouver, Canada. House, A. R. & Randolph, M. F. 2001. Installation and pullout capacity of stiffened suction caissons in cohesive sediments. Proc. 11th Int. Offshore and Polar Eng. Conf., Stavanger, Norway, pp 574–580. ISO 2000. Offshore structures: Part 4: Geotech., ISO 19900. Lehane, B. M., O’Loughlin, C.D., Gaudin, C. & Randolph, M. F. 2008. Rate effects on penetrometer resistance in kaolin, Geotechnique. 59(1), pp 499–511. LSTC 2006. LS-DYNA Theory Manual, Livermore. Leung, C.F., Gan, C.T. & Chow, Y. K. 2007. Shear strength changes within jack-up spudcan footprint. Proc. 7th Int. Conf. on Offshore & Polar Engineering, Lisbon, Portugal. Oasys. 2006. Safe Users Manual, Ove Arup, London, UK. Pinna, R., McRobbie, I., Altraide, B. & Razak, S. 2009. The first gravity based substructure (GBS) for the Caspian Sea. Proc. 4th Offshore Asia Conference, Bangkok, Thailand. Westgate, Z.J., Tapper, L., Lehane, B.M. & Gaudin, C. 2009. Modeling the installation of stiffened caissons in overconsolidated clay Proc. 28th Int. Conf. on Ocean, Offshore & Arctic Engineering, Honolulu, USA.

450

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Constructing breakwater with prefabricated caissons on soft clay S. Yan & X. Feng School of Civil Engineering, Tianjin University, Tianjin, China

J. Chu School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

ABSTRACT: A case history of constructing offshore breakwater on soft clay is presented. The breakwater was constructed near the Shanghai Port, China, for deepening of navigation channel along the Yangtze Estuary. The breakwater elements were designed as gravity retaining structures using prefabricated, semi-circular shaped concrete caissons. Some sections of the breakwater were installed on a thick layer of soft soils. During the construction, the caissons in one section failed under a heavy storm. The causes of failure were investigated by running dynamic triaxial tests on undisturbed soil samples taken from the construction site. It was found that the dike failure was induced by the strength weakening of the soft soil layer below the foundation. The design of the guide dike and the soil improvement works are described in this paper. Surcharge preloading and prefabricated vertical drains was adopted to improve the soft soils below the caisson. The soil improvement measure was proven to be effective in maintaining the stability of the breakwater against subsequent heavy storms.

1

INTRODUCTION

piece and thus make the subsequent repair or rehabilitation works relatively easier. (4) The construction process is relatively simple and speedy, as no on-site concrete casting is required.

Guide dike, literally, is the guide for navigation. And it is also constructed to reduce the effects of wave and prevent future sedimentation within the channels. As part of the Yangtze Estuary Development Project, a breakwater was constructed 40 km away from a busy port in Shanghai, China. The guide dike was approximately 50 km. The water depth ranged from 7.0 to 8.5 m. The design wave height was 3.32 to 5.90 m with a return period of 25 years. The breakwater was designed as gravity retaining structures using prefabricated, semi-circular shaped concrete caissons to resist the rough waves (Yan, 2005). The prefabricated, semi-circular shaped concrete caissons for guide dike construction were used in the Channel Deepening Project for the Yangtze Estuary, which has the following advantages: (1) All the wave and water pressures acting on the semi-circular shaped surface pass through the centre of the circle. Thus, the overturning moment becomes very small and the vertical pressure distribution at the base of the caisson becomes more uniform (Jia, 1999). (2) Study shows that the wave force acting on a semicircular shaped guide dike is smaller than that on a vertical guide dike (Xie, 1999). The internal stress induced by the load applied is relatively small for an arch structure and thus the costs for the construction of the breakwater can be reduced. (3) The guide dike is made of prefabricated segments which can be dismantled or replaced piece by © 2011 by Taylor & Francis Group, LLC

However, as the technique is relatively new, there is not sufficient experience on how to treat the foundation soil properly and yet economically when the caissons are to be placed on soft seabed soils. On one hand, these caissons have to be tall and heavy enough to match the water depth and provide stability against the wave forces. On the other hand, when the caissons are too heavy, they cause settlement, bearing capacity, or stability problems (Thiers and Seed, 1969;Yasuhara, 1988). This is particularly the case when the foundation soil is very soft. Treating seabed soft soils offshore in relatively deep water over a large area is difficult and costly. Therefore, it became a challenge on how to construct these large size gravity structures on soft soil in a cost-effective way. Based on the theory analysis and test research, soil improvement work was put forward on the soft soil below the caisson dike to prevent the soft soil from heavily weakening. These measures were proven to be effective in maintaining the stability of the guide dikes against subsequent heavy storms. 2

SOIL CONDITIONS

The typical soil profile of the soil below the guide dike is shown in Figure 1. It consisted of 1.5 m to 3.5 m thick silty sand followed by 2 m to 4 m thick muddy clay and approximately 30 m thick soft clay

451

Figure 2. Cross-section of the prefabricated, semi-circular shaped concrete caisson.

Figure 1. Soil profile beneath seabed soil along the navigation channel. Table 1.

Soil stratum Silty sand Muddy clay Soft clay

Soil properties. Water content %

Unit weight kN/m3

Void ratio –

Liquid limit %

Plastic limit %

29.3

19.0

0.827

–

–

57.5

16.4

1.672

45.2

23.6

51.5

16.8

1.470

45.8

23.8

Figure 3. Failed breakwater after a heavy storm.

underlying the muddy clay. The basic properties of the soils are given in Table 1. The silty sand was very fine with a mean grain size of 0.15 mm. Although the soil below the guide dike in this zone was weak, the initial design was not to treat the seabed soft soil, but to use an adequate strong rubble mount to support the caissons. This approach was based on the following four considerations: 1) It would be very expansive to treat the seabed soil offshore in relatively deep water; 2) The guide dike could tolerate relatively large settlement, as it would be used mainly as a divider; 3) As the guide dike was a strip load, the surcharge load would only be distributed to a certain depth. Under the surcharge load of the rubble mount, the geotechnical properties of the muddy clay near the top few meters would be improved with time, and thus the stability of the guide dike would be improved with time; 4) Similar design had been adopted for other projects with similar site conditions and no failure had occurred before. Such as the breakwater constructed in Miyazaki Harbor, Japan (Sasajima, et al., 1994; Tanimoto and Takahashi, 1994).

The base of each concrete segment was 17 m wide and 19.94 m long. The hollow caisson would be filled with sand after installation through a 600 mm diameter hole on top of the caisson. In order to enhance the stability of caisson and prevent the foundation soil from scouring, a layer of geotextile sheet and sand filled tube composite was placed onto the seabed to enhance the stability of the rubble mount. The sand tubes were 300 mm in diameter and spaced 500 mm apart at the edge and 1000 mm in the centre, as shown in Figure 2. Other detail of the sand tube will be described at a later section. A sand cushion of 700 mm thick was placed on top of the geotextile and sand tube composite layer before another getextile and sand tube composite layer was placed on top of the sand cushion. This was topped by crushed stones of 1∼100 kg in the centre and 200∼400 kg at the edge, which were used to form a rubble mount platform at an elevation of −3.60 m, as shown in Figure 2. The caisson was placed on top of this rubble mount. Berms were also placed on two sides of the caisson to enhance the stability of the guide dike. The berms were 1.7 m thick and made of 400 ∼ 600 kg crushed stones. The hollow caissons were then filled with sand to an elevation of −0.8 m.

4 4.1

3

ORIGINAL DESIGN OF THE GUIDE DIKE

The design of the guide dike is schematically shown in Figure 2. The caisson used was prefabricated, semicircular shaped concrete hollow segment, as shown in Figure 2. The radius of the semicircle was 5.7 m. © 2011 by Taylor & Francis Group, LLC

FAILURE CASE Failure description

The construction for the guide dike was completed in Oct 2002.A heavy storm took place in December 2002. During the storm, 16 segments of the caissons failed. Large settlements occurred suddenly and some caissons also moved laterally. A picture of the guide dike after the storm is shown in Figure 3. The maximum

452

Figure 4. Settlement of #1 caisson and wave height versus time curves.

settlement was 5.78 m. The maximum lateral movement was more than 20 m. The settlement versus time curve measured for caisson #1 is shown in Figure 4. The wave height versus time curve is also plotted in Figure 4. It can be seen that a sudden settlement took place at the time when there was a surge in the wave height on 6 December 2002. The guide dike had failed under the wave action. To understand the causes of failure, dynamic triaxial tests on the muddy and soft clay were carried out. 4.2 Dynamic triaxial tests on the muddy clay and soft clay 4.2.1 General It was understood that the muddy clay below the caissons was in a most critical state as the surcharge loads was just applied for 2 months. Furthermore, it was identified that the following two factors had also contributed to the failure. The first is the additional surcharge load due to the dynamic action of the wave. The second is the softening of soft soil below caissons; that is, a reduction in the undrained shear strength of the clay, under the wave loads. The possibilities of liquefaction of the top silty sand layer and scour at the base of the caissons were ruled out as the causes after the investigation. Based on elasticity theory, the vertical effective stress distributions due to both static and dynamic conditions are calculated as shown in Figure 5. The ratio of surcharge load increment due to the wave action, σv , to the surcharge load under static condition, σv , β = σv /σv , is also calculated and plotted in Figure 5. It can be seen that under the wave action, the vertical overburden stress could increase by 15% due to the wave load. However, this effect becomes much less significant with increasing in depth. 4.2.2 Load path and test equipment After the failure, some undisturbed soil samples were taken from the muddy clay layer and cyclic triaxial tests were conducted (Andersen, et al., 1988; Andersen, and Lauritzsen, 1988). The loading path for the soil samples is shown in Figure 6. Point A is the initial stress of the soil element. Path AB represents the static loading caused by the rubble mount and the caisson © 2011 by Taylor & Francis Group, LLC

453

Figure 5. Increase in surcharge load due to wave loads.

Figure 6. Load path before soil improvement.

and path BC denotes the action of the cyclic loading caused by the wave force. Where, τf -shear stress at failure; K0 -consolidation coefficient. τ-shear stress; σc
-confining stress.

Table 2.

Properties of soil sample SY3-4.

Soil type

Water content %

Unit weight kN/m3

Sample diameter cm

Sample height cm

Muddy clay

53

17.3

5.0

10.0

Table 3. Test conditions and results of soil sample SY3-4. Cell Static Dynamic Initial Residual Reduction pressure load load strength strength Ratio kPa kPa kPa kPa kPa – 30

20

8

15

10.8

0.72

The tests were performed with the HX-100 dynamic triaxial test equipment. According to the loading path, the sample was first consolidated with the in place initial stresses by self weight. Then the static load was applied in undrained condition. When the displacement of the soil sample was stable, the application of the cyclic load was applied also in undrained condition. The combinations of the applied static load and dynamic load were determined according to the FE analysis results before and after soil improvement (see the later sections). 4.2.3 Test results The triaxial dynamic test was performed on an undisturbed soil sample SY3-4 at depth 3.0–3.5 m. The details of the soil sample, typical test conditions and results are listed in Table 2 and Table 3. The dynamic stress was applied and repeated for 1000 times (with a period of 6 seconds) on the soil sample. The pore pressure (P.P) development is recorded and shown in Figure 7a; the accumulated strain developing with the number of cycles in Figure 7b. The strength of the soil samples were tested before and after the dynamic tests and shown in Figure 8. A total of 25 soil samples were tested in the same way. The strength reduction coefficients are summarized and given in Figure 9. In this figure, both the additional static stress and dynamic are normalized by the confining pressure. Therefore, the strength reduction ratio can be determined with different combinations of σc
, σj and σd . 5

SOIL IMPROVEMENT

Figure 8. Shear test results before and after dynamic loading.

Figure 9. The strength reduction ratio.

To ensure the stability of the guide dike against future storms, a soil improvement scheme was adopted to improve the soft soil layer below the caissons, which was 7.0 m to 8.5 m below the sea level. Several soil improvement methods were considered. The method to accelerate the consolidation process of muddy clay using prefabricated vertical drains (PVDs) was considered the most economical option. This method was © 2011 by Taylor & Francis Group, LLC

Figure 7. Typical dynamic triaxial test results.

454

also relatively easy to implement. The influence zone of the strip load was estimated to be 7.0 m below the seabed. Therefore, it was only necessary to install the PVDs to a depth of 10 m deep. The weight of the rubble mound was considered sufficient in providing surcharge. The PVDs were installed offshore from a specially designed PVD installation barge. There were

Figure 10. The revised design profile.

Figure 11. The constructed semi-circular concrete breakwater.

12 drain installation rigs on each barge with preset spacing of 1 m in square grid. The position of each drain was located using GPS. Each barge could carry 12 drain installation rigs and could install an average of 1185 drains per day. The procedure for improving the soft soil before the placement of the rubble mount and the caissons was as follows: (1) PVDs were installed from the PVD installation barge at a spacing of 1.0 m in square grid to a depth of 10 m below the seabed; (2) A geotextile and precast concrete block composite was used to cover the seabed around the toe of rubble mount that face the open sea for the prevention of scour. The concrete blocks were 400 mm × 400 mm in square and 160 mm thick. They were attached to the geotextile sheet to form a composite. For the rest, the geotextile and sand tube composite was installed on the seabed. Sand was filled into the 300 mm diameter geotextile tubes from a barge on site before the geotextile and sand tube composite was placed. The tubes were spaced at every 500 mm at the toe and 1000 mm at the centre of the rubble mount; (3) The 700 mm sand cushion was placed before the second layer of geotextile and sand tube composite was laid on top; (4) The crushed stones were laid on to the second geotextile and sand tube composite from a barge which also acted as a surcharge to consolidate the soil below; (5) The caisson segments were only placed after an average degree of consolidation of 80% was achieved, which took about 90 days after the placement of the rubble mound. The adopted soil improvement method has the following advantages: (1) The soft soil can be strengthened to increase the ability of resisting dynamic loads; (2) After consolidation, the weight of the foundation consisting of crushed stones becomes part of the consolidation stress, instead of part of the structure load. This means that after soil improvement, the soil becomes stronger and the static load becomes smaller. As a result, the consolidation stress increases and the vertical subsidiary stress caused by the upper structure decreases. Referring to Figure 9, the strength reduction ratio will increase obviously, so the constructed dike will become much safer. Using the soil improve method, the entire caissons of the navigation guide dike were completed in Dec © 2011 by Taylor & Francis Group, LLC

2003. The guide dike had experienced several strong storms caused by typhoons. Some of them were even stronger than the one that caused the failure. In spite of this, the guide dike was stable and there was no additional settlement caused by the storms. The total settlements were within the expected range. Therefore, the use of PVDs has proven to be a successful method for this project. The guide dike after construction is shown in Figure 11.

6

CONCLUSIONS

A case study of using prefabricated, semi-circular shaped concrete caissons to construct offshore breakwater on soft soil was presented in this paper. A failure case was also investigated by running dynamic triaxial tests on undisturbed soil samples taken from the construction site. It was found that the dike failure was induced by the strength weakening of the soft soil layer below the foundation. It is learned through dynamic triaxial tests that, under a heavy storm, the strength of the soft soil can be seriously reduced. This factor should be taken into consideration in the design. Surcharge preloading plus PVDs was used to improve the soft seabed soil before the installation of the guide dike. With the use of PVDs, the soft soil was consolidated much faster and gained sufficient strength quickly to maintain the stability of the caissons. This scheme has proven to be effective as the dike has experienced several storms since construction without any problems. REFERENCES Andersen, K. H., Kleven, A. & Heien, D. 1988., “Cyclic soil data for design of gravity structures”, Journal of Geotechnical Engineering, Vol: 114(5), pp.517–539. Andersen, K. H. & Lauritzsen, R., 1988, “Bering capacity for foundations with cyclic loads”, Journal of Geotechnical Engineering, Vol:114(5), pp.540–555. Jia, D.H., 1999, “Study on the interaction of water waves with semi-circular shaped guide dike”, China Ocean Engineering, Vol:13(1), pp.73–80.

455

Sasajima, H., Koizuka, T., Sasayama, H., Niidome,Y. & Fujimoto, T., 1994, “Field demonstration test on semi-circular shaped guide dike” Proceedings of the International Conference on Hydro-Technical engineering for Port and Harbor Construction, Yokosuka, Japan, (1), pp.593–615. Tanimoto, K., & Takahashi, S, 1994, “Japanese experiences on composite breakwater”. Proceedings of international workshop on wave barriers in Deep Waters, Yokosuka, Japan, pp.1–22. Thiers, G. R. & Seed, H. B., 1969, “Strength and stress-strain characteristics of clays subjected to seismic loading conditions”, Vibration Effects of Earthquakes on Soils and

© 2011 by Taylor & Francis Group, LLC

Foundations, Selig, E.T. ed., ASTM Special publication STP450. Xie, S. L., 1999, “Wave forces on submerged semi-circular shaped guide dike and similar structures”, China Ocean Engineering, Vol:13(1), pp.63–72. Yan, S.W., 2005, “Dynamic cyclic triaxial tests on undisturbed soft clay samples taken along the Yangtze Estuary waterway”, Project Report, Geotechnical Research Institute, Tianjin University, 2005 (in Chinese). Yasuhara, K., 1988, “Cyclic strength and deformation of normally consolidated clay”, Soils and Foundations Vol:22(3), pp.373–381.

456

6 Piled foundations

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Simplified analysis of laterally loaded pile groups F.M. Abdrabbo & K.E. Gaaver Structural Engineering Department, Faculty of Engineering, Alexandria University, Egypt

ABSTRACT: The response of laterally loaded pile groups is a complicated soil-structure interaction problem. Although fairly reliable methods are developed to predicate the lateral response of single piles, the lateral response of pile groups has attracted less attention due to the required high cost and complication implication. This study presents a simplified method to analyze laterally loaded pile groups. The proposed method implements p-multiplier factors in combination with the horizontal modulus of subgrade reaction. Shadowing effects in closely spaced piles in a group were taken into consideration. It is proven that laterally loaded piles in sand can be analyzed within the working load range assuming a linear relationship between lateral load and lateral displacement. The proposed method estimates the distributions of lateral loads among individual piles in a pile group and predicts the safe design lateral load of a pile group. The benefit of the proposed method is its simplicity for the preliminary design stage.

1

INTRODUCTION

The lateral response of pile foundations is critically important in the design of structures that may be subjected to lateral loads. It is worth noting that, lateral loads are in the order of 10%–15% of the vertical loads in the case of onshore structures, while this value may exceed 30% in case of offshore structures (Rao et al. 1998). The response of a laterally loaded pile is a complicated soil-structure interaction problem, because the pile deflection depends on the soil reaction and the soil reaction in turn depends on the pile deflection. Fairly reliable methods have been developed for predicting the lateral response of single piles, since the pioneer works of Matlock and Reese (1961), and Broms (1964). Frechette et al. (2002) reviewed the design methods for laterally loaded groups of drilled shafts and compared between methods employing a group reduction factor and a p-multiplier. Kumar and Lalvani (2004) analyzed the nonlinear load-deflection behavior of laterally loaded piles using p-y relationships. Full scale and centrifuge model tests on pile groups have been conducted by Brown et al. (1988), McVay et al. (1998), and Rollins et al. (2005). Laterally loaded pile groups may be analyzed using the elastic continuum approach (Poulos and Davis 1980), and the group equivalent pile procedure (Ooi et al. 2004). The p-y relationships, initially developed by Matlock (1970), have been used to model the pilesoil interaction, Reese et al. (1974). As a result of the interaction between piles in a group, the p-y relationship of single pile was modified to be implemented in pile group analysis. The modifications can be carried out by introducing p-multiplier, Ooi et al. (2004), and Rollins et al. (2005). The p-multiplier concept is an effective procedure for implementing in the pile group © 2011 by Taylor & Francis Group, LLC

analysis; nevertheless unique values of p-multiplier for a pile group are not standardized. It is worth noting that the p-y relationship is not a soil property, but rather pile-soil property, Ashour and Norris (2003). In recent years, several simplified approaches for the analysis of laterally loaded single piles or pile groups have been developed that can be used with little computational effort, Liyanapathirana & Poulos (2005), and Castelli & Maugeri (2009). This paper presents a simplified method, for analyzing laterally-loaded pile groups, using p-multipliers in combination with Winkler’s model. 2

GEOTECHNICAL DATA OF THE SITE

Before discussion of the proposed method, obtained geotechnical data, where laterally loading tests on vertical single piles were conducted, are presented. The proposed method was implemented to analyze pile groups constructed at this site. The site is located at the Northeast of Nile River Delta, Demiatta free zone district, Egypt. The soil profile at the site consists of a top layer of medium dense sand. N60% varies from 40 near ground surface to 20 at a depth of 10 m. Fine silty sand layer was encountered at a depth 10 m, and extended to a depth of 15 m below ground surface. This layer is underlain by a thick layer of soft to medium normally consolidated clay, which is extending up to a depth of 36 m. At a depth of 36 m, a very dense sand bed was encountered and explored to a depth of 60 m. The top sand layer has natural unit weight of 19 kN/m3 , angle of internal friction 37◦ , and relative density of 70%. These values were interpreted from the average value of SPT (N60% ) throughout the sand layer. For soft clay, the natural unit weight varies

459

between 13.1 and 16.7 kN/m3 , and the undrained shear strength varies from 6 to 20 kN/m2 . The values of unit weight and undrained shear strength increased with depth throughout the clay layer. For medium clay, the natural unit weight varies from 17.4 to 18.7 kN/m3 , and the undrained shear strength varies from 15 to 45 kN/m2 . For sand bed, the peak angle of shearing resistance is 45◦ , while the residual value is 36◦ . The shear strength of clay and sand bed was measured using the triaxial test apparatus. The ground water table is at 1.0 m below ground surface. Bored piles, of 600 mm diameter, were constructed to be seated at a depth 40 m below the existing ground surface. The pile is reinforced by nine bars of 18 mm diameter of steel grade 36/52 and the reinforcement is extending to 16 m below the ground surface. The piles are attached to a pile cap which is resting on ground surface. As the pile is considered a flexible pile, the safe design lateral load of the pile depends on structural capacity of the pile cross section and the allowable lateral deflection at pile head. Based on these design criteria, the safe design lateral load of single pile is 80 kN, dominated by structural capacity of the pile cross section. 3 THE PROPOSED METHOD The aim of the proposed method is to estimate the distributions of lateral load acting on a pile group among the piles in the group. The piles in the group are considered flexible piles. More likely, flexible piles in a group are embedded in a stratified soil, and hence the lateral load may be resisted by soil lateral stresses developed along the top portion of the pile, which is called the effective length (Lef ). One method of assessing the value of (Lef ) is by modeling a single pile as a beam in a soil represents by an elastic uncoupled spring modulus. Lef is assessed as the depth where the lateral deflection of the pile is effectively zero. But the effective depth of single pile differs from the effective depth of a pile in a group, due to pile-soil interaction. The effective length (Lef ) of a pile in a group was calculated by re-analyzing single pile but with softer springs. These spring moduli were obtained by multiplying the spring modulus of the single pile by p-multiplier values. In this analysis, the horizontal subgrade reaction (Kx(s) ) at a depth (Z) below the ground surface within the top sand layer is expressed as; Kx(s) = ηh .Z, where ηh is the modulus of horizontal subgrade reaction at top sand layer. (Kx(s) ) is expressed in units of force per unit area, while (ηh ) is expressed in units of force per unit volume. A constant value of horizontal subgrade reaction along the pile through the clay layer (Kx(c) ) is considered. It is important to note that the modulus of horizontal subgrade reaction is not a unique soil property, but depends on pile characteristics and the lateral displacement of the pile. Using the aforementioned soil profile and soil properties at Demiatta free zone district, the effective depth of single pile and for a pile within a group was founded to be less than the depth of the top sand layer and equal to about 16 times

© 2011 by Taylor & Francis Group, LLC

the pile diameter. In this study, ηh was considered for single pile equal to 16.346 MN/m3 , for medium sand. The relative stiffness factor (T) and the maximum value of depth coefficient (Zmax ) were calculated as; T = (E.I/ηh )0.2 , and Zmax = (Lef /T), where E.I is the flexural rigidity of the pile as un-cracked section. The maximum value of depth coefficient (Zmax ) was found to be 5, which means that the pile is flexible pile. The dimensionless relationships, developed by Reese and Matlock (1956), were used to determine the distribution of pile displacements, bending moments, shearing forces, soil resistances, and slope deflections along the effective length of a single pile due to the safe design lateral load of 80 kN applied at the pile head assuming fixed head piles, without any free length above ground surface. For the single pile, ηh was implemented directly, while for a pile within a group ηh was reduced due to shadowing effects. The shadowing effect depends upon the location of pile row within the group and the location of the pile within the row. McVay et al. (1998) concluded that in the same pile row, the middle pile develops slightly less lateral resistance than the side piles because it is subjected to more substantial shadow effects. However, the authors showed that the difference is not significant and no significant error is developed by assuming that all piles in the same row carry the same lateral load. Consequently, the multiplier factor (p) for all piles within a row was assumed to be the same value. To consider the effects of pile-soil-pile interaction in a group, a pile within a group was analyzed as a single pile and the distribution of pile deflections, bending moments, shearing forces, lateral soil resistances, and slope deflections along the pile were assessed for different values of ηhp where the reduced value of horizontal subgrade reaction (ηhp ) was obtained from; ηhp = p.ηh . The values of p-multiplier factors were obtained from McVay et al. (1998), and Ooi et al. (2004). As a result, a data base containing pile deflections, bending moments, shearing forces, lateral soil resistances, and slope deflections were formed for a single pile embedded in fictitious sand of different (ηhp ) values and subjected to different lateral loads. The data base was formed with the help of computer spreadsheets.

460

4 ANALYSIS METHODOLOGY Once the data base was compiled, the analysis of a laterally loaded pile group can be carried out. In the first case study, a pile group configuration containing n-rows of piles and subjected to certain lateral load (PH ) at the ground surface is studied. The piles in the group are attached to rigid pile cap, that is to say the lateral displacements of all piles in the group at their heads are equal. The properties and reinforcement of the piles are mentioned in section (2). The unknowns in this case study are the load distribution among piles in the group and the lateral displacement of the group at ground surface. The piles are considered long flexible piles. The relationship between lateral applied

load (PH ) and lateral displacement at pile head (yG ) is assumed to be linear. The p-multiplier for each row was assessed from documented literature assuming the leading row has the bigger p-multiplier while the trailing row has the smaller p-multiplier. The p-multiplier depends on the number of rows in the pile group, and location of the row in the group. Entering an assumed value of lateral displacement (yG ) and the p-multiplier into the data base, the lateral load acting on a pile (Pi ) in each row corresponding to the assumed lateral displacement of the group (yG ) and the given specified p-multiplier was obtained. If the sum of pile lateral loads (Pi ) is equal to the applied lateral load (PH ) acting on the pile group, the solution is obtained and the process is terminated. But if the sum of pile lateral loads (Pi ) differs from the applied lateral load (PH ), the assumed pile group displacement needs to be altered and the procedure continues in a trial and error process until the equilibrium between the sum of pile loads and the applied total load achieved, that is to say (Pi ) = (PH ). The assumed lateral displacement at equilibrium condition is the lateral displacement of the pile group. Once the equilibrium condition is achieved, the distribution of bending moment, shear force, and soil pressure in each pile in the group can be obtained from the data base. In the second case study, the number of piles in a group is known and the safe design lateral load of the single pile is also known. It is required to determine the safe design lateral load of the pile group. The properties and reinforcement of the piles are mentioned in section (2). This case study represents a practical case in which the safe design lateral load of single pile is evaluated and verified by field loading tests. Usually the pile group configurations are assessed by knowing the vertical applied loads, vertical working load of single pile, and group efficiency. Once the pile groups are arranged, the capability of pile groups to sustain the lateral loads safely becomes essential. To tackle this problem, the value of p-multiplier for each row in the group was assessed. Returning back to the compiled data base, the pile head displacement (y1 ) of the piles in the leading row was obtained corresponding to horizontal subgrade reaction of (p1 .ηhp ) and lateral applied load equal to design lateral load of the single pile. At the same displacement (y1 ) and horizontal subgrade reaction of (p2 .ηhp ), the pile load for the second row is obtained. The procedure is repeated for all rows in the group. Then the design lateral load of a pile group is equal to the sum of individual pile loads within the pile group. The only drawback in the proposed method is that the load of piles in the leading row is assumed equal to the design lateral load of single pile but with a corresponding bigger displacement compared to the displacement of individual pile. The p-multiplier of leading row varies from 0.75 to 1.00, Rollins et al. (2005), and from 0.65 to 1.00, Ooi et al. (2004). Truly at the same lateral displacement of a pile group and single individual pile, the lateral load applied on the leading row in the group is smaller than single individual pile.

© 2011 by Taylor & Francis Group, LLC

Figure 1. Pile arrangement, pile diameter = 0.60 m and length = 40 m.

5

NUMRICAL EXAMPLE

Consider a group of three piles, each of 600 mm diameter installed in one row. This arrangement represents 3 × n-pile groups, where n = 1, 2, 3, etc, figure (1). Single vertical pile was analyzed using curves, developed by Reese and Matlock (1956), under a lateral load of 80 kN. Fixed head pile was assumed, the resulted pile head displacement is 1.95 mm. It is required to determine the lateral load acting on each pile in the group under acting a lateral load of 240 kN. The analysis was started by assuming the p-multipliers as 0.8, 0.4 and 0.3 for leading pile, middle pile, and trailing pile respectively. These values were obtained from McVay et al. (1998). Dodagoudar et al. (2010) reported the values of p-multipliers published by their study, Rollins et al. (1998), Brown et al. (1987), Ilyas et al. (2004), Reese et al. (2006), and Mokwa and Duncan (2005). The reported data are for piles in all soil types. The p-multiplier for leading row varies from 0.60 to 0.93 with an average value of 0.79. The p-multiplier for second row varies from 0.40 to 0.78 with an average value of 0.58. For the third row, the p-multiplier varies from 0.40 to 0.63 with an average value of 0.46. For the fourth row, trailing row, the p-multiplier varies from 0.40 to 0.68 with an average value of 0.52. The reported values by Dodagoudar et al. (2010) excluded values published by McVay et al. (1998), which the present analysis was based. Also it is worth noting that the average value of p-multiplier reported by Dodagoudar et al. (2010) for the fourth row is bigger than the third row. From the compiled data base, the pile head lateral displacements under an applied lateral load of 80 kN and values of ηhp equal to 0.8ηh , 0.4ηh and 0.3ηh are 2.23, 3.40, and 3.98 mm for leading pile, middle pile, and trailing pile respectively. These lateral displacements violate the boundary conditions at the pile heads. Therefore by assuming the lateral deflection of the pile group at ground surface is 3.00 mm and each pile in the group exhibits this deflection, the lateral loads of leading pile, middle pile, and trailing pile shall be 107.62 kN, 70.58 kN, and 60.3 kN respectively. At this stage, the sum of the lateral loads of the piles in the group is 238.5 kN which is less than the applied lateral load of 240 kN. So the pile group displacement should be adjusted to be 3.019 mm to match with the applied load of 240 kN. The process was repeated and the corresponding loads acting on leading, middle, and trailing piles become 108.28 kN, 71.04 kN and 60.68 kN respectively. The sum of the pile loads in the group becomes equal to the applied lateral load. The corresponding lateral displacement

461

of the group is 1.548 times the lateral displacement of the single pile. It is worth noting that 9-pile group arranged in square pattern exhibits the same lateral displacement under an acting lateral load of three times of 240 kN. The shadowing approach in which the piles in a row have no interaction effects on piles outside this row is contradicted with the elastic approach. In the elastic approach, the interaction factor between two piles depends upon the angle in plan between the centers of these two piles among other factors, Poulos and Davis (1980). To estimate the safe design load of (3 × 1) pile group, the lateral load of the leading pile was assumed equal to the safe design lateral load of the single pile, which is 80 kN. From the compiled data base and introducing p-multiplier of 0.80, the corresponding lateral displacement at the pile head of the leading pile is 2.23 mm. By enforcing the piles in the group to exhibit the same deflection, thus the lateral loads of the middle pile and the trailing pile become equal to 52.49 kN and 44.83 kN respectively. These values were obtained from the data base using p-multipliers for middle and trailing rows. The maximum bending moments induced in leading, middle, and trailing piles under lateral loads of 80, 52.49, and 44.83 kN respectively and corresponding to p-multipliers of 0.80, 0.40, and 0.30 were obtained from the compiled data base. The structural capacity, expressed as the bending capacity of the pile cross section, was calculated and compared with induced values. It was found that the pile cross section is capable to resist the induced moments safely. If the pile cross section is incapable to resist the induced bending moment, the process is repeated but with a small value of lateral load on the leading pile. In flexible piles, the dominant factor in assessing the safe load of single pile and pile group is the structural capacity of piles. Consequently the safe design lateral load of the pile group is 177.32 kN. Nine-pile group arranged in square pattern carries three times the achieved value of lateral load, at the same value of lateral displacement. The corresponding group reduction factor is 0.739, compared by 0.67 that was reported by Frechette et al. (2002). This analysis indicated that the leading pile carries 45.1% of the applied lateral load acting on the pile group, while the middle and trailing piles carry 29.6% and 25.3% respectively. The corresponding lateral displacement of the group is 1.143 times the lateral displacement of a single pile. McVay et al. (1998) conducted lateral tests on pile groups founded in sand in a centrifuge machine. Their results indicated that, for 3 × 3 pile group the percentage of lateral load carried by lead, second, and trail rows were 43.3%, 31.5%, and 25.2% respectively in case of dense sand. For loose sand, the corresponding values were 46.6%, 29.3%, and 24.1% respectively. A comparison between the results of pile load distribution obtained by simplified method and the measured values revealed that results of the proposed method are in good agreement with the experimental results. The only shortcoming of the proposed method is that the effect of spacing between piles in the group is not considered. However, from

© 2011 by Taylor & Francis Group, LLC

Figure 2. Lateral load versus lateral displacement, test No. 1.

Figure 3. Lateral load versus displacement, tests No. 2 & 3.

a practical point of view, most of designers prefer to arrange the piles at minimum spacing in a group in order to minimize the size of the pile cap. Thus the proposed method is suitable to be implemented for pile groups having practical spacing of 2.5 to 3 times the pile diameter. In this situation it is important to note that the effect of spacing between piles in a group can be considered in the analysis if p-multiplier values for pile groups of different spacing to diameter ratios are developed. 6

JUSTIFICATION OF THE ASSUMPTIONS

The proposed method is based on linear relationship between lateral load and lateral displacement at pile head, which was confirmed by McVay et al. (1998), Yang and Liang (2006), Gaaver (2006), and field test results presented in figures (2) and (3). Figure (2) presents results of a test pile of 600 mm diameter and 40 m length installed at the site located at Northeast side of Nile River delta, having the same succession of soil strata as given before in section (2). Figure (2) illustrates a good agreement between theoretical p-y relationship and the experimental values up to the design lateral load of 80 kN. Two field pile loading tests were conducted on two individual piles at a site nearby Alexandria city, Egypt. The piles are of 500 mm diameter and 13 m depth below the ground surface. The lateral load was applied at the ground surface. Retrieved soil samples from boreholes indicated that the explored site consists of a top layer of sandy silty clay up to 9.5 m depth.The top layer overlies sand stone to 20 m depth. The top layer has a natural unit weight of 18 kN/m3 , undrained shear strength of 23 kN/m2 , and an angle of internal friction of 21◦ , measured using direct shear box apparatus. Figure (3) presents

462

Figure 4b. Comparison between measured and computed bending moments of a single pile, Ph = 89 kN, and ηh = 53 MN/m3 .

Figure 4a. Comparison between measured and computed bending moments of a single pile, Ph = 24 kN, and ηh = 53 MN/m3 .

the achieved test results. The safe design lateral load of the pile is 60 kN and the test load 120 kN. Cleary test (2) demonstrates that the relationship is linear up to 100 kN. At the same time, test (3) shows that the relationship is linear, without any appreciable residual displacement. Therefore linear analysis of laterally loaded piles is considered as a good simulation of real behavior of piles under lateral loads within the working load range. It is worth noting that the tested piles were constructed by boring the soil and cast in situ concrete. Therefore, there is a complete contact between the formed pile and the surrounding soil especially near ground surface. The gap that may be formed near ground surface between the pile and the surrounding soil during pile construction as well as the nonlinearity of soil stiffness are the main causes of nonlinearity response of a laterally loaded pile at small values of lateral loads. The proposed method was also based on that the lateral resistance of a pile in a group is a function of row location belonging to that pile within the group, rather than location within a row, contrary to expectation based on the elastic theory. Rollins et al. (2005) and McVay et al. (1998) confirmed the above assumptions. Validation of the proposed method is presented in figures (4a) and (4b). The horizontal subgrade reaction was considered to be increased linearly with depth, from zero at ground surface to (ηh ) at depth 10 m below ground surface. The selected value of horizontal subgrade reaction was used along with the compiled data base to predict the distribution of bending moment along the single individual pile. A comparison between the obtained distribution of bending moment and the measured values by Rollins et al. (2005) indicated that the selected value of horizontal subgrade reaction overestimated the maximum bending moment induced in the single pile by up to 17%, while LPILE (Reese et al. 1997) and SWM (Ashour et al. 2002) methods underestimated the induced values by up to 20%. Cleary the horizontal subgrade reaction can be used for pile

© 2011 by Taylor & Francis Group, LLC

Figure 5a. Bending moments versus p-multiplier.

Figure 5b. Lateral displacement at pile head versus p-multiplier.

group analysis. The induced bending moment in a pile within a group depends upon the location of the pile in the group. According to the proposed method, the distribution of the bending moment can be obtained by analyzing single individual pile using softening modulus of subgrade reaction that can be obtained by multiplying p-multiplier by (ηhp ). The effects of p-multiplier on induced bending moment and the lateral displacement at pile head are shown in figures (5a) and (5b). The applied lateral load at the pile head is 80 kN, while the horizontal subgrade reaction is increasing with depth from zero at ground surface to a value of 163 MN/m3 at depth 10 m below ground surface. The pile diameter is 600 mm. As p-multiplier decreased, the soil gets soft and consequently the pile head deflection and the bending moment increased.

463

7

CONCLUSIONS

The effective length of a long flexible laterally loaded pile is equivalent to about 16 times the pile diameter, for pile embedded in sandy soil. Laterally loaded piles in sand can be analyzed within the working load range assuming a linear relationship between lateral load and lateral displacement at pile head. The induced maximum bending moments and lateral displacements at pile head of laterally loaded piles decreased linearly as the values of p-multiplier increased. The paper presents a simplified method for the analysis of pile groups subjected to lateral loads.The proposed method estimates the distributions of lateral loads among piles in a group and predicts the safe design lateral load of a pile group. REFERENCES Ashour, M. & Norris, G. 2003. Lateral loaded pile response in liquefiable soil, J. Geotechnical & Geoenv. Eng., 129 (5): 404–414. Ashour, M., Norris, G., & Pilling, P. 2002. Strain wedge model capability of analyzing behavior of lateral loaded isolated piles, drilled shafts, and pile groups, J. Bridge Eng., 7 (4): 245–254. Brown, A., Reese, C. & O’Neill, W. 1987. Cyclic lateral loading of a large-scale pile group, J. Geotechnical Eng., Vol. 113 (11): 1326–1343. Brown, A., Morrison, C., & Reese, C. 1988. Lateral load behavior of pile groups in sand, J. Geotechnical Eng., vol.114 (11): 1261–1276. Broms, B. 1964. The lateral Resistance of Piles in Cohesionless Soil, J. Soil Mech. Found. Div., Vol. 90 (SM3): 123–156. Dodagoudar, G., Boominathan, A. & Chandrasekaran, S. 2010. Group interaction effects on laterally loaded piles in clay, J. Geotechnical & Geoenv. Eng., Vol. 136 (4): 573–582. Castelli, F., & Maugeri, M. 2009. Simplified approach for the seismic response of a pile foundations, J. Geotechnical & Geoenv. Eng., Vol. 135 (10): 1440–1451. Frechette, D., Walsh, K. & Houston, W. 2002. Review of design methods and parameters for laterally loaded groups of drilled shafts, Deep foundations 2002:1261–1274. Gaaver, K. 2006. Behavior of laterally loaded piles in cohesionless soils, The Tenth East Asia-Pacific Conference on Structural Engineering and Construction.

© 2011 by Taylor & Francis Group, LLC

Ilyas, T., Leung, F., Chow, K., & Budi, S. 2004. Centrifuge model study of laterally loaded pile groups in clay, J. Geotechnical & Geoenv. Eng., Vol. 130 (3): 274–283. Kumar, S. & Lalvani, L. 2004. Lateral load-deflection response of drilled shafts in sand, International Engineering journal, Vol. 84: 282–286. Liyanapathirana, D. S., & Poulos, H. G. 2005. Pseudostatic approach for seismic analysis of piles in liquefying soil, J. Geotechnical & Geoenv. Eng., Vol. 131 (12): 1480–1487. Matlock, H. & Reese, L. 1961. Foundation Analysis of Offshore Pile Supported Structures, 5 Int. Conf. on Soil Mech. and Found. Eng., Paris (2): 91–97. Matlock, H. 1970. Correlation for design of laterally-loaded piles on soft clay, Proc., 2nd Annual offshore Technology Conf., Vol. 1: 557–594. McVay, M., Zhang, L., Molnit, T. & Lai, P. 1998. Centrifuge testing of large laterally loaded pile groups in sand, J. Geotechnical & Geoenv. Eng., Vol. 124 (10): 1016–1026. Mokwa, L. & Duncan, M. 2005. Discussion of ‘Centrifuge model study of laterally loaded pile groups in clay’ by Ilyas, T., Leung, F., Chow, K., & Budi, S., J. Geotechnical & Geoenv. Eng., Vol. 131 (10): 1305–1308. Ooi, K., Chang, F. & Wang, S. 2004. Simplified lateral load analyses of fixed-head piles and pile groups, J. Geotechnical & Geoenv. Eng., Vol. 130 (11): 1440–1151. Rollins, M., Peterson, T. & Weaver J. 1998. Lateral load behavior of full-scale pile group in clay, J. Geotechnical & Geoenv. Eng., 124 (6):468–478. Rollins, M., Lane, D. & Gerber M. 2005. Measured and computed lateral response of a pile group in sand, J. Geotechnical & Geoenv. Eng., 131 (1): 103–111. Poulos, G. & Davis, H. 1980. Pile foundation analysis and design, John Wiley & sons, Inc., New York, N.Y. Reese, C., Cox, R. & Koop. D. 1974. Analysis of Laterally Loaded Piles in Sand, offshore Technology Conference, Houston: 473–483. Reese, C. & Matlock, H. 1956. Non-dimensional solutions for laterally loaded piles with soil modulus assumed proportional to depth, 8th Texas conference on Soil Mech. and Found. Eng., Austin: 1–41. Reese, C., Wang, T., Arrellaga, A. & Hendrix, J. 1997. LPILE plus 3.0 windows, Ensoft, Inc., Austin, Tex., USA. Reese, C., Wang, T. & Vasquez, L. 2006. Computer program GROUP version 7, tech. manual, Ensoft, Inc.,Austin, USA. Rao, S., Ramakrisha, V. & Rao, M. 1998. Influence of rigidity on laterally loaded pile groups in marine clay, J. Geotechnical & Geoenv. Eng., Vol. 124 (6): 542–549. Yang, K., & Liang, R. 2006. Numerical solution of laterally loaded piles in a two-layer soil profile, J. Geotechnical & Geoenv. Eng., Vol. 132 (11): 1436–1443.

464

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Behavior of piles under combined lateral and axial loading M. Achmus & K. Thieken Institute of Soil Mechanics, Foundation Engineering and Waterpower Engineering (IGBE), Leibniz University of Hannover, Germany

ABSTRACT: In many fields of application piles are used for the transfer of both axial and lateral loads into the subsoil. Combined loading leads to interaction between horizontal and vertical load bearing behavior, which is not usually taken into account in current engineering practice. Therefore numerical investigations were carried out to identify and quantify interaction effects for piles embedded in sand. In this paper the interaction effects are presented in terms of load-displacement curves and in terms of an interaction diagram, which represents the pile system behavior under arbitrary load inclinations. In a small parametric study, different pile diameters and different relative pile stiffnesses are considered to show the dependence of interaction effects on these quantities.

1

INTRODUCTION

Foundation piles are normally and favorably used to transfer axial loads. However, piles with a relatively large diameter can also be and often are exposed to lateral loads acting simultaneously. Due to this combined loading, interaction effects are to be expected, i.e. the horizontal load affects the vertical load bearing behavior and, vice versa, the vertical load affects the horizontal load bearing behavior. In current engineering practice, the interaction effects of combined loaded piles are not taken into account. The deformations in axial and lateral directions are calculated separately regarding only the loads acting in the corresponding direction. Numerous investigations have shown that significant interactions can result from the combined loading of piles in sand. However, it is not yet clear which features of the pile-soil system mainly affect the interaction behavior.

2

from θ = 0◦ (pure tension loading) to θ = 90◦ (pure horizontal loading). Based on their results, Das et al. proposed the following interaction approach (Eq. 1) with regard to the ultimate loads Qu

STATE OF THE ART

Leshukov (1975) carried out tests on 80 cm and 120 cm long piles with a quadratic cross section in silty fine sand. He applied oblique tension forces (combined tension loading) and analyzed the influence of inclination angle on the vertical ultimate load. He found that the ultimate load increases with inclination angles up to 45◦ against the vertical axis and decreases with larger inclination angles. Leshukov proposed taking the effects of combined loading at inclination angles between 10◦ and 40◦ into account. Das et al. (1977) carried out model tests with relatively short, practically rigid piles embedded in loose sand. The direction of the applied load varied © 2011 by Taylor & Francis Group, LLC

465

According to this approach, the ultimate horizontal load is relatively more reduced by a vertical load than inversely the ultimate load by a horizontal load component. Ismael (1988) presented the results of field tests with combined tension loading on small bored piles (length L = 1.5 m, diameter D = 0.101 m) in medium dense sand. He concluded that the horizontal component of the ultimate pile load is only slightly affected by a tensile axial force acting simultaneously. For piles under combined tension loading in sand Sharour & Meimon (1991) carried out numerical simulations and concluded that the horizontal load deformation behavior of a pile is hardly or not at all affected by an axial load, whereas a horizontal load leads to a stiffness reduction in the axial direction. Further investigations were carried out by Patra & Pise (2006). They conducted model tests with single aluminum piles D = 19 mm/L = 722 mm in medium dense sand under combined tension load. They found that the vertical ultimate load may significantly increase due to an additional horizontal component. However, it should be noted that the ultimate load in this study was determined at a very large pile head displacement of up to 15 mm (i.e. 75% of the pile diameter). Meyerhof & Sastry (1985) carried out model tests with piles in loose sand under combined compression

Table 1.

Soil parameters for medium dense sand.

Unit weight γ Oedometric stiffness parameter κ Oedometric stiffness parameter λ Poisson’s ratio ν Internal friction angle ϕ
Dilation angle ψ Cohesion c


Figure 1. Representation of system and loading parameters.

loads. With regard to the experimental results they proposed an interaction equation (Eq. 2).

Meyerhof and Sastry suggested calculating the vertical ultimate load Quv and the horizontal ultimate load Quh in Eq. (2) dependent on the load inclination angle, with a wall friction angle of δ = 0 for purely horizontal load (θ = 90◦ ) and δ = 0.6 ϕ
for purely vertical (compressive) load (θ = 0◦ ). As a result, small vertical loads lead to a larger horizontal ultimate load than purely horizontal loads do. This was also observed in the experiments. Additionally, reference is made to the investigations of Yoshimi (1964), Chari & Meyerhof (1983), Sastry & Meyerhof (1990), Meyerhof (1995), Amde et al. (1997) and Abdel-Rahman & Achmus (2006). Altogether, the existing investigations give no clear view of the effects and their quantitative significance on the interaction under combined pile loading in sand. In most of these investigations, furthermore, only the ultimate loads and not the system stiffnesses are considered.

3

19.0 kN/m3 400 0.60 0.25 35.0◦ 2.5◦ 1.0 kN/m2

The numerical calculation was carried out in three stages. In the first step the initial stress state is generated with vertical stress σz = γ z and horizontal stress σh = γ z k0 for the whole model by using only soil elements. For the coefficient of horizontal earth pressure at rest k0 the usual approach for sand with k0 dependent on the angle of internal friction ϕ
was applied: k0 = 1 − sin ϕ
. Subsequently, the pile installation process was modeled by replacing the soil elements located at the pile position by pile (concrete) elements (“wished in place”) and activating the contact conditions between pile and surrounding soil. The initial state of loading and deformation is defined after the small pile settlement due to own weight occurred. The unit weight of the concrete pile was thereby determined with γ = 25 kN/m3 . In the final stage, the load was applied to the pile head and increased gradually until the ultimate load was reached. A clear failure state was in general only reached under predominantly tensile axial loads. Regarding compressive axial and horizontal loads, the ultimate loads were defined – as usually done in practice – as the loads inducing deflections of 10% of the pile diameter in the corresponding direction.

3.1

Material and contact modeling

For the simulation of the stress-strain behavior of the soil an elasto-plastic material law including the Mohr-Coulomb failure criterion and stress-dependent stiffness was chosen. In order to account for the nonlinear behavior of the soil a dependency of the stiffness modulus for oedometric compression Es on the mean principal stress σm was implemented as illustrated in Eq. (3). Here pref = 100 kN/m2 is a reference stress. The parameter κ determines the soil stiffness for a current stress state with σm = pref and the parameter λ rules the stress dependency of the oedometric stiffness.

NUMERICAL MODEL

In order to clarify the interaction behavior and to quantify the effects for piles embedded in sand, numerical simulations were carried out. For this, a three-dimensional model of the pile-soil system was established using the finite element program system Abaqus (Version 6.8, Abaqus 2006). The system and loading parameters are depicted in Fig. 1. © 2011 by Taylor & Francis Group, LLC

466

For the reference system which is presented in the following, medium dense sand with the parameters given in Table 1 was considered. A small cohesion value was applied in order to enhance numerical stability.

different pile behavior. A negative effect of a horizontal load on the ultimate vertical load, as predicted by the interaction approach Eq. (2), was not confirmed.

4

Figure 2. Contact interaction approach between pile and adjacent soil.

Between pile and soil elements, an elasto-plastic contact behavior was simulated (Fig. 2). The elastic region was defined in a linear increase of the skin friction with relative displacement. This was assumed till reaching a limiting relative displacement uel,slip = 0.005 m, at which the skin friction was set to the maximum skin friction τfric,max resulting from the minimum of the product of horizontal stress σn and coefficient of friction µ and the limiting value of the skin friction τlimit . With regard to limiting values used in design practice (see API 2000), the limiting value was set to τlimit = 80 kN/m2 for medium dense sand. More details concerning configuration and discretization of the model can be found in Achmus et al. (2009). Additionally, it should be noted that geometric nonlinearity was taken into account in order to reflect the influence of relatively large pile deflections due to horizontal loading on the pile behavior.

As a reference system, a reinforced concrete pile (Ep = 30000 MN/m2 , ν = 0.2) with a diameter of D = 1 m and an embedded length of 15 m is considered. The general behavior of a horizontally loaded pile can be assessed by the ratio of the embedded length L to an “elastic length” Le . From the subgrade reaction theory with bedding modulus Ks increasing linearly with depth z (Ks = kr z) the following equation for the elastic length is obtained (e.g. Broms 1964).

With a bedding stiffness of kr = 40 MN/m3 typical for medium dense sand (cf. API 2000) the ratio of the embedded length to the elastic length is L/Le = 7.29 for the reinforced concrete pile considered. Flexible pile behavior with two zero deflection points, i.e. rigid clamping, is normally to be expected when L/Le > 4 to 5. Thus, the reference system reflects a flexible pile. To evaluate all interaction effects of a pile-soil system, numerous inclination angles between α = −90◦ (pure compression loading) over α = 0◦ (pure horizontal loading) up to α = 90◦ (pure tension loading) had to be analyzed. In the course of the study it became clear that the interaction effects depend less on the absolute inclination angle α and more on the normalized inclination angle αnorm (Eq. 5).

3.2 Comparison with experimental results In order to validate the numerical model, backcalculations of the well-reported model tests of Das et al. (1977) for inclined tension loading and of Meyerhof & Sastry (1985) for inclined compression loading were carried out. In these calculations, the soil parameters reported in the aforementioned papers were applied. Due to limited space, the results cannot be presented in detail here. However, the experimental and numerical results agreed well at least qualitatively and in general also quantitatively. Moreover, deficiencies in the experimental investigations were identified by the numerical simulations. Both the interaction approaches regarding ultimate loads proposed by Das et al. (Eq. 1) and by Meyerhof & Sastry (Eq. 2) were found to be unsuitable. In the experiments of Das et al., very large vertical ultimate loads led to a misinterpretation of interaction effects for small horizontal loading portions. The interaction approach according to Meyerhof & Sastry could not be confirmed either although the measured results of the model test agreed very well with the numerical results. However, the numerical analyses showed that the size of the test box used by Meyerhof & Sastry was not sufficient to avoid influences of the boundary on the pile behavior. Assuming a larger test box size in the numerical simulations yielded significantly © 2011 by Taylor & Francis Group, LLC

RESULTS FOR A REFERENCE SYSTEM

Here, for Vult the ultimate compression load (Vult,α=−90 = Vult,c ) or the ultimate tension load (Vult,α=90 = Vult,t ) must be used when combined compression or combined tension loading is considered. For the reference system Hult = 1.54 MN, Vult,c = −4.59 MN and Vult,t = 1.61 MN results. Thus, the absolute value of αnorm is smaller than α for both compression and tension loading. In Fig. 3 the horizontal load-displacement curves obtained for the reference system are presented. With only one exception, namely for the largest load inclination of 61.09◦ for combined tension loading, only minor interaction effects arise. Only the enlarged depiction of the beginning of the load-deflection curves on the right side of Fig. 3 shows differences in the system stiffnesses, which are nevertheless small. For combined compression loading first an increase in the horizontal system stiffness with increasing vertical load is obtained. The reason for this is that the mobilizable passive earth pressure and thus the stiffness in the horizontal direction increases due to vertical

467

Figure 3. Horizontal load-displacement curves (reinforced concrete pile D = 1 m, L = 15 m, medium dense sand).

Figure 4. Vertical load-displacement curves (reinforced concrete pile D = 1 m, L = 15 m, medium dense sand).

downward-directed shear stresses induced by the vertical load transferred via the pile shaft into the ground near the surface. This effect was described by AbdelRahman & Achmus (2006). For higher load levels and thus larger horizontal displacements the importance of this effect decreases. This results from the additional bending moment in the pile due to the moment arm of the vertical load, i.e. from geometrical nonlinearity, which was considered in this investigation. For combined tension loading the horizontal stiffness is decreased, since here the shear stresses acting on the pile shaft are directed upwards and thus reduce the mobilizable passive earth pressure.A distinct effect on the horizontal ultimate load is obtained only for combined tension loading with a large vertical load portion (α = 61.09◦ or αnorm = 60◦ , see Fig. 3). The decrease is induced by large deformations occurring when the ultimate vertical load is approached, which in this case happens due to the large vertical load portion. Besides, the interaction effects on stiffness and ultimate load in the horizontal direction are small, which is in agreement with the general findings of earlier investigations. In Fig. 4 the calculation results for the reference system are shown in terms of vertical load-displacement curves. Here, a larger dependence of the pile-bearing behavior on the load inclination angle than for the horizontal direction is obtained. © 2011 by Taylor & Francis Group, LLC

For combined compression loading a favourable effect of the horizontal load on the pile behavior in the vertical direction is found. The horizontal force increases the horizontal stresses in front of the pile, which leads to larger mobilizable skin friction stresses in the ultimate state. Also with small pile displacements a favourable effect arises, since the upward-directed movement of the passive earth pressure wedge induces a prestressing of the pile (see also Achmus et al. 2009). However, for the reference system this effect is of minor importance. Moreover, for greater horizontal loads the effect is counteracted by the increase in axial deflection due to geometrically non-linear effects. For combined tension loading the opposite effect appears. The upwards-directed movement of the passive earth pressure wedge leads to negative skin friction stresses and thus to a vertical stiffness reduction. Only when large vertical displacements (heave) occur is negative skin friction reduced, and finally even an increase in the ultimate vertical load can arise (α = 61.09◦ , s. Fig. 4). However, with a relatively large horizontal loading part (α = 31.12◦ , see Fig. 4) no increase in the ultimate load occurs, since even with a small pile heave the horizontal ultimate load is reached. For a clear and comprehensive presentation of interaction effects on the pile behavior, the calculation results can be depicted in an interaction diagram. From numerous load-displacement curves calculated with different load inclination angles, the load combinations (H/V) corresponding with certain displacements were derived. From that, lines of equal displacements can be constructed in a H-V interaction diagram. The respective diagram derived for the reference system is shown in Fig. 5. A presentation in a dimensionless form, i.e. the load components are related to the respective ultimate loads without accounting for interaction, was found to be suitable. The displacement values belonging to the different curves shown were taken from the load-deformation curves for purely axial and horizontal loading, respectively. A horizontal course of the curves ux = const or a vertical course of the curves uy = const would mean that no interaction effects occur. Accordingly, the deviation of the curves to the grey-coloured quadratical mesh is a measure of the importance of the H-V interaction effects. Moreover, the load-displacement curves for any load inclination angle can be obtained from an interaction diagram.

5

INFLUENCE OF PILE DIAMETER

The significance of interaction effects is influenced by numerous parameters regarding pile geometry, soil behavior and loading conditions. For practical design it is important to know in what conditions interaction effects become particularly important and should therefore not be neglected. Although a comprehensive parametric study is beyond the scope of this paper, the

468

Figure 6. Comparison of interaction curves for piles with different diameters; left: reference load defined at ux,y = 0.1 D; right: reference load defined at ux,y = 5 cm.

Figure 5. Interaction diagram for the reference system.

Table 2. Investigated systems with varying pile diameter.

Pile

L/D –

Le m

L/Le –

Class –

D = 0.5 m; L = 15 m D = 1.0 m; L = 15 m D = 2.0 m; L = 15 m D = 3.0 m; L = 15 m

30.0 15.0 7.5 3.0

1.18 2.06 3.58 4.95

12.70 7.29 4.19 3.03

Flexible (long) Flexible (long) Nearly rigid Rigid (short)

effect of varying the pile diameter and with that the relative pile stiffness on the magnitude of interaction effects is shown here. Based on the parameters of the reference system, the pile diameter was varied between 0.5 m and 3 m. The parameters of the systems considered are given in Table 2. The piles with D = 0.5 m and D = 1 m can be classified as flexible, whereas the piles with larger diameters are nearly rigid and rigid. Since the quality of interaction is dependent on the load level, two different load levels are considered in the following. On one hand, in order to elucidate interaction effects at a typical service load level, the interaction curves beginning at 50% of the ultimate vertical and horizontal loads are plotted and compared. On the other hand, the interaction curves for the ultimate loads are considered. Two different kinds of interaction diagram are given in Fig. 6. For the diagram in the left part of Fig. 6 the aforementioned definition of failure loads was used, © 2011 by Taylor & Francis Group, LLC

469

i.e. at a pile head deflection of 10% of the pile diameter. Since the pile diameter of the compared systems is different, this definition affects the course of the curves, which makes an assessment of the effect of only the pile diameter on interaction quality difficult. To eliminate this, a second comparison of the interaction effects is shown (Fig. 6 right). In this case the reference load is defined as the load which appertains to a deflection of 5 cm in both vertical and horizontal direction, constant for all considered systems. Obviously, the consideration of different horizontal deflections affects the interaction curves. First of all, it has to be stated that there is no uniform effect of the pile diameter and, with that, the pile stiffness on the quality of interaction, since for different loading type (load inclination and load level) different quantitative effects arise. However, with regard to the horizontal load-bearing behavior it can be stated that the effects are by trend more significant, the larger the diameter or the more rigid the pile behavior becomes. Regarding the vertical load-bearing behavior, only slight differences in the interaction effects with respect to pile diameter are obtained if the reference ultimate load is defined at horizontal and vertical pile deflections of ux,y = 5 cm. Larger differences occur when the reference load definition dependent on the pile diameter is used. This is in particular valid for the service load level in combined tension loading (Fig. 6 left). Regarding different pile diameters, the following effects influence the interaction behavior: • An increase in the pile diameter (at constant pile

length) increases the pile stiffness, i.e. leads to a more rigid behavior. With that, the depth of the area in which significant horizontal pile deflections occur is extended, which leads to larger passive earth pressures and so to a more pronounced interaction.

For rigid pile systems therefore greater interaction effects occur than for flexible systems. This was also found in a comparison of systems with constant pile diameter, but variable pile length (results not shown here). • Large horizontal pile deflection leads in all cases to a decrease of system stiffness. The reduction in vertical stiffness results from the maximum value of skin friction τlimit , which limits the skin friction on the passive side of the pile. Simultaneously the skin friction on the active side is decreasing due to horizontal deflection. With regard to horizontal stiffness, the influenced area increases with increasing horizontal deflection. This leads to greater interaction effects. Moreover, the consideration of geometrical nonlinearity leads to a further reduction of the system stiffness. All in all, a tendency to a greater significance of interaction effects for piles with large diameter-tolength ratio was established.

6

CONCLUSIONS

The numerical investigations showed that combined loading of piles in sand induces complex interaction effects. Depending on load inclination, load level and direction of the vertical load component both unfavourable and favourable effects regarding system stiffness and ultimate load of a pile-soil system can occur. For combined compression loading horizontal as well as vertical system stiffnesses are positively influenced, i.e. increased. By contrast, combined tension loading leads to a negative influence on horizontal as well as vertical system stiffness. However, if the vertical load portion dominates, an increase in the vertical ultimate load can happen due to larger normal stresses acting on the pile shaft induced by the horizontal load. A comparison of the interaction effects for piles with different diameters and otherwise identical system parameters showed that by trend interaction effects are the more important, the larger the pile diameter and thus more rigid the pile system is. However, since many different influences affect the interaction behavior, this statement cannot be generalized with regard to the pile behavior under vertical load. Further parametric studies and a more accurate assessment of the significance of pile geometry, soil type and loading conditions are the subject of ongoing investigations. Moreover, since experimental evidence regarding interaction effects under combined loading is scarce, a systematic experimental test programme would be highly desirable.

© 2011 by Taylor & Francis Group, LLC

ACKNOWLEDGEMENT The presented study was carried out as part of a research project funded by the German Research Council (DFG, project no. AC 100/4-1). The authors are grateful for the financial support. REFERENCES Abaqus. 2006. User’s Manual, Version 6.8. Abdel-Rahman, K., Achmus, M. 2006. Numerical modeling of the combined axial and lateral loading of vertical piles. 6th European Conference on Numerical Methods in Geotechnical Engineering., Graz, Austria, September. Achmus, M.,Abdel-Rahman, K & Thieken, K. 2009. Numerical study of the effect of combined loading on the behavior of piles in sand. International Symposium on Computational Geomechanics, Juan-Les-Pins, France, May Amde, A.M., Chini, S.A., Mafi, M. 1997. Model study of H-piles subjected to combined loading. Geotechnical and Geological Engineering (15): 343–355. API. 2000. American Petroleum Institute. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design. API Recommended Practice 2A-WSD (RP2A-WSD), 21st edition, Dallas. Broms, B. 1964. Lateral Resistance of Piles in Cohesionless Soils. ASCE Journal of the Soil Mechanics and Foundation Division 90 (3): 123–156 Chari, T.R., Meyerhof, G.G. 1983. Ultimate capacity of single rigid piles under inclined loads in sand. Canadian Geotechnical Journal (20): 849–854 Das, B.M., Seeley, G.R., Raghu, D. 1977. Uplift Capacity of Model Piles under Oblique Loads. ASCE Journal of the Geotechnical Engineering Division 102(9): 1009–1013 Ismael, N.F. 1989. Field Tests on Bored Piles Subject to Axial and Oblique Pull. Journal of Geotechnical Engineering, 115 (11): 1588–1598 Leshukov, M.R. 1975. Effect of Oblique Extracting Forces on Single Piles. Togliatti Polytechnic Institute. Translated from Osnovaniya, Fundamenty i Mekhanika Gruntov, No. 5, p. 15, Sept.–Oct. 1975 Meyerhof, G.G. 1995. Behaviour of Pile Foundations under Special Loading Conditions. 1994 R.M. Hardy Keynote Address. Canadian Geotechnical Journal (32): 204–222 Meyerhof, G.G., Sastry, V.V.R.N. 1985. Bearing capacity of rigid piles under eccentric and inclined loads. Canadian Geotechnical Journal (22): 267–276 Patra, N.R., Pise, P.J. 2006. Model pile groups under oblique pullout loads – an investigation. Geotechnical and Geological Engineering (24): 265–282 Sastry, V.V.R.N., Meyerhof, G.G. 1990. Behaviour of flexible piles under inclined loads. Canadian Geotechnical Journal 27(1): 19–28. Shahrour, I., Meimon, Y. 1991. Analysis of behaviour of offshore piles under inclined loads. International Conference on Deep Foundations, pp. 227–284. Yoshimi, Y. 1964. Piles in cohesionless soil subject to oblique pull. ASCE Journal of the Soil Mechanics and Foundation Division, 90 (6): 11–24.

470

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Investigations on the behavior of large diameter piles under cyclic lateral loading M. Achmus, J. Albiker & K. Abdel-Rahman Institute of Soil Mechanics, Foundation Engineering and Waterpower Engineering, Leibniz University of Hannover, Germany

ABSTRACT: Large diameter monopiles are an established foundation type for offshore wind energy converters (OWEC), although currently practical experience in the design and the behavior of these constructions is rare. Achmus et al. (2008) developed a method for estimating the permanent pile deformation under one-way loading which is to be expected over the lifetime of an OWEC. The procedure, called the stiffness degradation method (SDM), is based on the combination of cyclic drained triaxial test results with numerical simulations. Although it yields plausible results, further validation of the method by comparison with experimental results is very important to ensure its applicability. Recently, LeBlanc et al. (2010) presented the results of a series of 1g-model tests of a monopile in sand under cyclic horizontal loading. These test conditions were modeled with the SDM under drained conditions in order to verify the procedure. The comparison of numerical and test results proves the applicability of the stiffness degradation method. 1

INTRODUCTION

The planned offshore wind farms in the German parts of the North Sea and the Baltic Sea will be constructed in water depths varying from approximately 15 to 40 m. By means of suitable foundation constructions, the large horizontal forces and bending moments resulting from wind and wave loads must be economically and safely transferred to the sea soil. Monopile foundations can be used as one of these foundation types. This foundation method was already implemented for OWECs in the North and the Baltic Sea, but only in water depths of less than about 15 m. Its application is expected to be extendable for water depths up to about 25 to 30 m. However, the diameters of such monopiles will then vary between 5.0 and 7.5 m (Fig. 1). Since wind energy converters are relatively sensitive to deformations, in particular tilting, it is very important to estimate these as exactly as possible. For the mentioned large-diameter piles, there is to date no approved procedure for this. In this paper, the special numerical concept SDM is described, and the results are compared with experimental ones derived from 1g-model tests performed for monopiles under cyclic loading. This method was used in this paper to predict the behavior of monopiles under drained condations.

2

SIMULATION OF THE MONOPILE BEHAVIOR UNDER STATIC LOADING

A three-dimensional (3D) finite element model was established in order to analyze the behavior of © 2011 by Taylor & Francis Group, LLC

Figure 1. System and denominations for a monopile foundation.

monopiles embedded in sand soil. The computations were carried out using the finite element program system ABAQUS (ABAQUS 2008). The most important issue in geotechnical numerical modeling is the simulation of the soil stress-strainbehavior. An elasto-plastic material law with MohrCoulomb failure criterion was used. The soil stiffness is here represented by a stiffness modulus for oedometric compression Es and a Poisson’s ratio ν. To account for the non-linear soil behavior, a stress dependency

471

Table 1.

Material parameters used for dense sand.

Unit buoyant weight γ
Oedometric stiffness parameter κ Oedometric stiffness parameter λ Poisson’s ratio ν Internal friction angle ϕ
Dilation angle ψ Cohesion c


11.0 kN/m3 800 0.55 0.25 37.5 7.50◦ 0.1 kN/m2

of the stiffness modulus was implemented as follows, according to Ohde (1939): Figure 2. Deflection lines of a monopile (D = 2.0 m) calculated using Lpile (2000) and FEM.

Herein σat = 100 kN/m2 is a reference (atmospheric) stress and σm is the current mean principal stress in the considered soil element. The parameter κ determines the soil stiffness at the reference stress state and the parameter λ rules the stress dependency of the soil stiffness. The material parameters used here are typical for dense sand and are given inTable 1. For more details about the numerical modeling see Abdel-Rahman & Achmus (2005). The stress-dependency of the stiffness modulus given by Equation 1 is widely used in soil mechanics. However, no direct experience exists on the magnitude of the two parameters (κ, λ) to be used in the calculation of horizontally loaded piles. In order to calibrate these parameters in connection with the numerical model, firstly monopiles of smaller diameters were investigated. For such diameters the p-y-method is known to give a reasonable estimation of pile deflection, so the numerical results could be compared with the results of the p-y-method for calibration.The calculations with the p-y-method were carried out by means of the Lpile program (Lpile 2000). The calculated deformations of a monopile with a diameter D = 2.0 m with different embedded lengths varying from 20 to 40 m with a wall thickness of 3.0 cm under monotonic loading are shown in Figure 2 and compared with the p-y-method results. The loading consisted of a horizontal force acting at a height h above the soil surface. The magnitude of the load was varied between 0.5 and 4.5 MN. The results are valid for κ = 800 and λ = 0.55. This parameter combination was found to give the best matching results with respect to the p-y-method. This was also verified with similar comparisons for a pile of a diameter D = 1 m and embedded lengths of 20 to 40 m. Further investigations have been made by the authors; see Achmus et al. (2008). 3

SIMULATION OF THE MONOPILE BEHAVIOR UNDER CYCLIC LOADING

Figure 3. Degradation of secant modulus under cyclic loading in a drained triaxial test.

pile-soil system by numerical calculations, taking the behavior of soils under cyclic loading investigated in cyclic triaxial tests into account (Achmus et al. 2009, Kuo 2008). This method is based on the finite element model presented above and accounts for cyclic loading by a special stiffness degradation approach. A sketch of principles of the results of a stresscontrolled cyclic triaxial test under drained conditions is shown in Figure 3. The plastic portion of the axial strain εap increases with the number of load cycles. The increase rate of the plastic strain is mainly dependent on the initial stress state (confining stress) and on the magnitude of the cyclic load portion. The strain increase can be interpreted as a decrease in the secant stiffness modulus. When the elastic strain is negligible, the degradation of the secant modulus EsN can be formulated in the following way dependent on the plastic strain in the first cycle εap,N =1 and in the Nth cycle εap,N :

To investigate the lateral deformation response of a monopile under cyclic loading, a method was developed which yields the permanent displacements of a © 2011 by Taylor & Francis Group, LLC

472

The degradation of secant stiffness in a cyclic triaxial test with isotropic initial stress condition can be determined from the plastic strains measured with a regression equation. Such equations were presented, for instance, by Huurman (1996), Gotschol (2002) and Werkmeister (2004). Due to the approach of Huurman used here, the increase in deformation or the decrease in stiffness, respectively, can be described by the following equation:

Here N is the number of cycles, X is a stress-dependent variable (cyclic stress ratio), and b1 , b2 are regression parameters to be determined in triaxial tests. The cyclic stress ratio is defined as

wherein σ1,f is the main principal stress at failure in a monotonic test. Thus, the stress ratio is dependent on the initial stress state and on the cyclic load level. A problem to be dealt with is that the Equations (3) and (4) are valid for triaxial test conditions with isotropic initial stress conditions and a constant confining pressure σ3 during cyclic loading. In the pile-soil system, the initial stress conditions (before application of the horizontal load) are anisotropic and the minor principal stress in the elements as well as the direction of the principal stress axes in general change with the application of the load. To overcome this problem, a characteristic cyclic stress ratio Xc is defined here as

Here the index (1) indicates the cyclic stress ratio at loading phase and the index (0) at unloading phase. At the initial (and unloading) phase, only the vertical load V due to the tower weight is considered, and the lateral load H is applied subsequently in the loading phase. The characteristic cyclic stress ratio is derived from the difference between the stress ratios in the loading and the unloading phase. The accumulation of plastic strain and the degradation of stiffness of the soil element can be obtained from Equation (3) by replacing X by Xc . Figure 4 shows the lateral deflection of monopiles with different embedded lengths (L = 20 m and 40 m) and with a wall thickness tp of 9 cm calculated using the SDM. The long pile shows a better cyclic performance than the shorter one, which could be explained by the different loading levels (H /Hu ). The ultimate horizontal loading (Hu ) was determined here using the hyperbolic method by Manoliu et al. (1984). Figure 5 shows that the accumulation rate for the pile with longer embedded length and, with that, a lower cyclic loading level is smaller than for the shorter © 2011 by Taylor & Francis Group, LLC

473

Figure 4. Lateral pile deflection obtained from stiffness degradation model.

Figure 5. Accumulated displacement of a pile under cyclic loading (D = 7.5 m, h = 20 m, H = 15 MN).

pile with a higher cyclic loading ratio. This means that the accumulation rate is a function of the cyclic loading level (H /Hu ). This seems obvious, but with existing approaches an effect of the loading level is not reflected. The new stiffness degradation method is capable of taking soil, geometry and loading conditions into account. For detailed parametric study, refer to Achmus et al. (2008) and Achmus et al. (2009). 4 VALIDATION OF THE STIFFNESS DEGRADATION METHOD For validation of the described method, comprehensive comparison with experimental results is needed. Recently, LeBlanc et al. (2010) (see also LeBlanc 2009) presented the results of a series of 1g-model tests of a monopile in sand under cyclic horizontal loading and developed an approach for estimating the accumulation of plastic deformation with the number of load cycles. The test conditions were modeled with the SDM in order to compare the numerical results with the test results. 4.1 Test program carried out by LeBlanc et al. (2010) In the laboratory model a stiff copper monopile was installed in a basin filled with unsaturated sand. The pile was loaded by a sinusoidal cyclic load, which acted at the top of the pile and was generated by a specially constructed loading rig. The horizontal force creates a moment M = H × h acting at the ground surface,

Table 2. Properties of the small scale copper pile used in the laboratory by LeBlanc et al. (2010). Pile diameter, D (m) Wall thickness, tp (m) Penetration depth, L (m) Load eccentricity, h (m) Pile weight, V (kN)

0.08 0.02 0.36 0.43 0.035

Table 3. Dimensionless parameters introduced by LeBlanc et al. (2010). Moment loading Vertical force Pile rotation (degree) Load eccentricity Slenderness ratio

˜ = M /(L3 Dγ
) M V˜ = V/(L2 Dγ
) θ˜ = θ pa /(Lγ
) e˜ = M /(HL) = h/L η = L/D

where h is the load eccentricity. The secant rotation of the pile was measured by applying two deflectometers at different positions. The dimensions of the eccentrically loaded pile are shown in Table 2. The values of the parameters of the laboratory pile had to be determined accurately in order to be able to compare the test results to the deformations of a typical offshore monopile. For structures in sand the load response is governed by the frictional behavior, which in turn is dependent on the isotropic stress level. The much lower isotropic stress level in the laboratory, compared to a full-scale test, influences the values for the friction angle and the relative density. These issues of scaling are addressed by applying adequate scaling methods, which leads to the development of a complete non-dimensional framework. For a detailed description see LeBlanc et al. (2010). As a result the dimensionless value for the stiffness k˜ is expressed as a function of three further non-dimensional parameters, i.e. the vertical force V˜ , the load eccentricity e˜ , and the pile slenderness ratio η. The non-dimensional moment-rotation relationship in turn is expressed by the following equation:

Thus, a similar behavior of model and prototype is to be expected with regard to both stiffness and strength ˜ against when plotting the non-dimensional moment M the non-dimensional rotation θ˜ , while retaining the three parameters controlling the stiffness constant. Table 3 gives an overview of the definitions of the introduced dimensionless parameters. The parameter pa labels the atmospheric pressure of 100 kN/m2 . The test program carried out by LeBlanc et al. (2010) was developed to investigate the response of the pile and its dependency on the relative density of the sand and the characteristics of the applied loading. In order to choose appropriate values for the relative density in the model, a relation of Schnaid (1990) was used, which relates relative density, friction angle and © 2011 by Taylor & Francis Group, LLC

effective vertical stress. Two test series were conducted at relative densities of Dr = 4% and Dr = 38%, respectively. The peak friction angles used in the laboratory were estimated 35◦ and 43◦ . Under field conditions, these parameter values correspond to relative densities of Dr = 8% and Dr = 75%, respectively, due to the much higher effective vertical stress. The loading amplitude is expressed by the maximum moment of the applied loading in a cycle Mmax , normalized by the static moment capacity MR . Furthermore, a loading cycle is characterized by the values of the minimum and the maximum applied moment, Mmin and Mmax . Hence, for describing the specific loading conditions LeBlanc et al. (2010) introduced the parameters ζb and ζc :

ζb was chosen in a range of between 0.3 and 0.5 to reflect realistic loading conditions. ζc lies in a range of between −1.0 and +1.0, where values of ζc < 0 characterize two way loading. 4.2

Simulation with the SDM

For determining the static moment capacity of the model pile, LeBlanc et al. (2010) initially carried out static load tests in both loose and medium dense sand, ˜ against θ˜ . Here, in order to validate and plotted M the developed numerical model, these static tests were simulated in the original small scale dimensions, and the results were compared to the laboratory test results. Both relative densities of Dr = 4% and Dr = 38% were considered. The values for κ and λ were chosen as 400 and 0.6, respectively, for the loose state (Dr = 4%) and as 800 and 0.5 for the medium dense state (Dr = 38%). The values for the unit weights of the unsaturated sand were γ = 14.7 and 15.8 kN/m3 , respectively. The friction angles in the simulations were varied with ϕ = 32◦ and 35◦ for loose sand and ϕ = 37.5◦ , 40◦ and 43◦ for medium dense sand. The comparison of numerically obtained results and laboratory results is shown in Figure 6. LeBlanc et al. (2010) estimated the friction angles valid for small stresses to 35◦ and 43◦ , respectively. Applying these values in the simulation, the ultimate load is overestimated. Using smaller, but still reasonable values, a fairly good agreement is obtained. However, the agreement with regard to loads less than about 50% of the ultimate load is in all cases very good, which means that at least for service loads the numerical model yields very good results. This is the load range considered here. The validation of the developed numerical model with regard to static loading was therefore successful. For modeling with the SDM, here only the test series with the pile embedded in medium dense sand

474

Figure 6. Dimensionless moment-rotation curves obtained by numerical simulation and by LeBlanc et al. (2010). Table 4. Tests chosen for back-calculation with the SDM. e˜

Dr

ζb

ζc

N

1.19 1.19 1.19

38% 38% 38%

0.27 0.40 0.52

0 0 0

8090 7423 17532

Table 5. Pile properties of a typical full-scale offshore pile, scaled by applying the scaling law of LeBlanc et al. (2010). Pile diameter, D (m) Wall thickness, tp (m) Penetration depth, L (m) Load eccentricity, h (m) Pile weight, V (kN)

4.44 0.13 20 23.8 6284.2 Figure 7. Accumulated pile rotation θ(N )/θs against number of load cycles N, results of the laboratory tests and of the numerical simulations with SDM.

with Dr = 38% was regarded. Furthermore only oneway loading with ceasing of the applied moment after each load cycle can be simulated using the SDM, i.e. only simulations where ζc equals 0. Three different tests were chosen for back-calculation with the SDM, which are characterized by the parameters given in the following Table 4. The tests were simulated in the original small scale parameter configurations with the geometrical dimensions given in Table 2 and, furthermore, under application of the mentioned scaling law, in dimensions and parameter configurations that characterize a typical full-scale offshore pile. Table 5 shows the associated pile properties. LeBlanc et al. (2010) presented the results of the laboratory tests in terms of the evolution of the accumulated rotation of the pile by plotting the rotation resulting from cyclic loading θN , diminished by the rotation after the first cycle θ0 , in terms of the rotation θS that would occur in a static test when the applied load equals the maximum cyclic load:

© 2011 by Taylor & Francis Group, LLC

In Figure 7 a) to c) the laboratory results are compared to the results obtained numerically by applying the SDM. For the calculations presented here, the cyclic parameters b1 and b2 were chosen with regard to experience (Kuo 2008), since no suitable cyclic triaxial test results were reported. Values of 0.16 and 0.38, respectively, were used when calculating the system in original dimensions. These values are typical for a medium dense sand. For the system in full scale dimensions values of 0.2 and 5.76 typical for dense sand were used. LeBlanc et al. (2010) found that the pile behavior can be predicted by the following equation:

where Tb and Tc are dimensionless functions, depending on the load characteristics and the relative density. Based on the previous equation it follows that the accumulated rotation increases exponentially with N . For the test results given in Figure 7 (a, b & c) the exponent α was found to be 0.31. Also from the numerical

475

modeling the accumulated rotation increases exponentially, though this finding is not absolutely appropriate for low values of N , i.e. here for N = 10. Disregarding the results for this number of load cycles, the exponent describing the increase of the accumulated rotation in the simulations was found to vary between 0.22 and 0.24, slightly ascending with the amplitude of the applied cyclic load. However, it should be noted that the simulation results were obtained with estimated cyclic parameters b1 and b2 , which of course strongly affect the cyclic behavior. For future applications, it could be beneficial to search for a physical way to determine the values of these parameters, in order to make the SDM reflect more realistic results. However, from the presented calculations it is obvious that the numerical results match the model results of LeBlanc et al. (2010) quite well. The laboratory small-scale results show that the accumulated rotation is higher under a higher value of loading level (ζb ). From the numerical modeling using SDM, similar results were also obtained, whereby the dependency between cyclic loading level and accumulated rotation is not as strong as found in the laboratory results. 5

CONCLUSIONS

The stiffness degradation model as a numerical concept enables the estimation of permanent pile displacements and rotations under cyclic lateral loading. The comparison with the model presented by LeBlanc et al. (2010) model results shows good agreement both qualitatively and quantitatively. Despite of SDMsimplifications, this method is suitable to predict the behavior of monopiles under drained conditions in a realistic manner. It can be used to assess the influence of pile geometry and soil type on the performance of piles under cyclic loading. ACKNOWLEDGEMENTS

REFERENCES ABAQUS User’s Manual, Version 6.7. 2008. Simulia, Providence, RI, USA. Abdel-Rahman, K. &Achmus, M. 2005. Finite Element Modelling of Horizontally Loaded Monopile Foundations for Offshore Wind Energy Converters in Germany, International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, Australia. Achmus, M., Abdel-Rahman, K. and Kuo, Y.-S. 2008. Design of Monopile Foundations for Offshore Wind Energy Plants, 11th Baltic Geotechnical Conference – Geotechnics in Maritime Engineering, Gdansk, Poland, Vol.1, pp. 463–470 Achmus, M., Kuo,Y.-S. and Abdel-Rahman, K. 2009. Behavior of monopile foundations under cyclic lateral load, Computers & Geotechnics 36 (2009), pp. 725–735 Gotschol, A. 2002. Veränderlich elastisches und plastisches Verhalten nichtbindiger Böden und Schotter unter zyklisch-dynamischer Beanspruchung, Ph.D. thesis, Universität Kassel, Kassel, Heft 12. Huurman, M. 1996. Development of traffic induced permanent strains in concrete block pavements, Heron, Vol. 41, No. 1. pp. 29–52. Kuo, Y.-S. 2008. On the behavior of large-diameter piles under cyclic lateral load, Ph.D. thesis, Leibniz Universität Hannover, Hannover, Heft 65. LeBlanc, C. 2009. Design of Offshore Wind Turbine Support Structures, Ph.D. thesis, Aalborg University, Denmark, DCE Thesis No. 18. LeBlanc, C., Houlsby, G. T. & Byrne, B. W. 2010. Response of stiff piles in sand to long-term cyclic lateral loading, Geotechnique, Vol. 60, Issue 2, pp. 79–90. Lpile, 2000. User’s manual, Version Lpile plus 4.0. Manoliu, I., Dimitriu, D. V. & Dobrescu, GH. 1985. Loaddeformation characteristics of drilled piers, Proc. 11th International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 3, pp. 1553–1558. Ohde, J. 1939: Zur Theorie der Druckverteilung im Baugrund. Der Bauingenieur 20, pp. 451–459 (in German). Schnaid, F. 1990. A study of the cone-pressuremeter test in sand, Ph.D. thesis, University of Oxford. Werkmeister, S. 2004. Permanent Deformation Behaviour of Unbound Granular Materials in Pavement Constructions, Ph.D. thesis, Techn. Universität Dresden, Dresden, Heft 12.

The results presented in this paper were obtained as part of the GIGAWIND ALPHA VENTUS research group project funded by the Federal Ministry for the Environment, Natural Conservation and Nuclear Safety, Germany. The support is thankfully acknowledged.

© 2011 by Taylor & Francis Group, LLC

476

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

BP Clair phase 1 – Pile driveability and capacity in extremely hard till T.R. Aldridge & T.M. Carrington Fugro GeoConsulting Limited

R.J. Jardine Imperial College, London

R. Little Fugro GeoConsulting, Inc. (Formerly of Fugro-McClelland)

T.G. Evans EPT, BP Exploration

I. Finnie Advanced Geomechanics (Formerly of Lloyds Register)

ABSTRACT: Evans et al. (2010) describe how the foundation engineering for BP’s Clair Phase 1 Drilling and Production Platform, West of Shetland, UK, had to consider tills with unprecedentedly high undrained shear strengths and unit weights. Boulders were also present. This paper describes the technical approach taken by BP’s foundation assurance team in addressing these challenging conditions, focusing principally on driveability and axial capacity. Advanced field and laboratory investigations were conducted to allow a range of analyses that explicitly considered the effects of cyclic loading, group action, strain softening and possible pilot hole drilling. Instrumented advance driving trials were conducted, while the main jacket installation was also instrumented and back-up drilling options mobilised in case of harder-than-expected driving.

1

INTRODUCTION

A companion paper (Evans et al., 2010) describes the development planning for the Clair field and the strategy adopted by BP and their co-venturers in assuring foundation design under unprecedented geotechnical conditions. The four legged Clair Phase 1 jacket was installed with 3 or 4 2590 mm o.d. × 85–95 mm w.t. tubular sleeve piles driven per corner. This paper describes the technical approach taken in assessing foundation design from basic principles. The work was led by an international assurance team (IAT) assembled by BP to provide industrial and academic experience. The lessons learnt are now being applied in the new Clair Ridge development.

Bulk densities were up to 2.4 Mg/m3 despite ‘standard’ particle specific gravities (2.65–2.75), indicating very low void ratios. Geophysical data and well drilling records both indicated boulders at shallow depths. Boulders or cobbles occurred every 6 m on average in the geotechnical boreholes, but were more frequent close to the geological soil boundaries at 12, 22 and 33 metres below seafloor. The sandy clay tills above the 12 m boundary were the toughest, with some CPT qc pushes up to 120 MPa and some UU su tests
maxima >50. High exceeding 2500 kPa, giving su /σvo
su /σvo ratios dominated in the tills, but lower strengths
< 0.5) were apparent either side of the (with su /σvo 22 m boundary. These softened, possibly weathered, units were also associated with a very dense sand layer encountered between 24 and 26 m.

2 THE CLAIR 1 GEOTECHNICAL PROFILE 3 ADVANCED GEOTECHNICAL TESTING Standard drilling and sampling techniques gave poor results at Clair. A second investigation using the Geobore “S” rotary coring system provided far better samples, and high capacity CPT equipment provided a more accurate geotechnical profile. Fig. 1 summarises the sequence of very hard and dense low plasticity sandy and silty boulder clays, which originate from the Stormy Bank, Otter Bank and Ferder formations. © 2011 by Taylor & Francis Group, LLC

An advanced geotechnical test programme was performed to support ‘first principles’ foundation engineering. Data was required for effective stress design methods and the analysis of cyclic response. For the ‘MTD’ approach (Jardine and Chow, 1996) oedometer tests on remoulded and reconstituted soil were conducted, along with soil-steel ring shear interface

477

Figure 1. Soil parameters at Clair.

tests and clay sensitivity measurements. Design YSR
data, noting profiles were developed from the su /σvo that the K0 values required for stress path testing might be lower than expected by assuming a stress history of monotonic K0 overconsolidation. Cyclic direct simple shear tests, anisotropic triaxial tests with bender elements, triaxial permeability tests and suction probe tests were also performed. The possibility of cementation in the high strength soils was checked using mineralogy, carbonate content and SEM investigations. Bio-stratigraphic analyses including palynological and micro- and nano-fossil examinations were also used to improve geological understanding of the site. No evidence of cementation was found, and it was concluded that the high densities and strengths might have resulted from compaction under the shearing action of advancing and retreating ice sheets. 4 AXIAL PILE CAPACITY 4.1

Base case design approach by MEI

Mustang Engineering Inc. (MEI), the designer, noted API’s reservations on the use of their method for high su /σv
values, due to the lack of pile tests in soils with su /σv
values greater than three. Quirós et al. (2000) demonstrated that very high past consolida
)nc values substantially tion stresses can lead to (su /σv0 below the 0.25 value implicitly assumed in API. MEI therefore adopted the Randolph and Murphy (1985) approach, adding a constraint that the unit friction be limited to Kp σv
tan δ. This provided their ‘Baseline’ static capacity for a single pile. MEI assessed the effects of strain softening, cyclic loading, group action and strain rate, estimating the net effect of these considerations to be neutral, allowing their © 2011 by Taylor & Francis Group, LLC

Figure 2. Skew of pile test trends with YSR and L/D; showing Jardine and Chow (1996) data base points and API trends.

‘Operational’ jacket pile capacities to be equated to the ‘Baseline’ static single pile capacities. 4.2

Re-assessment of main text API by the IAT

Since standard methods of assessing axial group interaction and cyclic loading effects might not be appropriate for Clair, and the potential effects of pilot hole drilling also needed considering (Sullivan and Ehlers, 1973), the IAT decided to investigate these issues, along with potentially positive factors such as loading rate effects and skews in the pile API database with YSR and pile L/D ratio.
A multiple parameter analysis involving su /σvo , L/D and plasticity index showed that trends like those in
and L/D Fig. 2, when extrapolated to the Clair su /σvo ratios, would result in estimated static capacities up to
, and L/D ratios kept 38% higher than API. With su /σvo within the API pile load test data-base, static estimated capacities would be around 16% higher than calculated by the standard API approach.

478

4.3

MTD-ICP calculations by the IAT

Calibration checks indicated the MTD-ICP approach (Lehane et al., 1994, Jardine and Chow 1996, Jardine et al. 2005) should cope with the extreme YSRs and low L/Ds at Clair, and be extendable to incorporate cyclic and group effects. Specialist laboratory testing provided the data required to run the MTD-ICP method, and also confirmed the tills to be completely insensitive, with high

interface friction angles, δ
, of just below 30◦ . However, cyclic simple shear tests showed the tills to be strongly affected by high level cycling. Initially, an upper limit of 8 was imposed on parameter Kc to keep within the data-base of field-tests on instrumented piles. However, Chow’s (1997) database of industrial piles identified cases where the ICP procedure was consistent with higher Kc values, implying that the Kc restriction could be removed. Capacities 10% higher than API were calculated using a Kc limit of 8, with values 44% higher being calculated without this limit being imposed. 4.4

Cyclic effects

The jacket design indicated that the foundations would have significant cyclic loading. MEI’s non-linear inplace jacket analyses provided data to relate cyclic pile head loads to wave heights. Typical storm wave height distributions provided cyclic load build up and decay during the design storm. A programme of cyclic and post-cyclic static simple shear tests was designed to input into the Jardine (1994) and Jardine et al. (2005) approaches for predicting how the normal (radial) effective stresses σn
(and hence shear capacity) on
the shaft reduce from initial equilibrium values σnc under cyclic loading. Functions were fitted to the sim
to N, the number ple shear data that related σn
/σnc of cycles applied, and the cyclic shear loading level, expressed as τ/τmax , where τmax is the static shear capacity and τ is the cyclic load amplitude. Conducting the cyclic tests at appropriate normal stresses and keeping volume constant matched the undrained conditions expected close to the pile shaft. The results defined ratios σr
/σrc
expected after successive wave packages, with a cumulative ‘equivalent number of cycles’ model quantifying the overall effect. The analysis indicated axial capacity could reduce 12 to 14% during the design storm, although shaft friction recovery was considered probable with time, even after severe cyclic loading. 4.5 Strain-softening effects Local brittleness in shaft friction is implicit in the MTD-ICP approach. Peak and ultimate local δ
values from laboratory ring-shear tests are used to estimate the effect on axial capacity in ‘falling-branch’ t-z axial load-deflection analyses. The Clair ring-shear tests showed only a 4% difference between peak and ultimate δ
values (from 29.5◦ to 28.5◦ ). These were incorporated into the MTD-ICP capacity calculations. Strain softening effects were also assessed by performing API t-z analyses in which the postpeak t-z reduction was based on triaxial shear strength measurements taken to large strain. These indicated post-peak reductions of 7% in the clays to 15 m below seafloor and 2.5% in the underlying clays. The falling branch t-z analyses gave a peak shaft capacity 4.4% lower than that determined by assuming that all points on the shaft reached their local peaks simultaneously. © 2011 by Taylor & Francis Group, LLC

Figure 3. Axial capacity assessments.

The reduction in overall capacity (i.e. including base resistance) was 2.8%. 4.6

Rate effects

Studies by Bea and Audibert (1979) and Tang (1988) imply that the axial capacities of piles driven into clays may be 1.53 to 1.56 times higher under wave loading at 0.1 Hz than their static capacities. Laboratory loading rate tests were not conducted on Clair samples, but comparable positive effects are thought likely to apply. However, the IAT had to ensure the piles could withstand repeated loading cycles during storms without developing a slow creeping failure, which such positive rate effects might not prevent. 4.7

Group effects

The design spacing of the pile groups left less than one pile diameter between some piles. Equivalent pier analyses indicated that pile groups should not fail as a unit under axial loading. However, recent field tests have shown that shear stress fields emanating from closely-centred piles interact negatively, reducing individual pile capacity (Lehane et al. 2003). The Converse-Labarre equation, which gave a good match with the above field tests, indicated an axial group efficiency of approximately 87% for the Clair groups. Simplified theoretical calculations based on overlapping concentric shear stress fields gave a similar 12% reduction in group capacity. 4.8

Summary for pile capacity

Fig. 3 compares the pile capacities obtained by: 1. The MEI ‘Operational’ capacity; 2. The IAT-modified API method including strain softening, database bias, cyclic and group effects;

479

3. The IAT MTD-ICP calculations, including postpeak softening effects; 4. The IAT MTD-ICP capacity including cyclic and group effects, at the 29 m target penetration. At 29 m, approaches 1, 2 and 4 led to operational capacities falling within about 10% of each other, giving confidence in the approaches adopted and also in the direct adoption of the MEI operational capacity.

5 5.1

Well drilling and geophysics data indicated that the probability of a 2438 mm o.d. × 76 mm w.t. pile refusing on a large boulder was ∼10%, with refusals most likely in the upper 25 metres. The probability of one pile per group refusing on a boulder was ∼34%, whilst providing one spare slot per group offered a 92% overall probability of a successful installation. However, it was concluded that having drilling and grouting systems offshore would be a more cost-effective remedy for premature refusals. 5.5

Main considerations

Pile collapse

Consideration was given to the possibility that hard inclusions might cause piles to buckle or collapse inwards, possibly without showing initial refusal. The limiting obstruction size was estimated from the dynamic resistance generated when the pile struck the inclusion, following the approach developed to understand collapses seen in the Goodwyn A and Valhall IP installations (Aldridge et al., 2005). The analysis indicated that the pile tip sections should be at least 76 mm thick 440 MPa yield strength steel.

Soil resistance to driving (SRD)

The upper-bound SRD estimates used a conservative interpretation of the soils data, and because of the unusually high CPT values, upper bound SRD estimates included the cone-based ICP method (Lehane et al., 2000), which exceeded the best estimate by an order of magnitude, indicating possible refusal 10 metres below seafloor – not deep enough to provide adequate foundations. Options considered to reduce the uncertainty in SRD and improve confidence in meeting target penetrations included an internal driving shoe, pre-drilled undersized pilot holes, strengthened pile tips and pile driving trials. Steps to help manage residual installation uncertainties also included spare pile slots for additional piles and a backup drilling/grouting spread to allow for intervention in the event of premature refusals. Noting that shaft friction would be reduced by an unpredictable amount by any internal driving shoes, this option was not taken forward.

5.3

Boulders

PILE INSTALLATION

Factors addressed for pile driveability included whether “standard” approaches for hard clays, such as Toolan and Fox (1977) or Stevens et al. (1982) could apply at Clair, the feasibility of driving through cobble layers, the probability of impacting on large boulders, and the potential for pile collapse when driving through such hard soils.

5.2

5.4

Predrilled pilot hole

The option of a 1524 mm pilot hole pre-drilled through the upper very hard clay layers, to approximately 12 m below seafloor, was considered, to reduce friction during driving. The probable effect on long term pile capacity was estimated by referring to model tests on soft clays by Rojas (1993). For the Clair pile geometry and a pre-drilled 1524 mm pilot hole, Rojas’ tests suggested a most probable long term capacity reduction of 17%, with possible loss of 30%. Carefully designed tests at a heavily over-consolidated site would have been required to make a better assessment. It was concluded that pilot hole drilling should be avoided if possible. © 2011 by Taylor & Francis Group, LLC

6

PILE DRIVING TRIALS

A drilling template was to be installed one year before the jacket, with two 1829 mm o.d. × 75 mm w.t. jacket docking piles being driven at the same time. Instrumenting these piles and driving them beyond the required docking pile penetration to the jacket pile target 29 m penetration provided an opportunity to assess pile driving and set-up. One docking pile included an oversized internal driving shoe (80 mm). The MHU 3000 was used for the docking piles to provide the best information for the jacket sleeve pile driving, expected to use the same hammer. A comparison of observed blowcounts with those predicted using the Stevens et al. (1982) approach for both plugged and unplugged driving is presented for shoed and shoeless docking piles in Fig. 4. Docking pile driving was easier than predicted, blowcounts for MHU 3000 hammer efficiencies of 50 and 70% falling near the prediction for a hammer efficiency of 90%. The trend of actual blowcounts for the shoed DP-1 pile followed the predicted lower bound coring case closely to about 24 m penetration, then increased towards the lower bound plugged case prediction. The blowcounts for the shoeless DP-2 pile fell between lower bound coring case predictions for the shoed and shoeless piles, indicating internal friction for this pile not to be as high as expected, and only slightly higher than for the shoed pile. Below about 20-m penetration, DP-2 blowcounts were well below those predicted for the DP-2 coring case. Monitoring data and the dropping blowcounts predicted for the lower bound plugged case between 21 and 27 m

480

7

FINAL PILE DESIGN

The final decision was to install 2.591 m diameter pipe piles of 85 mm wall thickness, except for a 90 mm thick, 1.5 m long, driving head and a 95 mm thickened pile sleeve section. The pile tip was made of high strength (440 MPa) steel. The target pile penetration was 29 m, which satisfied both the standard API guidance, and the IAT’s assessment. The design load was assessed as the highest average load expected within any of the four pile groups under the design load condition. All 14 of the jacket’s piles were driven to the same penetration.

8

Figure 4. Docking Pile Predicted and Observed Blowcounts.

penetration indicate either that DP-2 was driving partially plugged throughout this interval, or that friction fatigue was higher than predicted, or possibly a combination of both factors. A review of the data collected during the docking pile installations revealed damping in the very hard Clair clays to be of the order of 0.23 s/m, i.e. similar to “normal” North Sea boulder clays, not at the higher 0.49 s/m value originally adopted. End bearing during driving was close to the static resistance estimates, instead of 1.67 times higher than the estimated static resistance, as originally adopted. The hindcast analyses, which used these revised damping and end bearing factors, indicated a large improvement in predictions if lower residual friction values and greater friction fatigue were adopted, particularly in the very hard clays in the upper 8 metres. Using these adjusted resistances, good lower bound blowcount matches were subsequently achieved for the jacket skirt piles. The marked friction fatigue effects observed during driving are compatible with the cyclic susceptibility observed in the cyclic simple shear testing and the dependency of static capacity on relative pile tip depth, ‘h/R’, that was implicit in the MTD effective stress capacity assessment. Whilst the pile monitoring instrumentation did not function during the re-strike tests, blow-count and hammer energy data indicated increases in shaft resistance (set-up) of 43% and 46% after pauses of only 14.5 and 22 hours. This is consistent with Fugro’s 30 year data-base of driven steel pipe piles in hard North Sea boulder clays (Bhattacharya et al., 2009). It was estimated from this set-up data that shaft frictions would increase by 50% within 24 hours at Clair. © 2011 by Taylor & Francis Group, LLC

OBSERVATIONS DURING JACKET INSTALLATION

Hindcast analyses of the docking pile driving trials allowed a Foundation Acceptance Plan (FAP) to be developed that specified limits to blowcounts, calibrated to hammer performance. Total counts of 2,350 blows, with the MHU 3000 hammer at 85% efficiency, were permitted. Drilling equipment was provided in case hard driving or boulders threatened pile fatigue or premature refusal. All 14 piles drove successfully, although two piles exceeded the 2,350 blow limit by 5 and 15% respectively. However, the hammer was running at less than 85% efficiency and imposed smaller driving stress cycles than anticipated, allowing the fatigue histories to be deemed acceptable. Noting that blowcounts reflect the frictional resistance along the entire shaft, lower bound blowcount rates were applied as the pile tips approached the design penetrations. The ratio of observed to lower bound rates (adjusted for the MHU 3000 hammer energy) was termed the Acceptability Ratio (AR) and values from 1.16 to 1.92 were recorded over the final metre for all 14 piles. Allowing for a minimum of 50% long-term shaft resistance set-up, all piles were expected to meet or exceed the design capacity.

481

9

POST-INSTALLATION VERIFICATION

Post-installation pile verification included comparison, for 10 piles, between hindcast Case Method SRDs from the field force-velocity-time data and the SRDs required to match the required capacity. Using the docking pile data, which showed that long term set up of at least 50% could be assumed, the required long term capacity requirement was reached comfortably in all cases. Signal-matching analyses were conducted for 18 selected hammer blows, including one for each of the ten successfully monitored piles as they approached target depths. A Jc value of 0.2 was used in analysing dynamic resistance. In each case the dynamic resistances exceeded the required static capacities.

Conservative projections that assumed that 60% of the SRD friction was on the external face of the pile, at least 50% long-term set up and a fully-plugged static failure mode led to lower bound static capacity estimates that marginally exceeded the required design capacity. This information, along with SRD increases observed during re-strikes after driving pauses, was important to the overall Foundation Assurance, as described by Evans et al. (2010). 10

CONCLUSIONS

This paper has described how the appropriate design axial capacity values for piled foundations were determined and assured in the unprecedented soil conditions encountered under the Clair-1 platform, West of Shetlands. Specific account was taken of the hard till’s very high shear strengths and potential boulder contents, as well as pile group effects, cyclic loading and the possible effects of pilot hole drilling. Advanced testing, analyses, driving trials and installation observations allowed BP and their partners to be confident in the performance of the foundation piles. ACKNOWLEDGEMENTS The authors are grateful to BP North Sea and its coventurers, ConocoPhillips, Chevron, Shell and Hess for permission to publish this paper and share the unique Clair Phase 1 foundation design and installation experiences with others. REFERENCES Aldridge, T.R., Carrington, T.M. and Kee, N.R. 2005. Propagation of pile tip damage during installation. International Symposium on Frontiers in Offshore Geotechnics, 19–21 September 2005. Bea, R.G. and Audibert, J.M.E., 1979. Performance of dynamically load pile foundations. Proceedings 2nd International Conference on Behaviour of Offshore Structures (Boss 79), Imperial College London, pp. 728–745. Bhattacharya, S, Carrington, T and Aldridge, T. (2009) Observed short term set-up of piles in over-consolidated North Sea clays, Proceedings of the Institution of Civil Engineers: Geotechnical Engineering, 162, (pp. 71–80)

© 2011 by Taylor & Francis Group, LLC

Chow, F. (1997). Investigation into displacement pile behaviour for offshore piled foundations. PhD Thesis, University of London (Imperial College). Evans, T.G., Finnie, I., Little, R., Jardine R.J. and Aldridge, T.R., (2010), ‘BP Clair Phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till, Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia Jardine, R.J.(1994) Review of offshore pile design for cyclic loading: North Sea clays. HSE Offshore Technology Report, OTN 94 157.85 Jardine, R.J. and Chow, F.C.(1996) New design methods for offshore piles. MTD Publication 96/103, MTD, London. Jardine, R.J., Chow, F.C., Overy, R.F. and Standing, J.R. (2005) ICP design methods for driven piles in sands and clays”. Thomas Telford Ltd, London p. 105. Lehane, B.M., Jardine, R.J., Bond, A.J. and Chow, F.C. (1994) The development of shaft resistance on displacement piles in clay. Proc. XIII ICSMFE, New Delhi, India, pp. 473– 476. Lehane, B.M., Chow, F.C., McCabe, B.A. and Jardine, R.J. (2000) Relationships between shaft capacity of driven piles and CPT end resistance. Geotechnical Engineering, Vol 143, No 2, pp. 93–102. Lehane, B.M., Jardine, R.J and McCabe, B.A (2003) Pile Group Tension Cyclic Loading: Field test programme at Kinnegar, N. Ireland. HSE Research Report 101; HSE Books, p. 42. Matlock, H., 1970. Correlations for design of laterally loaded piles in soft clay. Proceedings 2nd Offshore Technology Conference, Houston, Paper No OTC 1204. Quirs, G.W., Little, R.L. and Garmon, S. (2000). A Normalized Soil Parameter Procedure for Evaluating In-Situ Undrained Shear Strength. Proceedings Offshore Technology Conference Houston OTC12090. Randolph, M.F. and Murphy, B.S., 1985. Shaft capacity of driven piles in clay. Proceedings Annual Offshore Technology Conference, Houston, pp. 371–378. Rojas, E. (1993) Static behaviour of model friction piles. Ground engineering, May, pp. 26–30. Stevens, R.S., Wiltsie, E.A. and Turton, T.H., 1982. Evaluating pile driveability for hard clay, very dense sand and rock. Proceedings Offshore Technology Conference, Houston, Paper OTC 4205. Sullivan, R.A. and Ehlers, C.J, (1973) Planning for driving of offshore piles ASCE, JCD, Vol 99, CO1, pp 59–79 Tang, W.H. (1988) Offshore axial pile design reliability. Research Report for Phase 1 of the Project PRAC 89-29B sponsored by API. Toolan, F.E. and Fox, D.A., 1977. Geotechnical planning of piled foundations for offshore platforms. Proceedings of the Institution of Civil Engineers, 1, pp. 221–244.

482

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Photoelastic investigation into plugging of open ended piles J. Dijkstra, E.A. Alderlieste & W. Broere Delft University of Technology, Delft, The Netherlands

ABSTRACT: This paper presents the results of model pile load tests in a transparent photoelastic medium. This medium is composed of broken glass particles in an oil with matching refractive index. The setup allows for quantitative photoelastic measurements in the soil. Reliable interpretation of the stress data in the plug proved to be difficult. In the same test setup, displacements around the pile are measured using digital image correlation (DIC). As opposed to the photoelastic measurement method the DIC method was able to capture the soil deformation in the plug. During monotonic jacking of the transparent pile in loose (n0 = 0.446) and dense (n0 = 0.314) initial conditions only an 8% difference in pile head load was observed. The stress and strain distribution on the other hand show significant differences between the loose and the dense test, both in spatial distribution as in magnitude.

1

INTRODUCTION

Open-ended piles as used for deep offshore foundations can be installed more easily compared to closed ended piles at the penetration depth required for the design tension capacity. During installation of such a pile, given the limited cross section of the pile, only a limited amount of soil is pushed aside, deformed and compacted. However, plugging of soil in the pile can occur as a result of the inflow of soil into the pile. If this occurs, only a limited amount of soil enters the pile during further penetration and the pile will behave more like a closed-ended pile during further penetration. During installation the soil properties and stress state around the pile and in the soil plug are altered. These changes are even more pronounced if plugging occurs. This implies that a prediction of the pile bearing capacity or the driving resistance, based on the undistorted soil properties, should incorporate these installation effects in order to be as accurate as possible. However, before these mechanisms can be predicted reliably first the stress and strain evolution in the soil should be studied more carefully. The stress development in a tubular pipe pile is e.g. studied by De Nicola (1996). He placed strain gauges on several levels on the outside of a model pile in order to monitor the shaft friction distribution and the base load. The shaft friction in a pipe pile can be monitored by adding strain gauges on the inside of the pile (see e.g. Lehane & Gavin 2001 for model pile tests and Paik & Salgado 2003 for model and field tests) or by adding a shear force transducer, see Ogawa et al. (2008). These research efforts do not offer full understanding of the soil behaviour during and after installation. Performing full field stress measurements in natural soils is very difficult. For a decent spatial resolution © 2011 by Taylor & Francis Group, LLC

the size and amount of the required sensors embedded in the simply would negatively influence the characterisation of the stress state. The physical size of the sensors prohibits a realistic failure mechanism. Full field soil stress measurements can only practically be obtained with a photoelastic measurement method. This method uses the birefringent properties of the material for the stress characterisation in the material. In the current paper the photoelastic measurement technique is combined with digital image correlation in order to obtain a stress and strain field of the soil near an open ended pile. The paper is an extension on the work of Dijkstra & Broere (2009), as now displacement fields are measured as well as stress fields. The main aim of the paper is to investigate the plugging mechanism at a more fundamental level rather than directly extrapolating the results to design practice. 2

STRESS AND STRAIN MEASUREMENTS

Whilst photoelasticity has been used extensively to quantify stresses in homogeneous materials (e.g. Coker & Filon 1930, Dally & Riley 1991), the technique is not widely applied as an analogon in granular materials. In the current study the granular material, or soil, is replaced by grains of a photoelastic material. Crushed glass particles are immersed in a liquid with a matching refractive index, in order to prevent light scatter in the sample. The liquid in the current tests is Exxon-Mobil Marcol 82. This technique is similar to Allersma (1982), Drescher (1976) and Wakabayashi (1957). The other properties of this matching liquid, such as viscosity and density, are as close as possible to water. In Allersma (1987) and Dijkstra (2009) the similarity in mechanical behaviour between broken glass and sand is shown using triaxial test results. The

483

Table 1. Seven polariscope configurations for the photoelastic measurement. intensity I1 I2 I3 I4 I5 I6 I7

Figure 1. Arrangement of optical elements in a polariscope.

crushed glass grains, however, are more angular than sand and as a result the grain properties are slightly different (as are strength and stiffness of the grains, which renders the material more prone to crushing). On the whole, they are a reasonable substitute material to investigate sand behaviour. The broken glass in the current test setup has a grain size d50 = 2 mm. The remaining contrast in the sample, due a small mismatch in refractive index between the grains and the liquid, allows for the use of digital image correlation to capture the displacement fields next to the stress fields in the same test setup.

β (rad) π 4

− π4 π 4

− π4 − π4 − π4 π 4

γ (rad) 0 0

− π4 − π4 − π4 π 4 π 4

θ (rad) 0 0 − π4 − π4 π 2 π 2 π 2

with respect to the horizontal plane of the sample, the fast axis of the first λ4 plate has an angle β, the second λ plate an angle γ and the analyzer an angle θ. The 4 seven configurations are shown in Table 1. From the seven measured light intensities the isoclinic angle φ and retardation δ in the sample can be calculated by (1) – (3), for a full analysis one is referred to Dijkstra (2009):

2.1 Stress measurements The photoelastic effect is measured using a transmission polariscope. This apparatus projects light with a pre-defined polarization state on the sample. The stressed sample will alter the polarization of the incident light and this alteration of the light polarization is subsequently measured. The polarization state is derived from the measured light intensity as observed from seven preset orientations of the quarter wave plates and polarizers in the polariscope. The polariscope used in this research consists of two linear polarizers and two retardation plates (also called λ4 plate) and is shown in Fig. 1. The emerging light intensity can be calculated using Jones calculus (Theocaris & Gdouto 1979). The Jones vector of the incident light is multiplied with the Jones matrix of the optical element. Depending on the rotational position of the optical elements, the angle between the fast axis compared to the horizontal axis, the general form of the Jones matrix for the polarizer or retarder can be simplified. By the proper choice of the type, order and position of the optical elements the photoelastic properties in the sample can be derived from the emergent light intensities for at least four pre-set configurations of the optical elements in the polariscope. The proposed optical arrangement and configuration of Yoneyama & Kikuta (2006) is used for the measurement of the photoelastic parameters in the sample, i.e. the retardation δ and isoclinic angle φ in Fig. 1. This method allows to compensate for poor wave plate performance. Using this phase stepping technique a full polarization state is measured from at least seven configurations in which the polarization of the incident light and positions of the second λ4 plate and analyzer are varied. The current method incorporates the retardation  of the non-ideal retarders before and after the sample in the analysis. The fast axis of the first polarizer is set at an angle π2 (counter clockwise) © 2011 by Taylor & Francis Group, LLC

In these equations  is the retardation of the retarders. The sign of sin  is always taken positive as the retardation needs to be positive. At positions where  cannot be obtained for example when I1 = I2 , a representative average value for the domain is taken. Due to the trigonometric functions in (1) – (3) the phase data is still wrapped in the domains − π4 < φ < π4 and −π < δ < π respectively. Physically, for φ this means that the measurement method cannot differentiate between the first and secondary principal stress direction (Dally & Riley 1991). For full field analysis the data, therefore, needs to be further processed before subsequent analysis is possible. In the post processing of the data the phase wrap is corrected using the Goldstein branchcut algorithm Ghiglia&Pritt (1998), as opposed to the Lp algorithm used in earlier work (Dijkstra & Broere 2009). This algorithm is somewhat more timeefficient, whilst offering similar phase-unwrapping performance. Although phase continuity is recovered, sometimes an absolute phase shift is introduced in the data. This absolute shift (multiples of π) is manually corrected in the analysis. After unwrapping the isoclinic angle and retardation data the mechanical stress differences in the sample and the principal stress direction can be obtained. A necessary assumption in this case is that the stress optic law is applicable and that the isoclinic angle, after unwrapping, is equal to the angle of the main principal stress direction. The complete stress state

484

still cannot be derived directly from the measurements, as only the stress difference (σ1 − σ2 ) is derived from the measured photoelastic parameters. The values of the principal stresses need to be separated in order to obtain the complete stress state. Therefore, the individual stress components need to be derived using a full field stress-separation method. In the current research the method proposed by Quiroga & González-Cano (1998) is used. In this procedure the equilibrium equations for plane strain are solved with the measured field data as input (stress difference and principal stress direction). The equilibrium equations can be reformulated into the weighted Poisson equation which in turn is minimized using a multigrid solver. In the stress-separation step a material constant is introduced (4).

Figure 2. Schematized plan view of PE model test setup; laser source, plane strain sample, and photoelastic acquisition.

dimensional case, where Ebiot is a 2×2 matrix. The horizontal strain εxx , the vertical strain εyy , the shear strain εxy and volumetric strain εv are given by

where n is the porosity, the thickness d = 21 mm, the stress-optic constant is taken as 2.7·10−12 Pa−1 (Nissle & Babcock 1973). The wavelength λ of the HeNe laser is 632.8 nm. In both the phase-unwrapping and the stressseparation processing a mask is added in the analysis which masks out unreliable measurement points, e.g. the pile wall. 2.2 Strain measurements For the extraction of the displacement data from the recorded images the minimum quadratic difference method (Gui & Merzkirch 1996) is used for the digital image correlation (DIC).

3

where the subscript denotes to the column i and row j on a rectangular grid with M columns and N rows, and f (1) denotes to an intensity reading in the first image and f (2) is an intensity reading in the second image. The location of the minimum value for Cs,t yields the most probable displacement. The method proves to be more robust in case of illumination and noise differences in the subsequent images. Also the method proves to be more reliable with densely packed particles (soil). In the current research a Gaussian sub-pixel fit is used (Mori & Chang 2003). The strains are derived from the displacement fields after correction for rigid body rotations and translations by polar decomposition of the deformation gradient tensor (obtained from the displacements). This yields the stretch tensor U and subsequently the engineering strains from the biot strain tensor E:

The model test setup is designed to measure the stress field in the soil, composed of broken glass with d50 = 2 mm, near the pile by means of the photoelastic method. The displacement fields in the soil near the pile are measured using DIC, whereas the force on the pile head and the surcharge load are measured with load cells. The setup, as sketched in plan view in Fig. 2, can be divided in three parts. The source of the polarized light (laser source with galvano scanner to project the light on the sample), and the camera for the DIC are located at one side of the sample. The light source is offset from the center line of the sample to reduce reflections and glare effects that interfere with the measurements. The sample is contained in a glass and steel strongbox (H = 400 mm, B = 400 mm and S = 21 mm, and pile width W = 21 mm see Fig. 3) and loaded with two surcharges q1 = q2 = 200 kPa. The square tubular pile (outer dimensions 21 mm with 2 mm wall thickness) is centered in the strongbox , in order to prevent splitting the two side walls, a square tubular pile made out of perspex was used. This material is also photoelasticly sensitive, as a result the photoelastic readings in the pile plug are unreliable as the photoelastic effects in the pile’s front and back wall as well as in the soil plug aremeasured. The compressive stress in the pile walls from the load is influencing

where I is the identity tensor. The Biot strain tensor is composed of the engineering strains, for the two © 2011 by Taylor & Francis Group, LLC

MODEL TESTS

3.1 Test setup

485

Figure 4. Pile head load during installation.

Figure 3. Sketch of the strongbox.

the stress reading in the plug. In contrast the glass walls of the strong-box are only loaded in bending (by the soil in the strongbox), the stresses from bending in the front glass plate cancel out the stress in the back wall. On the other side of the sample the remaining components of the photoelastic setup (PE) are situated. As seen in the sketch only about half of the sample is within the field of vision of the cameras. The pile is completely in view, but only half of the soil surrounding the pile. This improves the spatial resolution of the DIC and PE analysis, as less surface area is covered with the fixed resolution of the camera. The light intensity measurements are made with a standard DSLR camera (Canon EOS 400D). In order to improve the reliability of the intensity measurements taken by this camera, only the raw image data was used, the initial CMOS offset was corrected by using black image substraction. In order to improve the intensity readings in each stress measurement the seven polariscope positions are repeated five times. In this way the positioning errors from the optical elements and the random errors of the camera sensor are reduced. 3.2 Test results

4

The strain and stress evolution around the tubular pile during monotic penetration problem is studied in loose conditions (n0 = 0.446) and dense conditions (n0 = 0.314). In these tests the maximum displacement was 36 mm with displacement increments of 1 mm, every 5 mm a photoelastic measurement was taken. The measured stress state, in the form of the shear stress σxy and the sum σxx + σyy , after 21 mm of penetration is shown in Fig. 5 for the loose and dense test. These plots show the full pile width, but only one side © 2011 by Taylor & Francis Group, LLC

of the soil surrounding the pile. The total height and width is 4.4 times the pile width. The scale is normalized on the pile dimension. The evolution of the pile head load is separately measured and shown in Fig. 4. For a similar region the shear and volumetric strain εxy and εv , which are the counterparts of the shear stress and the sum are shown in Fig. 6. Only an 8% difference in pile head load was observed. This difference is much smaller than the stress difference found from the photoelastic measurements. Possibly, the pile-strongbox resistance was much higher in the loose test. Reasonable values for the stress are only found below the pile base. The nonexisting measured shaft resistance is presumably due to the low photoelastic sensitivity of the broken glass. The readings in the plug are unrealistic. What can be noticed in the stress results is that in the loose test only one pile wall is supporting the load, whilst in the dense test two pile walls transfer the load. The loose initial conditions resulted in a somewhat inclined penetration. The loose test shows more loosening of the volumetric strains than the dense test, whereas the shear strains are comparable in distribution, but higher in magnitude for the dense test. The distorted zone in the stress results (with high stress magnitudes) is also found in the strain results. Unfortunately, the analysis of the stress evolution is still an ongoing research effort and could therefore not be presented.

CONCLUSIONS

During monotonic jacking of the transparent pile in loose (n0 = 0.446) and dense (n0 = 0.314) initial conditions only an 8% difference in pile head load was observed. This corresponds well with the observed stress level in the assembly below the pile base as obtained from photoelastic measurements. The stress and strain distribution on the other hand show significant differences between the loose and the dense test, both in spatial distribution as in magnitude.

486

Figure 5. From top to bottom: σxy;loose , (σxx + σyy )loose , σxy;dense & (σxx + σyy )dense ; loose: n0 = 0.446 (top), dense: n0 = 0.314, compressive stress is negative; W = pilewidth = 21 mm.

© 2011 by Taylor & Francis Group, LLC

Figure 6. From top to bottom: εxy;loose , εv;loose , εxy;dense & εv;dense ; loose: n0 = 0.446, dense: n0 = 0.314, negative volumetric strain is contraction; W = pilewidth = 21 mm.

487

ACKNOWLEDGEMENTS This research was financially supported by a grant from the international press-in organisation (IPA, http://www.press-in.org/), which is greatly acknowledged. REFERENCES Allersma, H. (1982). Determination of the stress distribution in assemblies of photoelastic particles. Experimental Mechanics 22(9), 336–341. Allersma, H. (1987). Optical analysis of stress and strain in photoelastic particle assemblies. Ph. D. thesis, Delft University of Technology. Coker, E. & Filon, L. (1930). A Treatise on Photoelasticity. Cambridge: Cambridge University Press. Dally, J. & Riley, W. (1991). Experimental Stress Analysis Third Edition. Singapore: McGRAW-HILL. De Nicola, A. (1996). The Performance of Pipe Piles in Sand. PhD Thesis, The University ofWestern Australia, Perth, Australia. Dijkstra, J. (2009). On the Modelling of Pile Installation. Ph. D. thesis, Delft University of Technology. Dijkstra, J. & Broere, W. (2009). Experimental investigation into plugging of open ended piles. In Proceedings of the ASME 28th International Conference on Ocean, Offshore and Arctic Engineering, Number OMAE2009-79299. Drescher, A. (1976). An experimental investigation of flow rules for granular materials using optically sensitive glass particles. Géotechnique 26(4), 591–601. Ghiglia, D. & Pritt, M. (1998). Two-Dimensional Phase Unwrapping. New York: John Wiley & Sons, Inc.

© 2011 by Taylor & Francis Group, LLC

Gui, L. & Merzkirch, W. (1996). A method of tracking ensembles of particle images. Experiments in Fluids 21(6), 465–468. Lehane, B. & Gavin, K. (2001). Base Resistance of Jacked Pipe Piles in Sand. Journal of Geotechnical and Geoenvironmental Engineering 127(6), 473–480. Mori, N. & Chang, K.-A. (2003). Introduction to mpiv. http://sauron.civil.eng.osaka-cu.ac.jp/ mori/. Nissle, T. & Babcock, C. (1973). Stress-optical coefficient as related to glass composition. Journal of the American Ceramic Society 56(11), 596–598. Ogawa, N., Ishihara, Y., Yokotobi, T., Kinoshita, S., Nagayama, T., Kitamura, A., & Tagaya, K. (2008, Dec). Soil Plug Behaviour of Open-Ended Tubular Pile During Press-In. In Proceedings of 2nd IPA International Workshop, New Orleans, Lousiana, pp. 15–22. International Press-In Assocation. Paik, K. & Salgado, R. (2003). Determination of Bearing Capacity of Open-Ended Piles in Sand. Journal of Geotechnical and Geoenvironmental Engineering 129(1), 46–57. Quiroga, J. & González-Cano, A. (1998). Stress separation from photoelastic data by a multigrid method. Measurement Science and Technology 9(8), 1204–1210. Theocaris, P. & Gdouto, E. (1979). Matrix Theory of Photoelasticity. Berlin: Springer. Wakabayashi, T. (1957). Photoelastic method for determining of stress in powdered mass. In Proceedings of the seventh Japanese National Conference on Applied Mechanics, pp. 153–158. Yoneyama, S.&Kikuta, H. (2006). Phase-stepping photoelasticity by use of retarders with arbitrary retardation. Experimental Mechanics 46(3), 289–296.

488

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Soil-pile interaction during extrusion of an initially deformed pile C.T. Erbrich, E. Barbosa-Cruz & R. Barbour Advanced Geomechanics, Perth, Australia

ABSTRACT: During installation of a pile into soil, an initial imperfection from the theoretical pure cylindrical geometry can progressively grow with increasing pile penetration when the stiffness of the surrounding soil exceeds the elastic stiffness of the pile. Eventually, even small initial imperfections can develop to the point where plastic yielding of the pile may occur, ultimately leading to total collapse. This paper addresses this problem by means of a specially developed numerical model (BASIL) which is implemented in a Python script that incorporates all the physics of the pile-soil model and which fires a procession of ABAQUS finite element analyses as required. The structural model of the pile is extruded as it penetrates into the soil, which is represented by a series of fully non-linear ‘p-y’ springs. The implemented numerical algorithm allows analyses of stratified soil profiles and incorporates pile tip forces acting on any tip chamfer type. The BASIL model is described in this paper and an example analysis is presented.

1

INTRODUCTION

At the end of 1992, the Goodwyn A (GWA) platform was installed on the North West Shelf of Australia. During installation of the drilled and grouted foundation piles for this platform it was discovered that 15 of the 20 driven ‘primary’ piles were severely crushed at depths of 80 metres or more below the seabed (Fig. 1; Barbour and Erbrich, 1994). In considering possible mechanisms for the GWA collapse, an ‘extrusion’ type of mechanism was conceived whereby an initial imperfection in the pile would have been forced to grow as it was pushed into a soil of higher stiffness than the pile. At the same time, the first and third authors were engaged in the design of bucket foundations in the North Sea and it became apparent that a similar mechanism might also occur during installation of the thin skirts of these foundations. While bucket foundations are much shorter and ‘squatter’ than the GWA piles, they are many times more slender and subject to high inward radial pressures due to the differential water pressures (‘suction’) induced by pumping water from the skirt compartment to aid penetration. To address this issue a soil ‘extrusion’algorithm was devised and a user element (BASIL – Bucket Adjusted Soil Installation Loading) developed to model this behaviour in the ABAQUS finite element code (Barbour and Erbrich, 1995). The BASIL element was used in conjunction with a shell based structural model of the bucket foundation. The model was carefully validated and then used extensively for verification analyses of the skirt installation process on the Europipe 16/11E buckets and for design of the Sleipner SLT bucket foundations. © 2011 by Taylor & Francis Group, LLC

489

Figure 1. Goodwyn A primary pile collapse details.

Attempts to model pile geometries with the original BASIL model were unsuccessful. However, the BASIL model has since been extensively reworked and this objective has now been achieved. As an aside it is noteworthy that Earl (2002) presents a finite element model based on the same principles as used in BASIL for modelling extrusion buckling. A series of parametric analyses were performed and presented along with a superficial investigation into the pile collapse at GWA. However, Earl’s model only considered linear elastic behaviour for both the pile and the soil, which are both inappropriate assumptions for this particular problem. In addition, forces acting at the pile tip and on any tip chamfer were ignored and only very simple soil profiles were considered (i.e. no discrete

layering). Aldridge et al. (2005) also considered this problem in a conceptually similar way, but using only a very simplified analytical implementation. 2. 2 2.1

DESCRIPTION OF BASIL MODEL 3.

Original formulation

The original BASIL model is described in Barbour and Erbrich (1995). The model comprised two parts; a structural finite element model (i.e. shell elements) of the foundation and the BASIL user element, which comprised a ‘brush’ of radial ‘hairs’ that radiated from the centre line through the foundation. Each ‘hair’ defined the location of a p-y soil spring, which acted on the structural model. The origin of each spring was defined by the position of the tip of the skirt or pile as it cut each hair. A permanent (plastic) initial imperfection was imposed on the skirt or pile, which was then pushed into the ‘brush’. Any growth of this imperfection was tracked to establish whether this ‘extrusion’ through the soil would lead to structural collapse. Experience using this model revealed a number of significant difficulties, which principally originate from the fact that the model was implemented into ABAQUS as a user element, written in Fortran: – Definition of soil parameters had to be done on a case-by-case basis, with specific Fortran code written for each new case. This was cumbersome and had a high risk of errors. – A stiffness matrix had to be defined for the BASIL element. This was achieved using numerical differencing, which was an approximate and inefficient process that was found to greatly slow the analysis. – Whilst the ABAQUS solver is very effective in many cases, it doesn’t deal particularly well where instabilities develop in a model. Experience showed that after a certain level of imperfection growth had occurred, it would be impossible to progress the analysis any further. – Using a pile sliding past discrete ‘hairs’ tended to set up oscillation patterns in the mobilised soil pressures, which were caused by the faceted nature of the linear shell elements used to model the pile. These oscillations were considered undesirable and unrealistic.

4.

5.

6.

This approach directly solves all the problems identified with the original algorithm:

For the work presented in this paper the model has therefore been completely re-written albeit the underlying physics remains identical. 2.2

New formulation

The ABAQUS user element is abandoned in the new formulation and is replaced with a Python script that incorporates all the physics of the BASIL element, and also fires a procession of ABAQUS analyses as required. The stages in each analysis are: 1. Define the pile geometry, initial imperfection and boundary fixities as per the ‘original formulation’. © 2011 by Taylor & Francis Group, LLC

However, in the new approach the pile is modelled with a uniform (user defined) element length up the pile. For the defined initial imperfection, use the Python script to determine an initial set of nodal forces to apply to the bottom pile element, assuming that this has penetrated fully into the soil. Trigger anABAQUS analysis to determine the compatible pile displacements for the applied set of nodal soil loads and then output these results to the Python script. Use the Python script to determine a new set of soil forces that are compatible with the updated pile displacements and compare these to the soil forces determined in stage 2. If the difference between the computed nodal forces is less than a defined tolerance then the step has completed. If the nodal force difference exceeds the allowable tolerance then trigger another ABAQUS analysis with this revised set of nodal loads and obtain a new set of pile displacements. Repeat stages 4 and 5 until convergence of nodal forces is obtained After convergence is obtained the analysis proceeds to the next step; this requires advancing the pile another element length into the soil and proceeding from stage 2 to stage 5 above. The only difference compared to the first step is that the deformed geometry obtained at the end of the previous step is used as the starting point for assessing the appropriate soil nodal forces for the next step and nodal forces are also defined for the next element up the pile.

490

– Any combination of pile or soil can be configured rapidly in a simple input file. – An iterative solution technique is used and hence no stiffness matrix is generated. The BASIL soil model is solved exactly at all points, rather than through an approximate numerical procedure. The iterative solution technique also means that the ABAQUS solver is not relied on for convergence of the BASIL user element. The ABAQUS solver is only used to determine the solution for the relatively simple problem of a pile shell model subject to defined nodal loads. – Since the solution advances in steps that are one element long, there is no longer any sliding past faceted elements and hence oscillations in soil pressure are effectively eliminated. Despite these advantages, implementation of the new approach proved extremely challenging. The main difficulty centred on the iterative solution technique. Many attempts were made to develop a robust iterative solution technique that would guarantee convergence and eventually after much trial and error we found a procedure that has proven highly successful. Amongst other things, this includes random generation of some of the iterative solution variables in order to prevent the solution getting ‘stuck’ in ‘local minima’.

One major advantage of the new formulation is that the problem can be stopped and restarted at any point, and the solver can be changed between different steps. In the example presented later we invoked a dynamic analysis procedure for a substantial part of the analysis, in which the nodal soil forces were applied to the pile over a time period of 0.04 seconds. This enabled the analysis to proceed into uncharted territory, with almost complete collapse of the pile demonstrated. This would have been impossible with the original BASIL user element since this had no dynamic formulation and the original ABAQUS solver would have failed at a much earlier stage due to the intrinsically unstable nature of the solution at such large deformations. It might have been expected that a dynamic procedure would slow down the propagation of tip deformation somewhat since some component of the applied nodal soil force would be balanced by the inertia forces. However, inspection of the rate of tip deformation suggests this to be a negligible effect; the dynamic analysis simply seems to have added enough ‘damping’ into the system to allow the iterative solver to find the correct, essentially static, solution. 2.3

Figure 2. Ramberg-Osgood model for external BASIL spring.

External BASIL soil spring

As discussed above, the BASIL model includes an external soil spring model (‘p-y’), which represents the response of the soil into which the pile is being penetrated. The original BASIL implementation adopted a simple linear elastic perfectly plastic soil spring model. However, the new formulation uses a fully non-linear Ramberg-Osgood (R-O) form of spring, with the stiffness for ‘unloading’paths defined as the initial (linear) stiffness of the R-O model (Fig. 2). As will be demonstrated in the next section this spring type can give a good match to the type of pressure-displacement response anticipated for the soils in this case. Note that the R-O springs only act in compression, being set to zero if they are displaced inward from their initial position. This implies that any effect from the internal plug has been ignored. The influence of the internal soil plug was explicitly evaluated in some cases, but was not found to be a significant contributor to the overall response. 3 VALIDATION OF BASIL SPRINGS The external BASIL soil springs have been determined assuming that cylindrical cavity expansion conditions apply. A series of small strain cylindrical cavity expansion analyses were performed using ABAQUS to validate this assumption and to calibrate the springs used. The first analysis comprised a validation of the FE model using an elastic perfectly plastic (Tresca type) model for the soil domain. The p-y response from the FE analysis was compared with the theoretical solution proposed by Carter, Booker andYeung (1986) and © 2011 by Taylor & Francis Group, LLC

491


Figure 3. Normalised shear strength of su /σv0 = 0.5.

Figure 4. p-y
= 0.5). (su /σv0

response

from

cavity

expansion

in Houlsby and Carter (1993) and was found to be essentially identical. For the same normalised undrained shear strength
of 0.5, a second FE analysis was performed of su /σvo using the generalised stress-strain properties defined on Fig. 3, which were considered more realistic for the soil under consideration. The resulting p-y response is presented on Fig. 4 along with the equivalent p-y response obtained for the simple linear elastic perfectly plastic Tresca model. Despite the fact that the same ultimate strength is adopted in both cases, it may be seen that the latter model gives a significantly softer response at all stress levels.

Figure 5. FE elliptical cavity into circular cavity analysis.

Figure 6. σr for elliptical cavity into circular cavity analysis.

Since a collapsing pile implies a progressively smaller cavity radius to be expanded, we have also investigated the influence of the initial cavity diameter on the final p-y response. Three small strain FE analyses were performed with initial cavity diameters ranging from 0.75 m to 2.65 m using the elasto-plastic Mohr-Coulomb hardening material shown on Fig. 3. From these analyses it was found that the cavity expansion p-y response is independent of the initial cavity diameter provided the wall displacement from the p-y curves is normalised by the initial cavity diameter. Based on these results, the R-O springs used in the BASIL model were therefore defined in a normalised displacement format. The final parametric study investigated the applicability of the cylindrical cavity expansion analysis to the actual pile buckling problem, where the applied soil springs involve deformation of an initial noncircular (elliptical) cavity back into a circular cavity. Two FE analyses employing an elastic perfectly plastic (Tresca) soil domain were performed for this purpose. The first was a small strain FE analysis of cylindrical cavity expansion starting from a nominal initial diameter (D0 ) of 1.325 m (the dashed line on Fig. 5). The second analysis was a large deformation finite element analysis starting from an initial elliptical section (the dotted line on Fig. 5). Both cases were expanded into the circle shown by the solid line on Fig. 5. In the latter case the perimeter of the evolving elliptical cavity section was also kept constant during the analysis, which is also likely to be the case for a collapsing pile. The radial stress (σr ) averaged across the soil element closest to the minor axis for the initial elliptical cavity geometry has been compared to the radial stress for the equivalent element in the initial circular geometry case. For the cylindrical cavity expansion case σr was found to be pretty uniform within the element, but varied significantly between the different gauss points for the ellipse-to-circle case. However, as shown on Fig. 6, the average value of σr for the four gauss points in this latter case exhibits a similar response, as a function of displacement, to that obtained in the cylindrical cavity expansion analyses. This finding supports the assumption made in deriving the BASIL springs that cylindrical cavity expansion conditions may be (approximately) assumed to apply.

4

© 2011 by Taylor & Francis Group, LLC

4.1

EXAMPLE ANALYSIS Structural model

The structural model of the pile has been developed using the ABAQUS S4R element, which is a 4-node doubly curved shell element with reduced integration, hourglass control and a large strain formulation (including finite membrane strains). These are a general purpose shell element using thick shell theory when the shell thickness is large and thin shell theory when the shell thickness is small. Only a quarter of the pile cross-section is modelled and hence 2-fold symmetry is assumed. The pile model used is 50.0 m long with an outer diameter of 2.65 m. Around the circumference of the pile, each element covers an 8.2◦ arc. Vertically, 0.2 m long elements were used over the full length of the pile. A homogeneous pile wall thickness of 45 mm has been adopted. The pile steel has been modelled as elastic-plastic with aYoungs Modulus of 207 GPa and a yield strength of 420 MPa. A post-yield hardening response was also included, equivalent to a modulus of 5 GPa, which is around 2.5% of the elastic modulus, which is considered reasonable for actual steel. The pile is pushed into the soil through displacement boundary conditions applied to the top and hence any applied vertical tip loads act over the entire length of pile modelled. No frictional resistance along the pile shaft was modelled.

492

4.2

External soil springs

For the example case considered here, a suite of FE cavity expansion analyses were performed to determine p-y curves for a number of generalised stressstrain curves, similar to those presented on Fig. 3,
. A best-fit R-O curve encompassing a range of su /σvo was then determined for each of the resulting ‘p-y’ curves, as in the example shown on Fig. 2, and the various R-O parameters were generalised such that appropriate values could be selected for any given
su /σvo . A strong cemented layer was also included in the analysis for which a separate FE analysis was performed to obtain the required R-O parameters.

Figure 7. Ramberg-Osgood parameters.

The R-O parameters adopted in this analysis are summarised on Fig. 7. Figure 8. Chamfer and tip stresses.

4.3 Tip stresses A chamfer at the pile tip has also been included in the example analysis. In theory a BASIL model could be generated that would automatically account for ‘wedging’ behaviour where a pile tip chamfer is penetrated into the soil. However, in practice this is not readily achieved due to the relatively steep inclination of a tip chamfer, as opposed to the much more gradual inclination expected of a pile wall that is undergoing extrusion buckling. Hence to deal with this problem, we separately estimate any pile tip chamfer forces and impose these, combined with the vertical tip loads, as a set of nodal forces at the tip of the pile. These chamfer/tip forces vary with changes in soil layering as the pile penetrates into the soil. The vertical pile tip forces are generally estimated as being equal to the net cone resistance in uncemented soil layers but rather lower in cemented soils, where the driving process tends to propagate fractures ahead of the pile tip. The same vertical tip pressure is assumed to apply over both flat and chamfered tip sections. However, for the latter, the vertical tip force must be transformed into a normal and frictional stress component acting on the chamfer face itself. The frictional component is determined using an interface friction angle, δ. The profiles of chamfer stresses (q1 ) and vertical tip stress (q2 ) used in the example analysis are presented on Fig. 8. 4.4 Initial imperfection The initial imperfection was defined based on the mode shapes derived from an eigenvalue extraction analysis and was used to adjust the initial pile geometry. The first buckling mode shape assuming a uniform inward radial ring load applied at the pile tip was used for the example analysis. The imperfection magnitude is defined as the maximum change in pile radius relative to the nominal pile radius. For the example analysis a 25 mm initial imperfection was defined (ie. 100 mm ovality). © 2011 by Taylor & Francis Group, LLC

Figure 9. Track of pile tip.

4.5 Analysis results A plot of the of the pile tip radius (‘track of pile tip’) is shown on Fig. 9 for the example analysis. Two lines are presented, one representing the minor axis of the pile and one the major axis. The track of the pile tip defines the origin for each of the BASIL springs at any depth. So if the actual pile radius is larger than defined by the track of the pile tip at any depth, a soil pressure will be imposed on the pile. Conversely, where the actual pile radius is smaller than defined by the track of the pile tip, no soil pressure will be mobilised since this means that the pile is moving away from the soil; as noted earlier no internal soil plug resistance is included. It can be seen that after an initial increase in tip displacement at the start of the analysis, a ‘steady state’ condition appears to arise just prior to the pile entering the well cemented calcarenite layer. On entering this hard layer the tip displacement rapidly increases, attaining a maximum of 48 mm inward displacement per metre of penetration at the base of this layer. After exiting the hard layer the tip displacement continues to increase, but initially at a much reduced rate of only about 9 mm per metre of penetration. However, the pile tip displacement never stabilises and rebounds; instead the pile is progressively crushed as it continues to penetrate into the soil. Down to a penetration of about 35 m, the tip displacement rate is relatively constant

493

at around 15 mm per metre, but starts to accelerate once again at greater depths. Initially, this acceleration seems to be initiated by the somewhat stronger layer which commences at a depth of about 35 m. However, we believe that the continued acceleration in tip displacement thereafter, even when this stronger layer has been left well behind, is a function of the softening in the structural response of the pile at the very large deformations that have now been imposed. At the end of the analysis, at a pile penetration of 45.6 m, the pile tip displacement has reached a rate of about 70 mm per metre of penetration, the total inward tip displacement is 761 mm and the final pile radius on the minor axis is 539 mm (ie. only 41% of its initial value). The pile deformed geometry, plotted at true scale, is presented on Fig. 10 at the final penetration depth. To aid visualisation this figure is derived using a feature in ABAQUS that allows the ‘full’ model to be plotted based on the stated axes of symmetry. In addition, the yielded and unyielded parts of the pile are shown by differential colour shading. It can be clearly seen that the pile has ovalised into a ‘peanut’ configuration and extensive plastic yielding has occurred along the pile.

processes could be allowed for by adopting a larger initial imperfection than might otherwise be assumed. Alternatively, the tip end bearing could be enhanced, thereby making some allowance for the inertia effects associated with driving the pile tip. Either way, it is recommended that the initial imperfection assumed in such an analysis should always exceed ‘normal’ fabrication tolerances.

5

REFERENCES

Figure 10. Pile deformed geometry at final penetration depth.

CONCLUSIONS

This paper has presented and demonstrated a sophisticated model that can be used to assess the risk of pile extrusion failure during penetration of an open ended pile into the ground. Whilst not commonly recognised, several major pile failures have arisen due to such behaviour in the past; the aforementioned GWA and the Valhall IP platform piles (Alm, et al. 2004) being the most notable examples. Even when complete collapse is not anticipated major problems might arise with only a small degree of ovalisation, if for example, some other tool had to be subsequently passed down the pile (e.g. if drilling out the pile). It is recommended that this type of analysis should be performed as a matter of course in such cases. When assessing the appropriate magnitude of initial imperfection for any given case it should be appreciated that the analysis does not explicitly account for dynamic pile driving effects; essentially the pile is steadily jacked into the ground. Model tests have indicated that driving rather than jacking is a more damaging process. Earl (2002) proposed that such

© 2011 by Taylor & Francis Group, LLC

Aldridge, TR, Carrington, TM, & Kee, NR. Propagation of Pile Tip Damage during Installation, Proc. International Symposium on Frontiers in Offshore Geotechnics, Perth, September 19–21, 2005. Proceedings. Eds: Gourvenec, S, Cassidy, M. Alm, T., Snell, RO., Hampson, K., and Olaussen, A. (2004). Design and Installation of the Valhall Piggyback Structures, Proc. Offshore Technology Conference, OTC16294, Houston. Barbour, R.J. & Erbrich C. (1994). Analysis of In-situ Reformation of Flattened Large Diameter Foundation Piles Using ABAQUS, UK ABAQUS Users Conference, Oxford, September 1994. Barbour, R.J. & Erbrich C. 1995. Analysis of Soil Skirt Interaction during Installation of Bucket Foundations Using ABAQUS, Proc. ABAQUS Users Conference, Paris, June 1995. Carter, J.P., Booker, J.R. and Yeung, S.K. (1986). Cavity Expansion in Cohesive Frictional Soils, Geotechnique, 36(3): 349–358. Earl R.J. 2002. Growth of Imperfections in Piles During Installation, PhD Thesis, University of Western Australia. Houlsby GT and Carter JP (1993). The Effects of Pressuremeter Geometry on the Results of Tests in Clay, Geotechnique 43(4), 567–576.

494

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

BP Clair phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till T.G. Evans BP Exploration Operating Company

I. Finnie Advanced Geomechanics (Formerly of Lloyds Register)

R. Little Fugro GeoConsulting, Inc. (Formerly of Fugro-McClelland)

R.J. Jardine Imperial College, London

T.R. Aldridge Fugro GeoConsulting Limited

ABSTRACT: BP’s Clair Phase 1 Platform is the first fixed structure on UK’s Atlantic Margin. The conventional steel platform is in 140m of water and is supported on groups of steel pipe piles driven into bouldery glacial clays with undrained shear strengths of up to 2000 kPa. These extreme conditions posed significant challenges for foundation engineering that were managed by a systematic process of design, design assurance and performance monitoring. This paper summarises the project history and foundation risk management process. The foundation engineering is described in more detail by Aldridge et al. (2010).

1

INTRODUCTION

The Clair field is about 75 km off the west coast of the Shetland Islands, UK Sector, North Sea (Fig. 1). Discovered in 1977, it is one of the largest hydrocarbon resources off northwest Europe, with 6 to 7 billion barrels of oil equivalent (bnboe) of reserves. Recoverable oil is limited by a fractured reservoir and low API gravity fluids, so development was not considered feasible until the 1990s, when advances in seismic imaging and engineering technologies offered economic production.

BP (with a 28.6% stake) and its co-venturers, ConocoPhillips (24.0%), Chevron (19.4%), Shell (18.7%) and Hess (9.3%) are developing the field in phases, the first of which extended from 1996 to 2005, as shown on Fig. 2. The second phase, Clair Ridge, is in the conceptual design stage and is expected to come on stream in late 2014. Clair Phase 1 comprised a single fixed steel drilling and production platform in about 140 m of water. Oil is exported to the Sullom Voe terminal in the Shetland Islands through a 22-in, 105 km long pipeline. The Clair Ridge development is about 6 km to the northwest of Clair Phase 1 and will consist of

Figure 1. Location of Clair Phase 1 Platform.

Figure 2. Nine year development of Clair Phase 1.

© 2011 by Taylor & Francis Group, LLC

495

Figure 4. Genesis of clair soils.

wall thickness pipe piles that were pre-installed with the template in summer 2003. The docking piles and 14 platform piles were driven to a depth of 29 m below the seabed using a Menck MHU 3000 underwater hydraulic pile driving hammer.

Figure 3. Clair jacket foundation layout.

two bridge-linked steel platforms – one for drilling and production, the other for quarters and facilities. The Clair field soils consist principally of hard clay tills, with gravel to boulder-size rock inclusions. The undrained shear strengths of the clay tills exceed those at all previous offshore sites where steel pipe piles had been driven successfully. This paper summarises the approach taken by BP to design and assure driven pile foundations for the Clair Phase 1 Platform in these exceptional conditions. It focuses on the project history, site characterisation and design and assurance process. The detailed engineering is covered by a companion paper to this Symposium (Aldridge et al., 2010).

2

CLAIR PHASE 1 PLATFORM

The Clair Phase 1 platform is the first fixed structure on the UK Atlantic Margin and is located about 100 km northeast of BP’s FPSO-based developments at Foinaven and Schiehallion (Fig. 1). The structure comprises an 8,800 t four-leg steel jacket with an 11,700 t integrated topsides deck. The jacket is founded on 14 No 2.59 m (102 in) diameter, 85–95 mm wall thickness open-ended steel pipe piles. The platform is designed for 100 year directional storm loads and a 25 year life. The foundations are asymmetrical, with each of the legs on the eastern side of the jacket supported on four skirt piles and the western legs each carried by three skirt piles (Fig. 3). The jacket and topsides were designed by Mustang Engineering Inc. (MEI) and fabricated at Aker Kvaerner’s Verdal yard in Norway. Fugro-McClelland Marine Geoscience (FMMG) was a specialist consultant to MEI for foundation design. The facilities were installed by Saipem UK Ltd. in late 2004 and early 2005 using their S7000 heavy lift vessel. The jacket was docked over a 28 slot drilling template using two 1.829 m diameter 75 mm © 2011 by Taylor & Francis Group, LLC

3 3.1

GEOLOGY Regional

The Clair field lies on the West Shetland Continental Shelf which was glaciated repeatedly during the mid to late Pleistocene Epoch. The platform is in the terminal moraine zones of two coalescing Weichselian-age ice sheets, as indicated on Fig. 1. The geological knowledge of the area pre-1997 was largely based on regional studies by the British Geological Survey (BGS) and exploration-quality seismic data and data from drilling hazard geophysical site surveys. Desk studies in 1997 indicated the depositional environment to be very complex, with the shallow geological conditions varying laterally and vertically. The near-surface soils of interest for platform foundations were inferred as lodgement tills and inter-morainal sediments that have been compressed and sheared by hundreds of metres of ice (Fig. 4). The soils were predicted to be very dense and hard and to contain cobbles and boulders. This prognosis was generally supported by the large number of anchor-dragging incidents and well top-hole re-spuds reported in the Clair field. 3.2

Field-specific

The first engineering-quality geophysical survey and intrusive geotechnical investigations were carried out in April and June 1997, respectively. The work was performed around the preferred platform location at the time, Site A. The main purpose of the geophysical survey was to help develop a regional stratigraphic model and to detect shallow geohazards, including boulders. The techniques used included side scan sonar for imaging the seabed, a Deep Tow Sparker (DTS) for shallow subsurface imaging (<40 m) and multi-channel UHR for deeper penetration (∼140 m). The survey succeeded in identifying the main soil boundaries for pile design

496

but was less successful in providing details of shallow soils, as a result of high acoustic impedances and poor DTS signal penetration. Gravel, cobbles and boulders were seen on the seabed, and a number of subsurface diffraction hyperbolae were detected at shallow depths but it was not possible to determine whether these signals were due to boulders and/or other features such as in-filled channels. The objectives of the geotechnical investigation were to calibrate the results of the geophysical survey and to provide geotechnical data for engineering foundations at Site A. The work was carried out by Fugro from the MV Norskald using standard offshore geotechnical equipment, comprising 5 inch API drill pipe with an open-centre drag bit, pushed and driven wireline sampling tools and 10 cm2 piezocones. The investigation proved to be very difficult and was curtailed prematurely due to a combination of worn drill bits, sheared drill pipes, poor sample recovery and numerous piezocone penetration test max-outs. The main conclusion drawn from the work was that the shallow sediments are mostly low plasticity un-cemented glacial clays with unit weights up to 24 kN/m3 (similar to concrete) and uniaxial strengths comparable to those of very weak and weak rocks. The investigation also revealed subsurface gravels and cobbles, but no direct evidence of boulders. 4

to occur singly or in clusters, and at any location or depth. 4.3

1. A range of foundation types including a gravity base, drilled and grouted piles, suction-installed caissons and hybrids. 2. Static capacity and drivability of steel pipe piles. 3. Effects of boulders on driven piles and the potential for premature refusal and tip damage. A conventional gravity-base structure was selected as the best foundation solution but was ruled out for non-geotechnical reasons. Driven piles were assessed as feasible by extrapolation of North Sea hard clay experiences and were considered to be the next best option. 4.4

4.1 Strategy

4.2 Geological review Further geological assessments of the Clair area were carried out by Fugro Limited and BGS in early 1998. This work drew on BGS’s regional knowledge and the shallow surveys performed by BP at Site A in 1997. The main objective was to develop a predictive geological model for the Clair field with special focus on the origins, nature and expected distribution of boulders. The study concluded that the high shear strengths at Clair could be explained by depositional and postdepositional histories but that the shallow soils were chaotic and highly variable. Boulders were predicted © 2011 by Taylor & Francis Group, LLC

Platform site options 1998, sites A, B and C

By early 1998, interest in Site A had reduced, and two other sites, Sites B and C, were also being considered. Site B, centred on an existing keeper-well about 1 km to the northwest of site A, was the new preferred location and Site C was a fall-back, about 600 m southeast of Site B. The final choice of site was put on hold pending the results of a further geophysical survey and conclusions about boulders.

PRELIMINARY ENGINEERING (1997–1998)

The presence of extremely hard clays and boulders clearly posed significant problems for the design and installation of the Clair platform foundations, especially for driven piles, and an Independent Assurance Team (IAT) of experienced cross-industry geospecialists and foundation engineers was engaged to advise the Project on how to manage these challenges. This team comprised specialists from BP, Fugro Limited, Brown & Root, Lloyds Register, Imperial College, the Norwegian Geotechnical Institute (NGI), BGS, Heerema and Saipem UK. Their main roles were to: (1) provide expert consensus views, (2) perform geo-engineering studies to improve the project’s fundamental understanding of geotechnical risks and (3) design geophysical and geotechnical surveys suitable for the Clair soil conditions.

Foundation engineering studies

Various engineering studies were performed throughout 1998 by Fugro Limited, NGI and Brown & Root. The work was based on SiteA conditions, and covered:

5 5.1

BESPOKE SURVEYS (1998 & 2000) Geophysical survey

A second engineering-quality geophysical survey was performed in the Clair field in March and April 1998. The main purpose of this survey was to investigate and compare the near-surface geology across Sites A, B and C. The main objective was to infer the numbers and spatial positions of boulders beneath the candidate sites. Additional seismic sources including a Pinger (chirp) and a Deep Tow Boomer were added to the survey spread to try to improve on the imaging achieved during the 1997 survey. The results were disappointing but were sufficiently reliable to conclude that boulders were not a significant site discriminator. Site B was chosen in mid 1998. 5.2

Geotechnical SI

A geotechnical investigation was planned for Site B for the summer of 1998, immediately after the 1998 geophysical survey. However, in May 1998 the project was put on hold following a significant dip in the oil price, and the investigation was postponed. The project was restarted in January 2000 and a purpose-designed geotechnical investigation was carried out at Site B in May 2000. The work was

497

performed by Fugro from the MV Bucentaur using techniques designed to cope with the anticipated soils. Five boreholes were drilled to depths of 48 m to 54 m using Seacore’s piggy-back Geobor S triple tube wireline core coring system with a PCD Diapax bit and a polymer mud. These produced nominal 1.5-m long, 102-mm diameter cores. A further eight 5-in API holes were drilled to a maximum depth of 83 m using a Stratapax diamond bit. The API phase included downhole wireline push sampling, 10 cm2 PCPTs and 5 cm2 CPTs. API-type push sample recovery was poor but good quality samples were obtained with the Geobor S equipment. Rock cores up to 300-mm long were recovered in some of the boreholes, confirming the presence of boulders. Cone point resistances were generally above 60 and locally up to 120 MPa in the top 14 m below seabed. Cone resistances of between about 40 and 60 MPa were measured below that depth. The soils encountered at Site B comprised 0.1 to 1.0 m of surface sand over low plasticity sandy clays. The undrained shear strengths of the clays were typically 1200 to 2500 kPa in the top 12 m below seabed and generally between 400 and 800 kPa below that depth. A 1.0 m thick layer of very dense sand was encountered at about 24 m depth. The saturated unit weights of the soils were consistently between 23 and 24 kN/m3 . 6

FOUNDATION ENGINEERING (2000–2004)

6.1 Engineering challenges The results of the surveys performed at Site B in 1998 and 2000 confirmed the geotechnical challenges inferred from earlier investigations: 1. Undrained shear strengths and su /σv
ratios for the soils are well outside the ranges measured at the pile test sites that form the API RP2A (2000) database, invalidating the traditional empirical method for estimating axial capacities of driven steel pipe piles in clays 2. Soil strengths exceeded those at sites of previous North Sea pile driving experience, and 3. Boulders could not be discounted. Of greatest concern was that hard driving through strong soils and/or rock obstructions might result in pile tip damage, excessive pile fatigue and/or premature pile refusal. Additional challenges were imposed by the design of a very light and flexible jacket, including closely spaced piles, to permit crane installation. The jacket’s flexibility increased the potential for two-way cyclic leg loading and degradation of pile skin friction. The tight pile groups increased the scope for negative pile group interaction effects. 6.2 Design and assurance plan The Clair Project adopted a systematic ‘learn-as-yougo’ approach to foundation engineering and assurance © 2011 by Taylor & Francis Group, LLC

498

Figure 5. Process used for clair phase 1 design.

based on the observational method of ground engineering described by CIRIA (1999). The aim was to manage geotechnical uncertainties and ensure economic, reliable pile foundations by a continuous process of design, installation monitoring and review, from concept selection through to detailed design and installation (Fig. 5). A key part of the plan was that the MEI-FMMG foundation engineering and the IAT work were performed independently but in parallel. Other important elements were the adaptation of the jacket docking pile installation as a driving trial, the preparation of a Foundation Acceptance Plan (FAP) and the monitoring of the main jacket pile installations against the FAP. 6.3

Static pile capacities

Despite the long-standing wording in the API RP2A recommendations, it has been common in routine offshore axial pile capacity assessment to take no explicit account of strain softening, cyclic loading or loading rate effects, unless the (1) soils show unusually large post-peak reductions in laboratory strength tests, (2) the design loads includes unusually high levels of two-way cycling, or (3) if other special features apply. Given the unusual soils at Clair, it was considered appropriate to address these issues explicitly as part of foundation assurance. The designer MEI-FMMG and the IAT used a similar two-stage approach for assessing pile capacities that involved first the estimation of ‘Baseline’ static single pile capacities and then ‘Operational’ capacities by applying correction factors to the ‘Baseline’

capacities to account for factors such as group effects and cyclic loading. MEI-FMMG estimated ‘Baseline’ capacities using Randolph and Murphy (1985), from which the API main text method for clays (API, 1993) was derived, and concluded that the net effect of other factors was neutral, so the ‘Baseline’ capacities could be relied upon as ‘Operational’ values. The IAT derived ‘Baseline’ capacities in two ways. The first was to review the API pile test database to assess the trends that existed for bias between predicted and measured capacities for the very hard Clair soils. For the subset of pile tests that best matched the Clair soil conditions, a conservative bias of about 16% appeared to apply to the API-derived capacity for the design case of 2.59-m (120-in) diameter piles driven to 29 metres. However, none of the tests reviewed approached the apparent OCRs encountered at Clair. Extrapolating the database suggested an even greater (38%) capacity bias for the API capacity at Clair. The ‘Baseline’ capacity was therefore assessed as between 16% and 38% higher than given by the API main text method. In view of this apparent conservatism, the IAT also adopted a first-principles ‘best-endeavours’ approach to pile design, taking account of recent research. They calculated ‘Baseline’ capacities by using the effective stress method developed at Imperial College, London (Jardine and Chow, 1996). A capacity 10% higher than the API main text method was found if the Kc values were limited to the maximum ratio (8) seen in field tests with the ICP instrumented pile, while a 44% margin was found without this limit being imposed. Both approaches therefore resulted in average ‘Baseline’ capacities about 27% higher than main text API. However, the IAT also considered the factors that might adversely affect operational capacities. Sitespecific analyses indicated that cyclic loading and group interaction would have the largest effects and that the operational capacities could be reduced about 20% by the combination of these effects. The overall operational capacities calculated by the IAT were on average only about 7% higher than given by API main text. All these factors led to a broad agreement between the ‘Operational’ capacities derived by the MEI-FMMG code-based and IAT ‘best-endeavours’ approaches, and the project therefore took the MEIFMMG pile design forward to the next step in the design assurance process; a driving trial. 6.4

Pushover analyses

As a special part of the verification scheme, Lloyd’s Register performed independent pushover analyses of the Clair jacket, simultaneously modelling the jacket structure and entire foundation. The hybrid foundation model used for this task took weighted account of the various pile design approaches considered by MEI-FMMG and the IAT teams, backed-up by observations made during pile driving. One key aspect of this model was that special attention had been placed © 2011 by Taylor & Francis Group, LLC

on phased development of friction capacity prior to full mobilisation of end-bearing capacity. These analyses demonstrated that although some individual piles and pile groups displayed capacities less than code requirements, using the results of in-place structural analyses, the entire foundation system displayed an ultimate reserve strength ratio that exceeded requirements for approval. Moreover, it was found that foundation capacity did not govern the ultimate capacity of the substructure.These findings were of key importance to the verification of the BP Clair platform foundations. 6.5

Pile driving trial

Uncertainties about pile installation were addressed by adapting the jacket docking pile installation as a pile driving trial. Two 1.829-m (72 in) diameter by 75 mm wall thickness piles were driven at the Clair site in the summer of 2003 under FMMG’s supervision, one year before the main jacket installation. One of the piles had an internal driving shoe and both were instrumented with strain gauges and accelerometers by Conewel b.v. Each pile was driven continuously to a target depth of about 29 m from Saipem’s S7000 vessel using an MHU 3000 underwater hammer. The shoed and unshoed piles were re-driven 14.5 hours and 22 hours after initial driving, respectively. No boulders were encountered during the test drives. Both piles drove much more easily than expected from conservative forecasts, but generally in line with predictions based on North Sea hard clay site experience. Soil plug behaviour was not monitored directly but signal matching indicated that the shoed pile drove unplugged and marginally more easily than the other pile, which was inferred to be unplugged to about 18-m penetration and to be partially plugged below that depth. Dynamic signal-matching and measured blowcounts were used to infer soil set-up and static pile capacities. Static shaft friction was estimated to have increased by about 35% after 14.5 hours and about 45% after 22 hours. The ultimate static compressive capacity of the uniform wall-thickness un-shoed pile was predicted to develop about 60 days after driving and to be about 55% higher than the static capacity at the end of continuous driving. The ultimate long term static capacity was estimated to be about 34% higher than the ‘Baseline’static capacity derived by MEI-FMMG using the modified API main text method, and was also in good agreement with the capacity calculated using the two IAT approaches, which averaged 27% above the capacity given by main text API. The results of the driving trials were used to (1) confirm MEI-FMMG’s reference design, (2) finalise risk reduction measures and (3) develop a Foundation Acceptance Plan (FAP) for the jacket pile installation. 6.6

Risk reduction

A range of measures was considered by the Project to mitigate pile damage, premature pile refusal and/or

499

considering the tip annulus and plug. As described earlier, these improved estimates of pile response were adopted into the pushover analyses. It should be noted that the detailed pile driving monitoring was considered essential to the verification scheme, both prior to installation and also retrospectively. The foundations were accepted by the project and IVB without any re-driving or remediation.

unexpectedly low static capacities. These included overdriving piles, spare piles, pre-drilled pilot holes, internal driving shoes, a thickened pile driving head, strengthened pile tip and a back-up drill rig for intervention. These options were investigated through the design and assurance process and some were shown to be unsuitable or unnecessary. Extra-long piles were inefficient and increased pile driving risks unduly; additional pile sleeves were precluded by jacket weight restrictions. However, the confidence gained from the pile driving trial allowed the Project to drop plans for pilot holes and internal driving shoes, both of which were expected to reduce pile capacities. However, boulders could not be excluded, so the options of thickened driving heads, toughened pile tip and a back-up drill rig were taken through detailed design into installation. 6.7

Pile installation and acceptance

No significant obstruction was encountered and all piles drove as expected. Set-up observed following operational delays confirmed average long term skin friction increases of at least 50%. Lloyds Register considered the external shaft frictions deduced from the driving data, including the set-up deduced from extrapolation of the (un-planned) re-strike tests. Peak frictions were based on the set-up friction, and residuals were based on frictions during driving. In addition, it was possible to revise estimates of end bearing,

© 2011 by Taylor & Francis Group, LLC

FUTURE INSTALLATIONS

Soil conditions at the Clair Phase 1 platform location comprise dominantly hard glacial tills whose undrained shear strengths are unprecedentedly high for an offshore driven steel pipe pile installation. A systematic observational approach was adopted to cope with uncertainties in foundation design and installation performance. The risks were managed by conducting high quality site investigations, analyses and trials, so that expectations could be updated against the monitored pile behaviour. The methodology worked well and, together with lessons learnt, is currently being applied for the second phase of development at Clair Ridge.

Foundation acceptance plan

The Project developed a Foundation Acceptance Plan (FAP) based on the observed and extrapolated behaviour of the docking piles. Pile acceptance was predicated on piles driving within the expected range inferred from the driving trial and external skin friction increasing by at least 50% after the end of driving. Each of the fourteen docking piles was fully instrumented and individual piles were accepted if they drove to the target penetration of 29 m with acceptable blowcounts, or were driven to refusal to a depth of least 27 m. No re-strikes were proposed other than for piles that drove to 29 m with lower blowcounts than expected. The FAP also called for a drilling rig to be on standby in case piles hung up prematurely, or were deliberately terminated to avoid pile damage. The intervention plan was to drill out the soil plugs, redrive to at least 27 m and form an internal grout plug to ensure plugged behaviour under design conditions. Provision was also made for accepting deficient piles based on satisfactory pile group response. 6.8

7

ACKNOWLEDGEMENTS The authors are grateful to BP North Sea and its co-venturers, ConocoPhillips, Chevron, Shell and Hess for permission to publish this paper and share the unique Clair Phase 1 foundation design and installation experiences with others.

REFERENCES Aldridge, T.R., Carrington, T.M., Jardine R.J., Little, R., Evans, T.G. and Finnie, I. BP Clair Phase 1 (2010), ‘BP Clair Phase 1 – Pile driveability and capacity in extremely hard till’, Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. API (1993), ‘Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – working stress design’, API recommended Practice RP-2A-WSD, twentieth edition, July 1993. CIRIA (1999), ‘The Observational Method in ground engineering: principles and applications’, Construction Industry Research and Information Association, Report 185, London. Jardine, R.J., Chow F. C. (1996), ‘New methods for Offshore Piles’, Publication 96/103, The Marine Technology Directorate.

500

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Pile driving experiences in Persian Gulf calcareous sands K. Fakharian & I.H. Attar Department of Civil & Environmental Engineering – Amirkabir University of Technology, Tehran, Iran

ABSTRACT: This paper presents the results of driven pile skin friction setup experiences in Persian Gulf calcareous sands. Dynamic load tests on steel pipe piles of the “New LPG & Condensate Loading Facilities” project, Sirri Island, Iran, were performed both at End Of Initial Drive (EOID) and at restrike conditions. The construction piles were 26

and 32

open-end pipe piles with total lengths in the range of 30 to 82 m. The embedment depths are in the range of 15 to 45 m. Signal-matching analyses were performed for the End Of Drive (EOD), Beginning Of Restrike (BOR) and/or End Of Restrike (EOR), to verify the pile capacity and compute the shaft resistance distribution. The results of the analyses show that the increase in pile capacity is mostly attributed to the shaft resistance in the study area. Two procedures were adopted to compensate for the non-mobilized shaft friction resistance of the lower parts of the piles, due to supposedly insufficient hammer energy. The results show that the setup factor has increased from 2.44 to 3.45 through 4.21. The restrike tests on one of the piles continued to more than 60 blows to verify the EOR resistance. It is noticed that the shaft resistance has reduced to half which is attributed to probable loss of resistance due to succession of blows. The toe resistance shows no significant reduction, however. 1

INTRODUCTION

Despite the fact that pile design approaches have advanced enormously within the past few decades, the most fundamental aspect of pile design, that of estimating the axial bearing capacity, relies heavily on empirical correlations, as discussed by Randolph (2003). Many studies have been reported within the past 40 years to improve the pile bearing capacity estimated with dynamic methods during pile driving and testing (e.g., Goble et al., 1967; Goble et al., 1970; Rausche et al., 1972). The results have shown that the variations of bearing capacity with time after the pile initial driving depend on different parameters of the pile and the soil. In the majority of cases, so-called pile setup has been reported with long-term increases of bearing capacity from 50 through 1000 percent of the End-of-Initial-Drive (EOID) resistance (Rausche et al., 2004). Occasionally, soil relaxation has been reported, in which the bearing capacity decreases slightly with time to below EOID. The studies have indicated that the long-term setup for shaft friction of piles embedded in non-cohesive soils is high (Axelsson and Hintze, 2000). The reported cases by Chow et al. (1998) also reveal that the shaft resistance has doubled over one to 100 days. Camp and Pamar (1999) reported setup ratios, BOR/EOD of up to 8, for overconsolidated hard Cooer Marl. Rausche et al. (2004) attributed the disturbance of the cemented structure of calcareous sediments to the set-up with time. It is not always practicable to evaluate the real soil setup from field measurements by comparing the EOID and restrike tests. This is in particular the case © 2011 by Taylor & Francis Group, LLC

once the setup is very high and therefore, requires longer time to develop. The selected hammer for pile driving is usually on the basis of the pile dimensions (length and diameter) and the maximum expected capacity during driving. On the other hand, the full mobilization of the pile bearing capacity in a restrike test may require a heavier hammer which may not necessarily be available at the construction site. Insufficient energy of the driving equipment may result in partial mobilization of the tip resistance as well as the lower elevations of the shaft resistance of the pile. In this study, pile dynamic test results are evaluated for long steel pipe piles, driven through the sandy layers of Sirri Island near-shore LPG export jetties. The acquired signals have shown that despite a considerable increase in mobilized resistance of about 2 times between EOID and restrike tests, only the upper parts of the shaft resistance is fully mobilized. To compensate for the non-mobilized resistance of the lower part, to estimate the real bearing capacity, two procedures are followed. In the first procedure, the restrike test results for lower parts of the pile are corrected with attention to the trend of EOID shaft resistance distribution. The shaft resistance has been corrected on the basis of SPT results in the second procedure. The estimated corrected capacities are compared and discussed. 2

SITE CHARACTERISTICS

The test site is in the south east of Sirri Island, on the Persian Gulf. Sirri Island is located between latitudes 25–56
and 25–53
N and longitudes 54–33
and

501

Figure 2. SPT N-value variations with depth for 6 boreholes of adjacent structures.

down to about 11 m, below which the N-values are increased to the range of 50 through 90 or in fact refusal. An abrupt increase of the N-value to 90 for BH6 at depth 12 m might be attributed to the cemented sub-layers described above. The pile driving records and the PDA test results confirm the frictional nature of the soil down to the embedment depths of 30 m in Dolphin structure.

Figure 1. The construction site and aerial photo.

54–29
E and is one of the six islands in Abu Mus â Island Group (Hormozg ân Province) in Iran, about 80 km south east of Kish Island and 50 km west of Abu Mus â Island. This island covers an area of 17.3 sq kms. The highest point on the island is 33 m above the sea level. Like the other islands in the Persian Gulf, Sirri Island enjoys a warm and humid climate. An aerial photo and a view of the construction site are shown in Fig. 1. A total of 27 open-end steel pipe piles were designed to support the structures of the “New LPG & Condensate Loading Facilities”, having three main components including “Piperack”, “Barrier” and “Dolphin”. The construction site is adjacent to a similar oil export loading facility which also includes all the three structural components. Two to three days of loading per week for the oil export facility interfered with the construction and pile driving activities and testing. The sea water depth across the pile driving zone is within 9 to 28 m. The foundation soils are defined in the drilled boreholes as dense to very dense calcareous sands with locally weakly cemented to cemented calcareous coarse sand to sandstone layers, which apparently do not exceed 1 m in thickness. During borehole drilling, SPT tests were conducted at 3-m intervals down to 15 m of depth. Figure 2 shows the variations of SPT N-value with depth for all the 6 boreholes. As shown, the N-values are between 40 to 70 © 2011 by Taylor & Francis Group, LLC

3

PILE INFORMATION

The pile diameters are of two pipe types, either 26

× 0.75

or 32

× 0.875

. All the barrier piles were driven vertically, but rest of the piles in the Dolphin and the Piperack were batter piles with slopes of 1/3.5 and 1/5.0, respectively. All the piles were driven openend without any shoe at the toe. The total pile lengths are in the range of 30 to 82 m. The construction method was pile driving with seato-sea operations, using a barge to support the hammer crane, vibro-hammer and the diesel hammer. Depending on the total pile length and the water depth, the shorter piles were fabricated to the full length on the barge by weld splicing before being lifted for driving. The longer piles were fabricated up to 42 and 48 m on the barge and driven. Then the last segment was lifted and spliced on top of the fabricated segment and re-driven. The skin surfaces of the upper part of all the piles were coated to prevent corrosion within the splash zone and above it. The pile was pushed by an ICE-416-L vibrohammer to at least half of the total embedment depth. Thereafter, a D-62 open ended diesel hammer (rated

502

© 2011 by Taylor & Francis Group, LLC

503

437 900 750 542 513 369 850 827 1,501 2,706 1,750 1,415 3,457 1,582 4,503 4,153 1,938 2,606 2,500 1,956 3,970 1,950 5,353 4,980 1,937 2,424 2,712 1,895 5,218 2,349 5,080 4,918 36.0 44.5 50.4 43.4 59.4 47.4 63.5 52.6 EOD BOR EOR Embed.

End of Driving Beginning of Restrike End of Restrike Length below adjacent grade at the time of testing

EMX Maximum energy transferred to the pile head PRES Penetration resistance (Blows per 25 mm) RX#/RS# RMX/RSP CASE Method with a J-Factor of #

44.0 43.0 45.0 46.0 45.0 45.0 43.0 43.0 8.8 5.0 5.5 3.8 40.0 4.0 35.0 58.0 22.50 30.00 30.00 15.00 15.00 15.20 15.00 15.00 1257 534 215 774 4 61 4 4 13/03/08 27/06/08 28/06/08 12/03/08 29/06/08 29/06/08 29/06/08 29/06/08 Project: New LPG & Condensate Loading Facilities, Sirri Island, Iran P-1 Pipe 26

× 0.75

, Bat. 35 EOD 13/03/08 P-11 Pipe 32

× 0.875

, Bat. 20 EOD 27/06/08 11 EOD 28/06/08 P-13 Pipe 32

× 0.875

, Ver. P-22 Pipe 26

× 0.75

, Ver. 20 EOD 12/03/08 BOR 12/03/08 EOR 12/03/08 P-25 Pipe 26

× 0.75

, Ver. 28 BOR 31/03/08 P-27 Pipe 26

× 0.75

, Ver. 49 BOR 31/03/08

EMX kJ Rate Bl./min. PRES Bl./25 mm Embed. m Blow No. Date tested Date Driven Test (E / R) PRES/EOD Bl./10 cm Pile type & size Pile No.

Total of 8 PDA tests were carried out on 6 piles. The tested piles were driven to a final embedment depth of between 15 m and 30 m, using the D-62 hammer. The penetration at the end of drive was as low as 11 Bl/10 cm for pile P-13 and as high as 49 Bl/10 cm for pile P-27. The hammer stroke varied between 2 m and 2.3 m at EOD, corresponding to blow rates of 43 to 45 Blows Per Minute (BPM). The details of PDA tests and CAPWAP analyses are presented in Table 1. Signal-matching analyses using CAPWAP were performed on selected representative blow records from EOD, BOR and EOR. The measured resistance of different test piles ranged from 1,900 kN to 2,500 kN, for embedment depths of 15 to 30 m. The measured resistance at BOR of different piles ranged from 3,900 kN to 5,300 kN, for embedment depths of 15 m. Therefore, a setup of about 2 (BOR/EOD) is observed with time. Signal-matching analyses on all the reported blows in Table 1 show that the toe resistance is in the range of 400 kN to 900 kN, which is far less than the friction or shaft resistances in the range of 1,400 kN through 4,500 kN. The shaft to toe ratios are in the range of 3.5 through 5 which is attributed to the fact that the piles are open-toe and the observations showed that no plug has formed and soil penetrates into the pipe pile during driving. The shaft frictions are hence higher as a result of outside and inside friction on the pile shaft. The soil penetration inside the pipe was measured as 14.5 m (out of 15 m embedment depth) for pile P-22 and 18.6 m (out of 22.5 m embedment depth) for pile P-1. The restrike test on pile P-22 continued to more than 60 blows to verify the EOR resistance which is lower than the BOR resistance to probable loss of resistance due to the succession of blows. It is noticed that the shaft resistance has reduced to half, which is down to around 2,000 kN from 4,000 kN. The toe resistance has also reduced from 510 kN to 370 kN. The substantial reduction in shaft resistance is probably attributed to increase in pore water pressure around the shaft. It could be also resulted from disturbance of the cemented structure of calcareous sediments, as pointed out by Rausche et al. (2004). Therefore, reduction factors may be applied to the pile ultimate resistance for post-cyclic responses like those after an earthquake. The reduction is shaft resistance in piles P-25 and P-27 after a succession of blows was not as severe as in pile P-22 though. The results of piles P-25 and P-27 show some unusual pattern of force and velocity signals at 0.85 of pile length below the installed gages, that is at about 5 m to the pile toe. This could be attributed to two factors: (1) damage at this elevation, or (2) lack of mobilization of shaft and toe resistance beyond this

RMX CASE Estimate (kN)

DYNAMIC TEST RESULTS

Table 1. PDA Data Table and CapWap Analyses.

4

Computed Resisitance (kN) Total Shaft Toe

energy of 223.3 kJ) was used to drive the pile to the specified embedment depth. The blow counts were recorded for all the piles during hammer driving.

EOD. The signals from PDA tests also show that the lower elevations of shaft resistance had not been fully mobilized. This seems to be attributed to the fact that the hammer energy has not been sufficiently high. Two procedures are introduced in the following subsections 5.1 and 5.2, to estimate the full shaft resistance of the pile. In the first procedure, it is assumed that the entire shaft resistance has been mobilized during EOD test. The shaft resistance distribution at EOD is calculated through signal matching analysis. The trend of variations is then applied to the BOR results. In the second procedure, the partially mobilized shaft friction has been corrected using the SPT-N values. 5.1

Figure 3. Shaft resistance distribution along pile 22 at EOD, BOR, and EOD (CAPWAP).

point due to very high soil setup. Comparison with results of P-22 during which the setup was completely reversed is in support of the 2nd postulate, as the pile is less likely to have experienced severe damage at this elevation. Had the driving continued at the restrike test for a few more minutes, it would most likely mobilize the rest of the shaft resistance beyond this elevation and a clear toe reflection would have been observed. Figure 3 shows the shaft friction distribution of P-22 for the three conditions of EOD, BOR and EOR, resulting from signal-matching analysis. Comparing the BOR and EOD results shows a setup ratio of 2.44, i.e. the BOR/EOD (3,460/1,420). The toe resistance on the contrary, has reduced from 540 kN at EOD to 510 kN at BOR. This is most likely attributed to the insufficient hammer energy to mobilize the ultimate toe capacity. The considerable reduction of shaft resistance at EOR after about 60 blows is discussed in subsequent sections. 5

CORRECTED SOIL SETUP MEASUREMENTS

It was explained in the previous section that the tip resistance of P-22 had reduced at BOR compared to © 2011 by Taylor & Francis Group, LLC

Shaft set-up estimate from restrike tests

The comparison between EOD and BOR results for P-22 in Fig. 3 shows that during EOD tests, the shaft resistance has linearly increased down to about 10 or 11 m of embedment depth, below which it has decreased. For BOR, however, the shaft resistance has linearly increased down to 7 m depth only, and then reduced below it. Signal-matching analysis has been carried out to calculate the shaft resistance distribution along the pile shaft. Supposing that the reduction in shaft resistance between 10 to 15 m at EOD is due to decrease in soil resistance in that zone, and not insufficient energy of the hammer, then it can be assumed that the same trend applies to the BOR test results. This implies that in spite of shaft resistance reduction between 10 to 15 m related to soil resistance reduction, because of insufficient hammer energy the resistance has further decreased between 7 to 10 m, for example for P-22. To compensate for this reduction, the trend of variations of shaft resistance for EOD is extended to BOR between 7 to 10 m. The corrected result is presented in Fig. 4, together with non-corrected results of BOR and the EOD variations. Calculation of the total shaft resistance has shown that there is an increase from 3,457 kN in noncorrected results up to 4,931 kN in the corrected ones. The comparison shows that in fact 43% of the pile capacity has not been mobilized during the BOR test. The setup ratio is therefore increased from 2.44 (BOR/EOD non-corrected) to 3.48 (BOR/EOD corrected). 5.2

Shaft set-up estimate from SPT N-value

As SPT results with depth are available from the adjacent site (Fig. 2), it would be possible to calculate the shaft frictional resistance distribution from the socalled direct methods on the basis of SPT results. There are many methods available in literature to calculate tip and shaft resistances on the basis of N-value from SPT. Fakharian & Vaezian (2007) compared 12 of the well-known methods for different conditions of pile installation method (bored and driven) and soil type (sand and clay). A database was collected for welldefined site conditions for which soil layering, SPT

504

Figure 4. Correction of BOR shaft resistance distribution from EOD trend for pile 22 (1st procedure).

Figure 5. Correction for BOR shaft resistance distribution from EOD trend and SPT N-value for pile 22 (2nd procedure).

profile, static and dynamic tests (with distinguished tip and shaft resistances) were available. A ρ parameter was defined as the ratio calculated to measured bearing capacity. The average and standard deviation were calculated for each category of data. It was found that the Aoki and Vellosos (1975) method results in the best estimate for driven piles in non-cohesive layers with an average and standard deviation of 1.05 and 0.14 for ρ, respectively. The method is described in appendix A. One important aspect for utilizing direct methods to determine the pile bearing capacity is the proper estimation of the necessary parameters. As the upper 7 m of the pile shows full mobilization of the frictional resistance, the parameters are back-calculated from the signal-matching analysis in this zone and then applied for the shaft resistance calculations below 7 m. The method was utilized for p-22 and resulted in a shaft resistance of 5,946 kN. The shaft friction distribution from SPT and its comparison with EOD, BOR and corrected BOR are presented in Fig. 5. The friction distribution resulted from SPT is overall higher than BOR and corrected BOR, but the trend of variations compiles with the BOR results. It should be noted that SPT results are collected from an adjacent site and

there might be some variations with the location of the driven pile. In this method, the setup factor has increased to 4.2, as compared to 2.44 for EOD and BOR factors and 3.48 for EOD and corrected BOR factors. Of course, an instrumented pile statically loaded up to the ultimate capacity would have been required to fine-tune the shaft frictional resistance distribution and measure the tip resistance, but unfortunately, it was not accepted by the client to carry this out due to time and budget restrictions of the project. One of the initial concerns of the design consultant and the client was the possibility of soil relaxation, instead of soil setup, as the deposits at the construction site are constituted of calcareous sediments. Since such granular materials are known for their degradation potential of frictional resistance once subjected to axial cyclic loading, for example resulted from an earthquake, it was not really clear before construction whether setup would occur. The EOD, BOR and EOR dynamic test results revealed that the setup is considerably high, but the EOR results proved that degradation of shaft resistance is highly probable and this point has to be anticipated in selecting the allowable capacity for different loading combinations.

© 2011 by Taylor & Francis Group, LLC

505

6

SUMMARY AND CONCLUSIONS

Chow, F.C., Jardine, R.J., Brucy, F., and Nauroy, J.F. 1998. Effects of Time on Capacity of Pipe Piles in Dense Marine Sand. Journal of Geotechnical and Geoenvironmental Engineering, Vol. 124, No. 3, ASCE, pp. 254–264. Fakharian, K. and Eslami A., 2006. Axial Bearing Capacity of Piles, Transportation Research Institute, Ministry of Transportation, Tehran, Iran. Fakharian, K. and Vaezian, H., 2007. Applicability of SPT results to bearing capacity determination of piles, 13th Asian Regional Conference on Soil Mechanics and Geotechnical Engineering, December 10–14, 2007, Kolkata, India. Goble, G.G., and Rausche, F., 1970. Pile load test by impact driving, Highway Research Record 333, 1970, 123–129. Goble, G.G., Scanlan, R.H., and Tomko, J.J., 1967. Dynamic studies on the bearing capacity of piles, Highway Research Record 167, 1967, 46–47. Randolph, M.F. 2003. Science and Empiricism in Pile Foundation Design. 43rd Rankine Lecture, Geotechnique, 54(1). Rausche, F., Moses, F., and Goble, G.G., 1972. Soil resistance predictions from pile dynamics. ASCE, Journal of Soil Mechanics and Foundations, SM9, September, 917–937. Rausche, F., Robinson, B., Likins, G. E., August, 2004. On the Prediction of Long Term Pile Capacity From End-ofDriving Information. Current Practices and Future Trends in Deep Foundations, Geotechnical Special Publication No. 125, DiMaggio, J. A., and Hussein, M. H., Eds, American Society of Civil Engineers: Reston, VA; 77–95.

The results of dynamic tests on open-end steel pipe piles driven at Sirri Island in Persian Gulf were presented. The setup factor for shaft resistance was evaluated. Two procedures were followed to compensate for the non-mobilized shaft resistance at the lower elevation of the pile. The main conclusions of the presented study are summarized below: 1. A very good setup for shaft resistance is observed in the calcareous sands of the region at the beginning of restrike. 2. The setup might be the result of dissipation of excessive pore pressure generated during initial drive or from disturbance of the cemented structure of calcareous sediments. This requires further evaluation or field monitoring during driving, to clarify. 3. The setup factor seems to be even higher than what is directly calculated from BOR/EOD shaft resistance. 4. The two procedures used to correct the BOR shaft resistances are compatible in trend and suggest that the setup factor could be even above 4. 5. The loss of the increased shaft resistance after about 60 blows during the restrike tests suggest that the degradation potential when subjected to cyclic loads is extremely high which is an important design consideration.

ACKNOWLEDGMENTS The authors would like to thank IOOC, the project owner, OIEC, the EPC contractor, Fars Scout Industrial Co., the sub-contractor, and Pars GeoEnviro Co., the pile dynamic testing consultant, for providing the necessary data for conducting this study.

APPENDIX A SPT in situ based direct methods are available in literature through which the tip and/or shaft resistance of the pile are calculated directly on the basis of N value. These methods calculate the ultimate capacity according to following Equation:

REFERENCES

Aoki and Vellosos method (Aoki and Vellosos, 1975) is used for bored and driven piles and clay, silt and sand soils, the rt and rs in this method are calculated as following:

Aoki, N. and Velloso, D.A. 1975. An approximate method to estimate the bearing capacity of piles, Proc. 5th Pan American Conference on Soil Mechanic and Foundation Engineering , Buenos Aires, 1, 367–375. Axelsson, G. And Hintze, S. 2000. Evaluation of pile setup from penetration per blow, Proc. 6th Int. Conf. on the Application of Stress-Wave Theory to Piles, Sa˜o Paulo, Brazil, pp. 665–672. Camp III, W.M., and Parmar, H.S. 1999. Characterization of Pile Capacity with Time in the Cooper Marl: A Study of theApplicability of a PastApproachTo Predict Long-Term Pile Capacity, Emre, TRB, pp. 1–19.

in which: Nb = Average of SPT N -value near the pile tip Nsi = SPT N -value in subjected layer For α, K, F1 , F2 See Fakharian & Eslami (2006).

© 2011 by Taylor & Francis Group, LLC

506

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

FLAC3D analysis on soil moving through piles E.H. Ghee & W.D. Guo Griffith University, School of Engineering, Gold Coast, Australia

ABSTRACT: Piles may be utilised as deep foundations or employed to support offshore structures, which may be subjected to passive loading due to lateral soil movements. The safety of the piled foundations depends on the additional stresses induced. This issue has been investigated recently by conducting model tests on single piles and on pile groups with a new apparatus developed by the authors. Typical test results were analysed and reported previously. In this paper three-dimensional finite difference analyses are reported: (1) to predict the results of two model tests (with and without axial load); and (2) to investigate the effect of the moving and the stable depths of soil on the pile response. Typical results of the comparison between the FLAC3D analysis and the model tests for single piles in sand are presented in terms of three profiles namely: bending moment; shear force; and pile deflection profiles. A unique linear relationship between the maximum shear force (thrust) and the maximum bending moment induced on the piles is obtained regardless of the ratios of the moving depth over the stable depth.

1

INTRODUCTION

2

Pile foundations designed to support offshore structures and services are often subjected to lateral soil movements and axial load simultaneously. There have been active studies on piles subjected to vertical load, and on piles subjected to lateral soil movement. However, little information is available for evaluating the response of vertically loaded piles due to soil movement. This response is important, in particular, for offshore foundations, API (2000) specifies that possibility of soil movement against foundations should be investigated and the forces caused by such movements, if anticipated should be considered in the design. Studies on piles due to liquefaction induced lateral soil movement have found that the influence of axial load can cause on the piles (1) additional bending moment; (2) additional compression stress; and (3) additional lateral displacement (Bhattacharya 2003). The first two findings are consistent with those observed from the model tests conducted by the author (Guo & Ghee 2004), however, lateral displacement of a free-head pile is found to have reduced rather than increased. With the support of the Australian Research Council, a new apparatus was developed to simulate axial load and lateral soil movement on the pile. The details of the apparatus and typical test results were previously reported in Guo & Ghee (2004) and Guo et al. (2006). In this paper three-dimensional finite difference (FLAC3D ) analyses were conducted: (1) to predict the response of two model piles (with and without axial load); and (2) to investigate the effect of the moving and the stable depths of soil on the pile response. © 2011 by Taylor & Francis Group, LLC

FLAC3D ANALYSIS

FLAC3D version 2.1 (Fast Lagrangian Analysis of Continua in 3 Dimensions) was used to perform the numerical analysis. One of the main features of the FLAC3D is that it can operate in small or large strain mode. The large strain mode occurred when the grid point (mesh) coordinates are updated at each strain or movement increment, according to the computed displacement. This is particularly important to the present study involving large soil movements. In order to model piles subjected to lateral soil movements, a lateral velocity (defined in FLAC3D as a unit of movement per step or iteration, mm/step) is applied in the direction parallel to x-axis as shown in Figure 1c. Experience gained from calibrating the model pile indicates that a minimum of 150,000 steps is required for the unbalanced force to arrive at a small value, and also remain constant without significant changes. Therefore, the velocity was applied over two intervals: firstly by an velocity of 10−6 mm/step to bring the pile and soil into equilibrium at 30 millions steps which is equivalent to 30 mm of lateral soil movement; subsequently using a 10 times higher velocity in order to minimise the accumulation of truncation errors that arise when very small displacement increments are added to generate coordinate values in large strain mode.

507

2.1

Geometry and constitutive model

In order to capture the effect of soil arching, and soil flowing on response of passive piles, a three-dimensional model was deemed necessary. The

Figure 2. Physical model test setup.

zero cohesion and dilation; and 4) zero tensile strength. The interface elements (with limiting tensile stress set to zero) allowed the soil to slip and/or seperate from the pile. In most analyses, slip between the pile and the soil occurred due to the relative movements, most notably at the surface of the soil.

Figure 1. Mesh of soil and pile used in FLAC3D analysis.

symmetrical nature of the model test allows using only half of the actual pile-soil system to model overall pile response, as shown in Figure 1. The soil strata were modelled with eight-noded brick shape elements and the pile using six-noded cylindrical shape elements. Interface elements were placed between the soil and the pile. For the standard model, with a single pile, the mesh comprised of 1856 elements with 2894 nodes or gridpoints. With the current computer capacity (Intel® Core™2 Duo), the computation took approximately one day for a single analysis. It was found that should a higher number of elements been used, only a small increase in accuracy would be achieved, and would require longer computing time. The bottom face of the mesh in Figure 1b was fixed in all three directions x, y and z. Both faces parallel to the yz-plane were fixed in the x direction and the other faces parallel to the xz-plane were fixed in the y direction. The surface of the mesh is not fixed in any direction. The sand strata were modelled with an elasto-plastic Mohr-Coulomb model and using a non-associated flow rule. The Mohr-Coulomb criterion does affect the behavior in the moving soil layer at large sand movements. Interface elements were attached to the outer perimeter of the pile shaft, hence separating the pile and the adjacent soil. These interface elements were defined by the Coulomb failure criterion with the following parameters: 1) friction angle of 28◦ ; 2) normal stiffness and shear stiffness of 1.0 × 108 N/m2 /m; 3) © 2011 by Taylor & Francis Group, LLC

2.2

Soil properties

The sand properties used are taken directly from that reported in the model tests (Guo & Ghee 2004). In both the moving (Lm ) and the stable (Ls ) layers (see Fig. 1c), the sand was assigned the following: • A friction angle of φ = 38◦ (peak angle obtained

• •

•

•

from the direct shear box test). This angle was set to remain constant in the Mohr-Coulomb model. A dilation angle estimated using the equation (after Chae et al. 2004): ψ = φ-30◦ . A Young’s modulus Es of 572 kPa obtained from oedometer test data (assuming Poisson’s ratio of 0.3) for a vertical stress of 6.5 kPa over the middle depth of the sand in the shear box. An initial stress calculated by specifying a density of 16.27 kN/m3 and by a coefficient of earth pressure at rest, Ko (= 1 − sin φ, see Jaky 1944) estimated using the friction angle. zero tensile strength.

2.3

Pile model

The maximum bending moment in the pile in this study was expected to be much less than the yield moment of the pile (Myield = 1,340 Nm). Therefore, the pile was modelled as an isotropic elastic hollow pile that consisted of cylindrical elements (Fig. 1a). The mesh of the FLAC3D model (Fig. 1) has been setup to suit the dimensions of the experimental apparatus (Fig. 2), and for the two model tests detailed in

508

Table 1.

Details of FLAC3D analyses.

Axial load Pile diameter Moving layer, Stable layer, Test (N) (mm) Lm (mm) Ls (mm) 1 2

0 284

50

400

300

Table 1. For comparison purposes, the results obtained from the FLAC3D analyses and the model tests will be presented together, in this paper. As a calibration, a FLAC3D analysis was first conducted on a cantilever pile, i.e. the pile was first fixed at one end into ground, and without soil around. The length of the pile was 1.2 m and the internal and the external diameters were 50 mm and 48 mm respectively. The aluminium pile had Young’s modulus of 7.0 × 1010 N/m2 and Poisson’s ratio of 0.3. Note these are the actual dimensions and properties of the pile used in model tests. A lateral load of 300 N was applied on the free end of the cantilever pile. The obtained deflection from FLAC3D shows a maximum 5 % difference from analytical results. Thereby, the FLAC3D analysis is sufficiently accurate. 3

PREDICTION OF MODEL TESTS

In the first stage, this paper will simulate two model tests reported by Guo & Ghee (2004) and Guo et al. (2006). The shear apparatus used to conduct the model tests is shown in Figure 2. It is mainly made up of a shear box, a loading system, and a data acquisition system. • The shear box is of 1 m (width) × 1 m (length) ×

0.8 m (height). The upper section of the shear box consists of 25 mm deep square laminar steel frames. The frames, which are allowed to slide, contain the “moving layer of soil” of thickness Lm . The lower section of the shear box comprises a 400 mm height fixed timber box and the desired number of laminar steel frames that are fixed, so that a “stable layer of soil” of thickness Ls ( ≥400 mm) can be guaranteed. Changing the number of movable frames in the upper section, the thicknesses of the stable and moving layers are varied accordingly. Note that the Lm and Ls are defined at the loading location, and they do vary across the shear box. The actual sliding depth Lm around a test pile is unknown, but it would not affect the conclusions to be drawn. • The loading system encompasses a loading block that is placed on the upper movable laminar frames, and some weights on top of the test pile. The loading block is made to different shapes in order to generate various soil movement profiles. A uniform loading block is shown in the figure that enforces a prespecified sliding depth of Lm . A hydraulic jack is used to drive the loading block. The surcharge is exerted by the weights through the loading plates. © 2011 by Taylor & Francis Group, LLC

509

Figure 3. Response of 50 mm pile Lm /Ls = 400/300 (Q = 0 N).

Response of the pile is monitored via strain gauges, and via the two linear variable displacement transducers (LVDTs) above the model ground.The test readings are recorded and processed via a data acquisition system and a computer. 3.1 Test 1 at Lm /Ls = 400/300 and Q = 0 N For Test 1, Figure 3 presents the pile responses when subjected to soil movement ws , of 60 mm. For comparison purposes ‘three’ profiles namely the bending moment, shear force and pile deflection obtained from both the FLAC3D analysis and the test data are presented. A spreadsheet program was written to analyse the displacement data (output from FLAC3D ) obtained at every 100 mm interval along the pile. The bending

moment profile along the pile was first derived from the 2nd order numerical differentiation (finite difference method) of the pile deflection profile. Subsequently, the shear force profile was derived from the 1st order differentiation of the bending moment profile. Figure 3a shows 104% difference in the maximum bending moment (located in the stable layer) between FLAC3D and test data, although the bending moment profiles show similar shape with double curvatures (negative and positive bending moments). Generally the FLAC3D analysis underestimated the shear force and the pile deflection. The pile deflection at the soil surface is much less than measured value of 79.7 mm. The deflection profiles indicate that the pile deformed like a rigid pile with a rotation point at the depth in between 600 mm and 700 mm. 3.2 Test 2 at Lm /Ls = 400/300 and Q = 284 N Test 2 was performed under identical conditions to Test 1, but with an axial load of 284 N applied on top of the pile at 500 mm above the soil surface (Fig. 1c). Figure 4 presents the three profiles of the pile responses subjected to soil movement, ws of 60 mm, which were obtained from both the FLAC3D analysis and model test, respectively. A similar shape of bending moment profiles is noted again. The maximum bending moment (Mmax ) obtained from FLAC3D analysis is 49% higher as compared to that obtained from the model test. This similarity is also noted in the maximum shear force (Smax ) and the deflection profiles (Figs. 4b, c). The deflection obtained from FLAC3D analysis at the soil surface is 61.1 mm, compared to the model test of 59.5 mm, showing only 3 % difference. However, at the pile toe, the deflections from the model test and FLAC3D analysis are 32.8 mm and −7.6 mm, respectively. The FLAC3D analysis predicts the pile rotates as the soil movement increases, while, the model test shows that the pile first rotates and later translates. These differences are attributed to the way axial load is applied on the pile head (by placing weight in the model test and by applying a uniformly distributed vertical stress in FLAC3D ).

Figure 4. Pile response on 50 mm pile at Lm /Ls = 400/300 (Q = 284 N). Table 2.

Summary of FLAC3D and the model test results.

3.3 Summary of results and discussion Table 2 provides a summary of the maximum bending moment shear force and maximum pile deflection. Further comparison between the results from the FLAC3D analysis and the model tests shows : (1) the 104 % difference in the bending moment on the pile without axial load; and (2) the 3% (lowest) difference in the maximum pile deflection with axial load; (3) the 96∼99% difference in shear forces. FLAC3D model over-predicted and underpredicted the magnitude of the pile response (Mmax , Smax , Pile def.), respectively, for the tests with and without axial load. The differences are attributed to the following factors used in the FLAC3D analysis: • the selection of stiffness (Es ) and strength parame-

ters (φ, ψ) of the sand © 2011 by Taylor & Francis Group, LLC

Model 1 Max. value Mmax (Nmm) Smax (N) Pile def. (mm) Model 2

FLAC3D 37,094 235.5 55.5

Model tests 75,672 461.5 79.7

Difference 104 % 96 % 44 %

Max. value Mmax (Nmm) Smax (N) Pile def. (mm)

FLAC3D 52,700 317.9 61.1

Model tests 35,424 159.7 59.5

Difference 49 % 99 % 3%

• the selection of Mohr-Coulomb model, which may

be not able to the real soil behaviour; • using an uniform distributed vertical stress on the

pile elements (Fig. 1a), which is different from the

510

way of placing weight blocks (axial load) on pile head in the model tests (Fig. 2). • the difficulty in capturing the actual mechanics of the moving soil and the stress induced on the pile at large soil movements, when the soil starts to flow around the pile. Parametric analysis has also been carried out to investigate the effect of soil properties on the agreement between the FLAC3D analysis and the model test. This analysis is reported in Ghee (2009). 4

EFFECT OF LM /LS RATIO

The aforementioned two analyses for the model tests have Lm /Ls of 400/300. The Lm /Ls ratio has major impact on the pile response. This is investigated through six FLAC3D models, and presented in form of bending moment and pile deflection profiles. These models have ratios of Lm /Ls ranging from 100/600 to 600/100. The 50 mm diameter pile is again studied here with no axial load on the pile-head. The numerical results at ws = 60 mm are presented in Figures 5a, b for bending moment (smoothed with B-Spine method for better comparison) and pile deflection profiles, respectively. It should be noted that the ws = 60 mm is applied at the boundary of the shear box, and may not represent the actual soil movement as it reaches the pile. In fact, FLAC3D analysis and model test indicate the actual soil movement at the pile location decreased to approximately half of the ws applied at the boundary of the shear box. The shape of the bending moment profiles show either a single curvature at Lm ≤ 200 mm, or a double curvature at Lm = 300 ∼ 500 mm when the pile deflection exceeds the actual soil movement of ∼ws /2. At Lm = 600 mm, the bending moment profile shows a single curvature shape with the M− max of −40,374 Nmm at the depth of 400 mm. The magnitude of this negative moment is approximately the same as M+ max obtained at Lm /Ls = 300/400. In summary, for a fixed pile length, by increasing the Lm /Ls (from 0.17 to 6.00), the M+ max and M− max change, but did not exceed ±40,374 Nmm, respectively. However, the pile may not be stable when Lm /Ls > 400/300, due to its excessive deflection (>50 mm or 1 pile diameter). The deflection profiles show rotation, or rotationtranslation of the pile as the Lm /Ls increases. The deflection attains maximum at the surface, and increases with Lm /Ls ratio. Starting at Lm /Ls = 500/200, the pile also begins to translate through the stable layer (Ls ), as the soil movement increases (the pile deflection exhibits an initial rotation, and then rotation-translation at a higher ws ). Figure 6 shows the Mmax against Smax for the FLAC3D models having Lm /Ls of 100/600 to 600/100 (see Table 3). Each model shows a linear correlation between the Mmax and Smax for any magnitude of soil movement, as is noted in model pile tests (Guo & Qin 2010). © 2011 by Taylor & Francis Group, LLC

Figure 5. The pile response at different Lm /Ls ratios.

Figure 6. The relationship between Mmax and Smax (ws = 30∼120 mm)

511

The gradient of each line (model), α, is obtained and provided in Table 3. The following are noted: • The α generally reduces with the increase in Lm /Ls

until Lm /Ls = 400/300;

• The α of 0.14 (line 5) and 0.25 (line 6) are correct

for the Lm /Ls of 500/200 and 600/100 respectively,

Table 3.

Gradient α of Mmax against Smax relationship.

Lm /Ls

α

Line

100/600 200/500 300/400 400/300 500/200* 600/100*

0.17 0.19 0.16 0.14 0.14 0.25

1 2 3 4 5 6

layers, independent of soil movement level; and (3)The ratio α for the investigated Lm /Ls ratios is 0.17 ∼ 0.25. ACKNOWLEDGEMENTS The work reported was supported by Australian Research Council (DP0209027) and Griffith University School of Engineering.These financial assistances are gratefully acknowledged.

Note: *Both S max and M max are negative

REFERENCES as negative magnitude of both the M−max and the S−max are noted (see Figure 5(a) for the bending moment profile). • The lines 1 to 4 rotate around the origin in an anticlockwise direction, as Lm /Ls increases. This reflects progressive change in the pile movement mode (discussed previously), as Lm /Ls increases. • The ratio α (= Mmax /Smax ) is 0.14∼0.25 from all the FLAC3D models. This range of values are consistent with 0.13∼0.28 obtained theoretically and experimentally by Guo and Qin (2010), regardless of magnitudes of soil movements. 5

CONCLUSIONS

FLAC3D analysis was conducted regarding the model pile tests subjected to lateral soil movement. The predictions show some difficulty in modeling the magnitude and the profile of the measured pile response. However, the ratio α of maximum bending moment Mmax over shear force Smax induced in each pile is well simulated. The FLAC3D analysis shows that: 1) the bending moment and pile deflection profiles change with the increase in Lm /Ls ratios; 2) The ratio α is unique for each model in the stable and moving soil

© 2011 by Taylor & Francis Group, LLC

American Petroleum Institute. 2000. Recommended practice for planning, designing and construction fixed offshore platforms-working stress design, API RP 2A-WSD. Bhattacharya, S. 2003. Pile stability during earthquake liquefaction. PhD Thesis, University of Cambridge. Chae, K. S., Ugai, K. and Wakai, A. 2004. Lateral resistance of short piles and pile groups located near slopes. International Journal of Geomechanics, Vol. 4, No. 2, pp. 93–103. Ghee, E. H. 2009. The behaviour of axially loaded piles subjected to lateral soil movements. PhD Thesis, Griffith University. Guo, W. D. and Ghee, E. H. 2004. Model tests on single piles in sand subjected to lateral soil movement. Proceedings of 18th Australasian Conference on the Mechanics of Structures and Materials, Perth, Vol. 2, pp.997–1004. Guo, W. D. and Qin, H. Q. 2010. Thrust and bending moment for rigid piles subjected to moving soil. Canadian Geotechnical Journal, Vol. 47, No. 1, pp. 180–196. Guo, W. D., Qin, H. Q. and Ghee, E. H. 2006. Effect of soil movement profiles on vertically loaded single piles. International Conference in Physical Modelling in Geotechnics, Hong Kong, pp. 841–846. Itasca. 2002. FLAC3D version 2.1. Fast Lagrangian analysis of continua in three dimensions manual, Itasca Consulting Group, Inc., Minneapolis. Jaky, J. 1944. The coefficient of earth pressure at rest. Journal of the Society of Hungarian Architects and Engineers, pp. 355–358.

512

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Cyclic loading of barrettes in soft calcareous rock using Osterberg cells C.M. Haberfield, D.R. Paul, M.C. Ervin & G.A. Chapman Golder Associates, Melbourne, Australia

ABSTRACT: There remains considerable uncertainty with respect to the cyclic performance of drilled and grouted piles founded in soft calcareous rock. Static and cyclic load testing of three test barrettes founded in a weak carbonate siltstone has been carried out using Osterberg cells. The load test results indicate the shaft resistance performance depends on load and displacement history, and that loss of shaft resistance performance under load reversal may not occur unless the ultimate shaft resistance is achieved during load cycling. 1

INTRODUCTION

There remains considerable uncertainty with respect to the cyclic performance of drilled and grouted piles founded in soft calcareous rock. Recent static load testing of 2 No. 65 m and 1 No. 95 m long test barrettes with cross-sectional dimensions of 1.2 m × 2.8 m for the Nakheel Tower project in Dubai provided valuable data on the performance of cast-in-place deep foundations in calcareous sediments under cyclic load. The testing was carried out using two levels of Osterberg cells placed in the bottom 20 m of the barrettes. As part of the testing programme, an 8 m long section of the shaft of the test barrettes (between the two levels of Osterberg cells) was subjected to cyclic loading. This paper describes the ground conditions and properties of the soft calcareous rock, the construction and testing of the barrettes and the measured performance of the barrettes when subjected to static and cyclic loading.

material is generally massive with no significant joints or discontinuities. A conformable sedimentary sequence (Unit D) lies below Unit C and extends to greater than 200 m depth. This material comprises calcareous siltstone or calcisiltite (depending on carbonate content) and is characterised by layers (upto 3.5 m thick) and nodules (cobble size) of gypsum. The thicker gypsum layers could be correlated between boreholes and suggest a general dip within this material of about 8◦ . UCS testing of the calcareous siltstone and calcisiltite generally indicated strengths between about 1 MPa and 5 MPa. Bulk density varies between 1.8 t/m3 and 2.2 t/m3 , void ratio between 0.5 and 0.7 and carbonate content between 50% and 70%. The gypsum is stronger with unconfined compressive strength varying between about 5 MPa and 15 MPa. Unit D material is also massive with no significant joints or discontinuities. 3

2

GEOLOGY AND SUBSURFACE STRATIGRAPHY

The material in which the test barrettes are located is of Quaternary age, comprising shallow marine sediments deposited as the sea level fluctuated during the onset and decline of ice ages. Recent aeolian deposits form a 20 m thick capping over the site. The ground water is highly saline and ground water level is at a depth of about 2.5 m below ground surface level. A unit comprising predominantly variably cemented, calcisiltite (Unit C) of very low to low rock strength underlies the recent aeolian deposits and extends to a depth of about 72.5 m below ground surface level. Unconfined compressive strength (UCS) tests on samples recovered from high quality coring generally range between 0.5 MPa and 4 MPa. Bulk density varies between 1.6 t/m3 and 2.0 t/m3 , void ratio between 0.6 and 0.8, carbonate content between 40% and 80% and hydraulic conductivity is about 10−7 m/s. The Unit C © 2011 by Taylor & Francis Group, LLC

513

CONSTITUTIVE BEHAVIOUR

Specialist laboratory testing was undertaken to better understand the constitutive behaviour of the Unit C and D materials. Testing comprised cyclic and monotonic constant normal stiffness direct shear testing on concrete/rock interfaces, resonant column testing, drained triaxial testing, cyclic triaxial testing and high pressure oedometer testing. An extensive programme of insitu testing comprising pressuremeter testing, cross hole seismic testing and water pressure testing was also undertaken. As indicated below, from our observations of the core sample and comparison of the insitu and laboratory test results, it became apparent that, when sampled and brought to the surface, the Unit C and D materials underwent significant stress relief. This resulted in samples tested in the laboratory showing significantly lower strength and deformation properties than measured by insitu testing. Significant emphasis was therefore placed on the results of the insitu testing.

Nevertheless, the laboratory tests provided useful insights into the constitutive behaviour of the Unit C and D materials. Unit C and D materials have a relatively high stiffness below a “bond yield strength” after which the compressibility of the material increases significantly and exhibits properties similar to an uncemented, normally consolidated material at the same void ratio. Prior to reaching the bond yield strength the rock displays approximately linear elastic behaviour with deformations occurring essentially instantaneously. As the bond yield strength is approached, deformations become time dependent and consolidation and creep displacements dominate. Strength testing indicates the behaviour of the rock is dominated by intergranular cementation with little apparent frictional component to strength. The Tresca yield criterion was found to provide a reasonable basis for modeling the behavior of the Unit C and D materials up to bond yield strength. 4

ENGINEERING PROPERTIES

Figure 1 compares the Young’s modulus values estimated from the pressuremeter (initial loading modulus), cross hole seismic and laboratory UCS tests. The pressuremeter test results displayed similar initial loading and unload-reload moduli values which is consistent with the absence of jointing in the rock and the domination of the cementation. The Young’s modulus values obtained from the pressuremeter and cross hole seismic tests show reasonable agreement (see Figure 1) if the small strain modulus values obtained in the cross hole seismic tests are reduced by a factor of five (which is consistent with published data on a range of soil and rock types which compare modulus at different strain levels). Figure 2 compares the shear strengths measured in the UCS tests (taken as UCS/2) and those estimated from the pressuremeter tests assuming a purely cohesive strength criterion (Tresca criterion). We note that the use of the Tresca criteria for rock would be considered unusual as it assumes no frictional component to strength. However, for this material it would appear to be reasonable on the basis of the constitutive behavior observed in the laboratory tests (including drained triaxial tests) where pre-peak strength behavior was dominated by cementation rather than friction). The Tresca criterion also provided a very good fit to the pressuremeter curves up to significant strain levels (8% cavity strain) and was adopted for all subsequent modeling of the test barrettes (see below). Figures 1 and 2 show that stiffness and strength properties measured in the laboratory were significantly less than obtained from insitu tests, and supported a hypothesis that the core samples were undergoing significant stress relief even with the care that was undertaken during the drilling, retrieval, storage, transportation and testing processes. The load testing carried out on the test barrettes (see below) confirmed that the properties obtained from the insitu testing were reasonable and that the © 2011 by Taylor & Francis Group, LLC

514

Figure 1. Young’s modulus variation with elevation (surface level RL + 2.5 m DMD).

Figure 2. Shear strength variation with elevation (surface level RL + 2.5 m DM).

laboratory test results significantly under-estimated the properties of the insitu rock. Constant normal direct shear testing was undertaken on concrete/rock interfaces using a range of interface roughnesses (including smooth), normal stiffnesses and initial normal stresses. The results of the tests indicate a residual interface friction angle of about 37◦ .

5 TEST BARRETTES Three test barrettes with cross-sectional dimensions of 1.2 m × 2.8 m were installed to depths of 65 m (TB02 and TB03) and 95 m (TB01) and tested in accordance with specifications provided by Golder Associates Pty Ltd. Test barrette TB02 was installed at the same location as the investigation borehole BH208. Test barrette TB01 was installed about 12 m south east of TB02 and TB03 about 8 m due south of TB02 resulting in a minimum clear distance between test barrettes of about 6 m. The lengths of the barrettes were chosen to provide information on barrette performance in the Unit C and D materials. The test barrettes were installed by a SoletancheBachy/Intrafor Joint Venture using hydrofraise equipment with polymer support. The hydrofraise cutting action results in a relatively smooth excavated surface and hence a concrete rock interface which is essentially devoid of roughness. High slump concrete was placed by tremie. Concrete design characteristic 28 day strength was 60 MPa. Load testing of the barrettes was carried out by Loadtest International Inc under the direction of Golder Associates Pty Ltd. The load tests comprised two levels of Osterberg cells in each test barrette as shown in Figure 3. The Osterberg cells were positioned to measure performance of the lower 20 m or so of the barrettes. The test barrettes were instrumented with displacement tell-tales and strain gauges. In addition, instrumentation was also located in the rock below the toe of the barrettes to directly measure the displacement of the rock at this location. 6

STATIC TEST RESULTS

The measured load versus displacement performance of the two shorter test barrettes (TB02 and TB03) for loading at the lower (LOC) and upper (UOC) levels of Osterberg cells are shown in Figures 4 and 5 respectively. Also shown are predictions of the performance. The predictions were obtained on the basis of the design strength and deformation properties for the site (residual friction angle from CNS tests, strength and deformation properties from pressuremeter tests) and on the as-constructed barrette geometry. The predictions of performance were completed prior to testing of the barrettes. For the Class A prediction, the rock-socket software ROCKET97 (Seidel, 2000) was used to calculate the shaft resistance versus deformation response of the relatively smooth, short test sections of the test barrettes. These analyses assumed a relatively smooth barretterock interface. The shaft resistance versus deformation response calculated in the ROCKET97 analyses were then adopted as the barrette-rock interface behavior in an axisymmetric PLAXIS V8 non-linear finite element model to calculate the load versus displacement responses shown in Figures 4 and 5. The comparison between the measured and predicted response is excellent, which provided further confidence that the design © 2011 by Taylor & Francis Group, LLC

515

Figure 3. Shear strength variation with elevation (surface level RL + 2.5 m DMD).

strength and stiffness properties adopted on the basis of the insitu pressuremeter testing were appropriate. Similar comparisons between measured and predicted results were obtained for the deeper test barrette TB01.

7

CYCLIC TEST RESULTS

Following the initial static load testing, both the upper and lower Osterberg cells were sufficiently “open” to allow cyclic testing of the shaft between the upper and lower levels of Osterberg cells. This was undertaken, for example, by pressurising the UOC, while allowing the LOC to bleed off any pressure caused by the shaft between the UOC and LOC moving downwards. Once the nominated downward displacement had been achieved, the process was reversed by depressurising the UOC then pressurising the LOC while allowing the UOC to bleed off any pressure caused by the upwards movement of the shaft between the UOC and LOC. This process was repeated a number of times to investigate the cyclic loading behaviour of the shaft. The initial loading of test barrettes TB01 and TB02 was undertaken by increasing the load in the UOC, whereas the initial loading for test barrette TB03 was undertaken by increasing the load in the LOC. This has implications for the cyclic loading results set out

Figure 6. TB01 – Average shaft resistance versus displacement response for barrette shaft between UOC and LOC. Figure 4. Measured vs predicted performance for loading at upper Osterberg cells.

The average displacement is the average of the measured displacements at the bottom plate of the UOC and top plate of the LOC. Annotations shown on these figures describe various aspects of the tests and shaft resistance behaviour. The cyclic load testing also incorporated significant hold stages during which the load in the Osterberg cells was kept constant. No significant creep displacement was observed during these hold stages. The results for the deepest test barrette TB01 shown in Figure 6 indicate essentially elastic behaviour with a maximum mobilised shaft resistance obtained in the test (for both upwards and downwards loading) of about 1250 kPa. We note the ultimate shaft resistance was not achieved in either upwards or downwards loading and that the response is very stiff with a maximum displacement of less than 3.5 mm at maximum load. The results for TB02 shown in Figure 7 indicate an ultimate average shaft resistance for downward loading of about 550 kPa. At some stages of the test, higher apparent average shaft resistances were measured. However, these were due to contributions from elsewhere along the shaft and should be ignored. Figure 7 also indicates the following behaviour: Figure 5. Measured vs predicated performance for loading at lower Osterberg cells.

below. The average shaft resistance versus displacement performance of the barrette shaft between the UOC and LOC are shown in Figures 6, 7 and 8 for TB01, TB02 and TB03 respectively. The average shaft resistance was calculated by dividing the load difference between UOC and LOC by the perimeter area of the shaft between UOC and LOC. © 2011 by Taylor & Francis Group, LLC

1. A significant reduction (from about 550 kPa to about 250 kPa) in ultimate shaft resistance on complete shear reversal from downwards loading to upwards loading. We note that this was not observed with TB01 and may suggest that the reduction in shaft resistance under tension loading may only occur if the ultimate shaft resistance is exceeded. 2. No reduction in shaft resistance performance during cyclic loading for downwards loading provided the displacement on reversal (i.e. in upwards

516

Figure 7. TB02 – Average shaft resistance versus displacement response for barrette shaft between UOC and LOC.

Figure 8. TB03 – Average shaft resistance versus displacement response for barrette shaft between UOC and LOC.

loading) is less than required to achieve peak shaft resistance (for upwards loading). 3. No significant reduction in ultimate shaft resistance with displacement under downwards loading. 4. Shaft resistance versus displacement behavior during cycling appears to depend on the magnitude of post-peak displacement experienced during upwards loading. 5. Following cycling, full shaft resistance is achieved with further downwards displacement. © 2011 by Taylor & Francis Group, LLC

6. The stiffness of the unloading response (whether in upwards or downwards loading) is relatively constant across all loading cycles.

517

The results for TB03 shown in Figure 8 are reasonably consistent with those for TB02 and indicate an ultimate average shaft resistance of 600 kPa for downward loading and about 300 kPa to about 100 kPa for upward loading. Note that during some cycles higher (and increasing) negative shaft resistances were

measured and these were due to errors in the testing procedure which allowed resistance contributions from other parts of the barrette. Comparison of Figures 7 (TB02) and 8 (TB03) indicates that the shaft resistance in TB03 is mobilised at significantly lower displacement than for TB02. The ultimate shaft resistance measured in TB03 is also slightly higher than for TB02. This is consistent with other data obtained from the tests (e.g. base resistance) and indicates the ground at TB03 may be stronger and stiffer than at TB02. The results for TB02 and TB03 appear to indicate a significant loss in shaft resistance on reversal of loading. However, it is interesting to note that loading of the test section for pile TB01 and TB02 initially occurred by pushing the test section downwards, whereas TB03 the test section was pushed upwards. Ultimate shaft resistance was not achieved during testing of TB01 and there was no loss in shaft resistance on reversal of load. For TB02, ultimate shaft resistance was achieved during initial downwards loading and on load reversal a significant reduction in ultimate shaft resistance was observed for upwards loading. For TB03, ultimate shaft resistance was not achieved in initial upwards loading, but on load reversal, ultimate shaft resistance (about 600 kPa) was achieved in downwards loading. On load reversal again to upwards loading, a significant reduction in ultimate shaft resistance was observed. It would appear that the reduction in ultimate shaft resistance may only occur if ultimate shaft resistance is achieved prior to load reversal. The measured values of ultimate shaft resistance (downwards loading) for the test barrettes TB02 (550 kPa) and TB03 (600 kPa) are reasonably consistent with those estimated using the estimated pressure of the fluid concrete and the residual friction angle of the concrete – rock interface (580 kPa). This is probably to be expected due to the relatively smooth barrette-rock interfaces formed using the hydrofraise equipment. The measured ultimate shaft resistance for TB01 (1250 kPa) is higher than estimated using this

© 2011 by Taylor & Francis Group, LLC

simplistic approach (1000 kPa) and this may be potentially explained by greater roughness of the interface within the gypsum layer and the higher strength of the gypsum compared to the calcisiltite.

8

CONCLUSIONS

The results of load tests on three test barrettes in calcareous siltstone indicates that for the barrettes tested, it would appear that: 1. a reasonable estimate of ultimate shaft resistance for smooth barrette-rock interfaces can be made from the pressure applied by the fluid concrete and the residual friction angle of the concrete-rock interface. 2. the ultimate shaft resistance measured on load reversal (following loading to ultimate shaft resistance) is significantly less than measured under static loading. 3. this reduction in ultimate shaft resistance may only occur if ultimate shaft resistance is achieved prior to load reversal. 4. the stiffness of the unloading response (whether in upwards or downwards loading) is relatively constant across all loading cycles.

ACKNOWLEDGEMENTS The authors gratefully acknowledge the assistance of Nakheel, Soletanche-Bachy/Intrafor Joint Venture and LoadTest International; and Foundation QA for the use of Rocket.

518

REFERENCE Seidel J.P. (2000) ROCKET97 Help Manual. Department of Civil Engineering, Monash University

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Shaft capacity of drilled and grouted piles in calcareous sandstone B.M. Lehane University of Western Australia & Consultant to Arup Australasia

ABSTRACT: The paper addresses the significant shortage of full scale load test data for drilled and grouted (D&G) piles in weak calcareous rocks by presenting results from an instrumented pile test programme that involved tension testing of 240 mm, 340 mm and 450 mm diameter D&G piles at a calcareous sandstone site north of Perth, Australia. It is shown that the capacities predicted using a variety of approaches employed by local practitioners are often significantly larger than measured static capacities, and that these approaches themselves yield a wide range of capacities. The testing programme highlighted the considerable uncertainties associated with the prediction of both static and cyclic pile response in variable soft rock deposits such as the Tamala Limestone. 1

INTRODUCTION

Drilled and grouted (D&G) piles are presently the preferred pile type to resist axial loads in the coastal limestones and calcareous soils present along the west coast of Western Australia. There is, however, a great shortage of reported field load tests on such piles and consequently a diverse range of design approaches are in current use. Motivated by the need for actual field data to test the validity of various approaches, Arup International, supported by Belpile Pty. Ltd., commissioned three static tension load tests (followed by cyclic testing) on instrumented D&G piles installed in a calcareous sandstone. This paper presents the results of these tests and compares the observed capacities with Class A predictions made by local practitioners. 2

SITE LOCATION

The site selection process was constrained by the need to be within easy reach of Perth city centre (thereby reducing costs associated with piling). The selected site was a limestone quarry in Pinjar, which is about 25 km north of Perth. The rock in this quarry was described by its owner as a “medium grade limestone”, which forms part of the Tamala Limestone formation. Piles were installed in the centre of the quarry, which had been excavated some years ago to a depth of about 10 m below the surrounding ground level. 3

GROUND CONDITIONS

Cone Penetration Tests (CPTs) were conducted in advance of the piling and included tests within 1m of each of the pile test locations. The CPT end resistance (qc ) profiles at these locations are shown on Figure 1 and indicate qc values typically varying from about © 2011 by Taylor & Francis Group, LLC

Figure 1. CPT qc profiles in the vicinity of the test piles.

15 MPa to 50 MPa; some horizons with low qc values (<10 MPa) and other harder layers with qc in excess of 80 MPa are also in evidence. It should be noted that no CPT data were available below a depth of 3m at the location of pile P450 (discussed later), due to refusal at this depth. All 10 CPTs conducted at the site indicate very significant lateral and vertical variability in the qc values. CPT friction ratios (Rf ) were consistently between 0.5% and 1.0%, except in the low qc layers (qc < 10 MPa), where Rf values were typically between 1.5% and 2%. The qc values shown on Figure 1 are at least double typical values shown by

519

Figure 2. Photos of typical cores recovered.

the (un-cemented) Spearwood dune sand in Perth city centre, which overlies the Tamala Limestone. Two rotary cored boreholes were drilled in close proximity to the pile test area, although the recovery from these boreholes was poor (generally less than 75%). The boreholes, which remained open and stable during drilling, indicated that the deposit comprised weakly cemented ‘cobbles’ in an un-cemented or very weakly cemented sandy matrix (as shown in upper photo in Figure 2; it would appear that the drilling process washed out the un-cemented material. About half of the core recovered was relatively intact as shown by the lower photo on Figure 2, although solid core lengths were generally less than 100 mm long (i.e. the rock quality designation, RQD, was close to zero). Tests revealed that the deposit’s average grain size is about 0.3 mm and that it has an average calcium carbonate content of 42% with a measured range of 16% to 55%. The deposit therefore classifies as a calcareous sandstone according to the Clark & Walker (1977) classification system. Further tests indicated that: • The Point Load Index (Is,50 ) values for the ‘cobbles’

was typically 0.15 ± 0.05 MPa, which equates to an unconfined compressive strength (UCS) of at least 1 MPa (assessed based on local experience in this type of rock, e.g. Cocks 2010).

© 2011 by Taylor & Francis Group, LLC

Figure 3. Measurements obtained in Constant Normal Stiffness (CNS) tests with k = 400 kPa/mm. • The degree of saturation of the samples was typi-

cally in excess of 75% and therefore suction pressures are unlikely to be significant. • Two UCS tests were performed, but these gave very low strengths (less than 50 kPa), perhaps reflecting the stratified nature (i.e. cemented and un-cemented layers) of the deposit and of the samples. Constant normal stiffness (CNS) shear tests were performed on samples from between 2.6 m and 3.6 m depth. The results from tests on two samples are shown on Figure 3; these were consolidated to an initial vertical effective stress of 50 kPa and then sheared to failure with a normal stiffness of 400 kPa/mm. Inspection of Figure 3 indicates that, although these samples were obtained from a soil horizon with a CPT qc value of about 25 MPa, their shear strengths are compatible with relatively low c
(cementation) values

520

Table 1.

Details of pile tests at Pinjar.

Table 2.

Steel tube Steel tube Diameter Length outer diam. wall thickness Pile No. (mm) (m) (mm) (mm) P240 P340 P450

240 340 450

5 5 4.9

168 273 356

4.8 4.8 6.4

Pile capacities at Pinjar.

Pile

Tmax (kN)

Wstatic (mm)

Tcyclic (kN)

Wcyclic (mm)

Tpost−cyclic (kN)

P240 P340 P450

1290 1500 940

12 25 40

0–600 0–1050 0–600

1 7 60

1155 1200 550

of between 5 kPa and 20 kPa (for an assumed friction angle, φ
, of 37 ± 3◦ ); the higher strength of the sample from 2.75m is seen to arise primarily from its more dilatant response. The samples’relatively low c
values are consistent with the low UCS values measured, suggesting that either (i) the UCS and c
for material with a CPT qc < 25 MPa is very small or (ii) poor sampling destroyed some of the sample’s in-situ structure and cementation. Standard CPT correlations with relative density for un-cemented sand cannot explain the high levels of qc observed suggesting that some degree of sample disturbance took place. Cyclic CNS tests on 2 samples, involving 2-way cycling with a displacement amplitude of 5 mm, a normal stress of 50 kPa and a normal stiffness (k) of 400 kPa/mm, indicated that the samples’shear strength had reduced by over 90% after only about five cycles (i.e. accumulated relative displacement ∼50 mm). 4

PILE TEST PROGRAMME

5.1 m deep bores for the piles were drilled in September 2009 by Belpile Pty. Ltd. using 225 mm, 325 mm and 440 mm diameter augers. Experience with drilling using these augers in similar ground conditions indicated that the final bore diameters formed were 240 mm, 340 mm and 450 mm. The holes were tremied grouted with a 30 MPa (28 daycube strength) grout before insertion of the steel tubes detailed in Table 1. These tubes were beaded to ensure good shear transfer between the grout the tube. The tubes were instrumented with 20 quarter bridge strain gauges with two gauges at 10 levels. The piles were tested in tension six weeks after their installation. As seen in Figure 4, tension was applied using a hydraulic jack sitting on a reaction beam pushing a steel plate that was attached to another plate at the pile head using 4 high strength Macalloy steel bars. The plate at the pile head was bolted to another four, 1 m long, Macalloy bars which were cast into the centre of each pile. The load was increased in increments with typically between 10 and 20 maintained load increments being required to attain ultimate tension capacity; the holding/creep periods during increments were between 5 and 10 mins. Subsequent to this static phase, the piles were unloaded and then subjected to 10 one-way cycles varying from 5 kN to 47%, 70% and 64% of the respective peak tension capacities of P240, P340 and P450. This cycling was followed by monotonic loading of each pile to failure. © 2011 by Taylor & Francis Group, LLC

Figure 4. Tension pile testing set up at Pinjar.

5

PILE TEST RESULTS

One of the pile load displacement responses observed is shown on Figure 5 for the case of the 450 mm diameter pile (P450). Once the tension load on this pile reached 940 kN, when the head displacement was about 10 mm, the pile continued to move steadily out of the ground. Significant pile head movement is observed after unloading and subsequent cycling to a maximum tension load of 600 kN, with over 100 mm displacement accumulated after application of 10 cycles. Subsequent static loading of this pile indicated a post-cyclic capacity of 550 kN i.e. less than 60% of the initial static capacity. The capacities measured in the three pile load tests are listed in Table 2. Tmax is defined as the maximum applied tension load and wstatic is the pile

521

Figure 5. Pile head load displacement response for P450.

head movement experienced before unloading while attempts were made to maintain a load of Tmax at the pile head. The pile head movement accumulated during ten cycles is referred to as wcyclic in Table 2 while Tpost-cyclic is the post-cyclic capacity. Some observations from Table 2 include: (i) Surprisingly, P450 has significantly lower tension capacity than both P340 and P240; (ii) There is a clear trend for the post-peak capacity to be related to the degree of relative displacement between the pile and soil; (iii) Displacements accumulated during cycling increase as the cyclic load level increases e.g. wcyclic for P240 (with Tcyclic /Tmax = 0.47) is far less than that of P340 (with Tcyclic /Tmax = 0.7); At broadly the same Tcyclic /Tmax value, the cycling induced displacement (wcyclic ) of P340 is substantially less than that of P450. The variations of the average shaft shear stress with pile head displacement, normalised by the pile diameter, (w/D) recorded during the (initial) static component of the three tests are shown on Figure 6 (assuming zero base capacity for all piles). It is apparent that displacements in excess of 0.03D are required to generate peak average frictions (τavp ) for P240 and P340 while peak friction is mobilized at 0.012D for P450; these peak values reduce with pile diameter. If β is defined as the ratio of the average CPT end resistance (qc,avg ) to the peak shaft shear stress, for an approximate qc,avg value in the vicinity of the piles of 28 MPa, β was 90 ± 10 for P240 and P340 and 200 for P450. The β value for P450 is more than 2.5 times the average β value reported by Randolph et al. (1996) for 440 mm diameter D&G pile sections in the calcarenite at Overland Corner – and is actually greater than the β value of 160 ± 20 indicated by the database compiled in Lehane (2009) for bored piles of the same diameter in un-cemented sand (constructed by the same Contractor and equipment). Given the spatial variability of qc values indicated by the CPTs and the absence of © 2011 by Taylor & Francis Group, LLC

Figure 6. Development of shaft shear stress in three pile tests.

qc data below 3 m at the location of P450, it is therefore probable that a high proportion of material in the vicinity of the shaft of P450 was either un-cemented or very weakly cemented sand. This inference is also supported by the significant cyclic degradation of this pile and the parallel degradation shown in the cyclic CNS tests in samples with relatively low c
values (mentioned in Section 2). The trends evident from Figure 6 reflect the importance of dilation and cementation coupled with a complicated dependence of shaft friction on the CPT qc value. These characteristics are discussed in Lehane (2009) and are also examined in a separate paper (in preparation) which presents the results from the strain gauge data. These strain gauge results show that peak local frictions (τp ) on piles P240 and P340 range from between 140 kPa and 500 kPa, with the highest frictions developed between 2 m and 3 m depth in both piles and also below 4.5 m in P240. Values of τp for P450 were relatively uniform at 140 ± 20 kPa and appear to correspond to the available friction in the absence of any significant cementation. 6

PREDICTED AND MEASURED CAPACITIES

Well known practitioners, involved in piling in the Perth area and in the Australian NWS, were invited to provide Class A predictions of the capacities of the three test piles. All information described in the foregoing that related to the ground conditions, the laboratory tests and the piles was provided to the predictors. Calcium carbonate contents were not, however, known

522

7

Figure 7. Comparison of predicted with measured static pile tension capacities.

by the predictors and it was suggested, when invitations were sent out, that the average content was likely to be higher than the actual mean measured value of 42%. Six predictions were received from four separate geotechnical companies. To assess peak static capacity, two predictors employed correlations between peak friction and the unconfined compression strength (UCS) proposed by Williams & Pells (1981) and Abbs & Needham (1995), where the UCS was taken between as qc /18 in one case and as qc /25 in the other. Two predictors used their own company’s proprietary software and the two remaining predictors did not reveal the predictive method employed. Predicted static capacities (Tmax ) are compared in Figure 7 with the measured capacities. It is evident that there is a wide range of predictions. This range translates to a mean coefficient of variation (COV) of the predicted to measured capacities of 0.48; this COV is well above standard expectations and presumably reflects both the range of different predictive methods employed and the significant site variability evident from the CPT data. Predictors were also invited to assess the likely degree of degradation of shaft capacity under 10 cycles (to 66% of the static capacity) and to provide a prediction for the post-cyclic capacity (Tpost-cyclic ). Despite the quite dramatic loss in cyclic strength indicated by the (weakly cemented) samples tested under cyclic CNS conditions, all predictors estimated a Tpost-cyclic /Tmax ratio of 90 ± 5%; this ratio contrasts with measured ratios of 80% and 58% observed for P340 and P450 respectively. It is probable that predictors took a view that the CNS data were not representative of the majority of the in-situ material. © 2011 by Taylor & Francis Group, LLC

CONCLUSIONS

The investigation presented in this paper illustrated some of the difficulties associated with making reliable predictions of pile capacity in weak calcareous rocks. A relatively intensive CPT investigation revealed significant variability in the mechanical characteristics of the rock and interpretation of appropriate parameters was further complicated by poor core recovery in boreholes and potential significant disturbance to recovered samples. Such features are often encountered in practice and need to be accommodated in design. The tension pile tests indicated that the average shaft shear stress at peak static capacity (τavp ) reduced with an increase in pile diameter. This trend is consistent with the influence of dilation at the shaft interface, but the specific diameter dependence observed was probably influenced by varying levels of cementation at the specific pile test locations. Correlations between shaft friction and CPT qc end resistances are seen to be more appropriate than those using UCS values in the type of weak rock present at the test site. Measured ultimate peak shaft frictions were typically between 0.5% and 1% of qc and were lower than expected based on extrapolation from other field D&G pile tests in calcarenite. The prediction exercise highlighted both the range of different predictive methods employed and the influence that significant site variability (as indicated by the CPT data) can have on the capacity range. The available shaft friction degraded during cycling and reduced by about 40% after an accumulated relative displacement of 100 mm. The testing programme results demonstrated the considerable uncertainties associated with the prediction of cyclic pile response in variable soft rock deposits such as the Tamala Limestone.

ACKNOWLEDGMENTS The author is grateful to Mr. Colin Zampatti for the permission provided to use his quarry for the test programme. The author also gratefully acknowledges the support provided by the Arup Design and Technical (DTX) research fund and the contribution of Belpile Pty Ltd to the overall success of the project. The assistance provided by Bhavikhjit Singh, who is currently undertaking postgraduate studies at UWA, is appreciated – as is the time taken by the six practitioners who provided predictions of pile capacity.

REFERENCES Abbs, A.F. and Needham, A.D. (1985). Grouted piles in weak carbonate rock. Proc. Offshore Technology Conference, Paper No. OTC 4852, Houston, 105–112. Clarke, A.R. and Walker, B.F. (1977). A proposed scheme for the classification and nomenclature for use in the

523

engineering description of Middle Eastern sedimentary rocks. Geotechnique, 27(1), 93–99. Cocks, G. (2010). Personal communication. Lehane, B.M. (2009). Relationships between axial capacity and CPT qc for bored piles in sand. Keynote Lecture, Proc. 5th International Symposium on deep foundations on bored and auger piles, Ghent, 1, 61–76, Taylor Francis Group, UK.

© 2011 by Taylor & Francis Group, LLC

Randolph, M.F., Joer, H.A., Khorshid, M.S. and Hyden, A.M. (1996). Field and laboratory data from pile load tests in calcareous soils. Proc. Offshore Technology Conference, Paper No. OTC 7992, Houston, 327–336. Williams, A.F. and Pells, P.J.N. (1981). Side resistance of rock sockets in sandstone, mudstone and shale. Canadian Geotechnical Journal, 18(4), 502–513.

524

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Numerical analysis of mudmat contribution to capacity of piled offshore platforms L.S.D. Lorenti Senior Geotechnical Engineer, Arup Pty. Ltd., Perth, Australia

M.A. Ismail Formerly Arup Pty. Ltd., Perth, Australia

B.M. Lehane School of Civil Engineering, University of Western Australia, Consultant to Arup Australasia

ABSTRACT: Mudmat foundations are traditionally used for the temporary support of steel jacket structures prior to installation of the piled foundation system. The contribution of the mudmats is typically excluded from the analysis of structural response post installation. However, there are ground conditions that lend themselves to inclusion of the mudmats contribution to foundation capacity, particularly in situations where extension of the production life of existing facilities is demanded. Inclusion of the mudmats will have an effect on displacements, structural forces and the dynamic amplification of environmental loads. The response of a piled foundation system, including and excluding a mudmat for a ground profile typical of that found in the North West Shelf of Australia (NWS) was predicted using numerical techniques. The effect of including the mudmat in the analyses is shown to be generally beneficial.

1

INTRODUCTION

to a frame analysis of the assumed jacket structure to study the structural response.

Mudmat foundations are traditionally used to provide temporary support for steel jackets prior to installation of the piled foundation system. In the North West Shelf of Australia these piles are usually founded in competent rock strata at depth. This paper explores the possibility of using mudmats as part of the permanent foundation system for ground profiles where weakly to moderately cemented calcareous sand overlies well to very well cemented calcarenite or calcareous sandstone, which is typical of the NWS. The paper first describes the hypothetical problem under analysis including the assumed stratigraphy, material parameters, foundation systems and jacket configuration. A simplified model with one pile per leg was chosen. The paper then describes the finite element analyses performed in Plaxis 2D to study the axial behaviour of the assumed foundation systems. This allowed comparison of the vertical stiffness of the individual foundation elements with the combined foundation system. Plaxis 3D Foundation analyses were then undertaken to study the combined horizontal, moment and vertical loading of a single pile and mudmat system. The horizontal, rotational and vertical stiffness values derived from these analyses were finally applied © 2011 by Taylor & Francis Group, LLC

2

PROBLEM UNDER ANALYSIS

The problem analysed in the 2-dimensional and 3-dimensional finite element analyses performed was for a hypothetical (but typical) layered stratigraphy supporting a steel jacket structure.

2.1 Assumed stratigraphy The stratigraphy used in the numerical modelling is presented in Table 1 and consists of a layer of weakly to moderately cemented calcareous sand overlying well to very well cemented calcarenite or calcareous sandstone. Table 1. Assumed ground profile. Material

UCS (MPa)

Thickness (m)

Weakly to Moderately Cemented Calcareous Sand Well to Very Well Cemented Calcarenite/Calcareous Sandstone

0.2–0.4

20

5.0

Base of model

525

2.2 Assumed jacket structure The jacket structure assumed to be supported by the hypothetical foundation systems is presented in Figure 1. In general terms, the jacket is 40 m tall and consists of two identical frames spaced 20 m apart, each with two main legs raked at 1H: 4V (in one direction only). The plan area at the top of the jacket is 20 m × 20 m. The jacket has four equally spaced (vertically) sets of lateral bracing with one diagonal member bracing each bay. The plan area at the base of the jacket (at the connection with the mudline) is 40 m × 20 m. The total buoyant weight of the assumed jacket is 552 tonnes (5.4 MN). The diameters and wall thicknesses of the three structural elements assumed to form the jacket are presented in Table 2. 2.2.1 Assumed loading In addition to self weight (5.4 MN), the jacket has been assumed to carry a topsides load of 8000 tonnes (80 MN) and withstand environmental applied loads of 5.8 MN (horizontally) and 232 MNm (overturning). 2.2.2 Assumed foundation system The foundation systems assumed to carry the jacket, topsides and any applied environmental loading are (i) isolated piles and (ii) pile and mudmat in combination. The piles modelled to support the jacket structure were assumed to extend to a depth of 30 m below the mudline with a 10 m long rock socket formed in the strong calcarenite/calcareous sandstone, with a diameter of 2 m. It was also assumed that the base of the pile does not contribute to either the capacity or stiffness

of the foundation system. This has been explicitly considered in the numerical modelling performed. The mudmats modelled were assumed to have a diameter of 10 m and included a central hole of 2 m in diameter to permit penetration of the pile. A stiffness of 200 GPa was used for the plate element representing the mudmat, although a thickness of 2 m was specified to simulate fully rigid conditions. The pile and mudmat were considered rigidly connected by the program. For foundation system (i), one pile was assumed to support each of the jacket’s main legs. In foundation system (ii), one pile and mudmat in combination was assumed at the end of each main leg. 2.2.3 Analysis sequence The analysis sequence used in the 2D numerical analysis is presented in Table 3. In each case the foundation system was “wished in place” and a vertical prescribed displacement of 40 mm was applied.All materials were declared drained throughout the analysis sequence. The analysis sequence used in the 3D numerical analysis is presented in Table 4. In each case the foundation system was “wished in place” and the loads described in Section 2.2.1 were applied. The calcareous sand layer was declared undrained throughout the analysis sequence, whilst the calcarenite/calcareous sandstone at depth was retained drained. Table 3.

Follows Stage Stage Description 1 2

– 1

3

1

4

1

Table 4.

Figure 1. Assumed jacket structure. Table 2. Assumed jacket structural elements.

Structural Element

Outside Diameter (mm)

Wall Thickness (mm)

Main Leg Lateral Bracing Diagonal Bracing

1829 762 508

51 38 25

© 2011 by Taylor & Francis Group, LLC

2-Dimensional model analysis sequence.

Initialisation Install mudmat and apply vertical prescribed displacement of 40 mm Install pile and mudmat and apply vertical prescribed displacement of 40 mm Install pile and apply vertical prescribed displacement of 40 mm 3-Dimensional model analysis sequence.

Stage

Follows Stage

Description

1 2 3 4 5 6 7 8 9 10 11 12 13

– 1 2 3 4 5 6 7 1 9 10 11 12

Initialisation Install mudmat Apply 1.35 MN to mudmat ( = 5.4/4) Install pile Apply 20 MN to pile ( = 080/4) Drained conditions Apply lateral load of 1.45 MN ( = 5.8/4) Apply moment of 58 MNm ( = 232/4) Install isolated pile Apply 21.35 MN ( = 85.4/4) Drained conditions Apply lateral load of 1.45 MN ( = 5.8/4) Apply moment of 58 MNm ( = 232/4)

Note: Pile axial load produced by the overturning moment was ignored

526

The calcareous sands were declared undrained inthe 3D analysis to capture the reduced strength and stiffness of this layer during the application of the horizontal and moment loads, assumed to be associated with storm events.

3

NUMERICAL MODELLING

3.1 2-Dimensional numerical analyses 3.1.1 General Numerical analyses were performed to calculate the vertical stiffness of the isolated pile, isolated mudmat and combined pile and mudmat foundation system. These analyses were performed in the 2D (axisymmetric) version of Plaxis (Plaxis, 2008). This software was chosen as it provides advanced soil constitutive models capable of predicting the non-linear loaddisplacement response of the foundation system.

It should be noted that the non-linear, Hardening Soil constitutive model was used for the calcareous sand in the numerical modelling. However, because cemented soils show reduced dependency of stiffness on effective stress (e.g., Ismail et al., 2005) it was decided to use m = 0.0 (Table 5) for the cemented calcareous sand, implying a constant stiffness. As shown in Table 5 the Mohr-Coulomb constitutive model was used for the calcarenite/calcareous sandstone at depth. The rock therefore has a fixed soil modulus for initial loading, unloading and reloading events. In addition, a parametric study, designed to investigate the effect of modifying the strength and stiffness of the cemented calcareous sand was undertaken. The alternate parameters used in these analyses are presented in Table 7.

Table 5.

3.1.2 Finite Element model The Finite Element mesh used is presented in Figure 2. The mudmat was modelled using plate elements with the dimensions provided in Section 2.2.2. The pile was modelled using solid elements. In order to exclude end bearing from the contribution to pile vertical stiffness, a “hole” was created below pile toe level. This modelling detail accords with the assumed foundation details presented in Section 2.2.2. The 2D axisymmetric model was built using the internal meshing functionality in Plaxis and in total there were 7,290 elements in the model. The assumed soil stratigraphy presented in Table 1 was replicated in the numerical model. The far-field boundary was placed 55 m from the edge of the mudmat, a distance of ∼5 mudmat diameters. 3.1.3 Analysis parameters The analysis of the vertical stiffness of the pile only, mudmat only and combined pile and mudmat foundation systems was performed using the “base case” set of analysis parameters provided in Tables 5 and 6.

Base case parameters: Cemented calcareous sand.

Parameter Description

Value

Saturated Unit Weight (γsat , kN/m3 ) E50−ref (MPa) (Modulus at 50 % strength at pref ) Eoed−ref (MPa) (Constrained Modulus at 50 % strength at pref ) Eur−ref (MPa) (Unload-reload modulus) m (dimensionless exponent) (Determines dependency of stiffness on σ3
) c
(kPa) (cohesion) φ
(degrees) (Friction angle) ψ
(degrees) (Dilatancy angle) νur (Unload-reload Poisson ratio) tan δ/tan φ
(interface strength ratio)

16 80 80 240 0 100 40 0 0.2 0.7

Note: pref = in situ effective horizontal stress Table 6. Base case parameters: Calcarenite/Calcareous sandstone. Parameter Description

Value

Saturated Unit Weight (γsat , kN/m3 ) E (MPa) c
(kPa) φ
(degrees) ψ
(degrees) ν tan δ/ tan φ


20 1,000 900 33 0 0.2 0.7

Table 7. Parametric study parameters: Cemented calcareous sand.

Figure 2. Plaxis 2D analysis model.

© 2011 by Taylor & Francis Group, LLC

527

Parameter Description

Value (Study 1 and 2)

E50-ref (MPa) Eoed-ref (MPa) Eur-ref (MPa) c
ref (kPa)

17 and 250 17 and 250 51 and 750 20 and 300

3.1.4 Numerical prediction of vertical stiffness The results of the base case and two parametric study numerical analyses are shown on one chart presented as Figure 3.

3.2

3-Dimensional numerical analyses

3.2.1 General Numerical analyses performed was to calculate the vertical, horizontal and rotational stiffness of the isolated pile and combined pile and mudmat foundation system. These analyses were performed in the 3D (Foundation) version of Plaxis (Plaxis, 2008). 3.2.2 Finite Element model The Finite Element mesh used is presented in Figure 4. The mudmat was modelled using plate elements with the dimensions provided in Section 2.2.2. The pile was modelled using 15-node solid elements. In order to exclude end bearing from the contribution to pile vertical stiffness, a “hole” was created below the pile toe level. This modelling detail accords with the industry practice for design of

drilled and grouted piles in the North West Shelf of Australia. The 3D model was built using the internal meshing functionality in Plaxis 3D Foundation and in total there were 14,000 elements in the model. The assumed soil stratigraphy presented in Table 1 was replicated in the numerical model. The far-field boundary was placed 20 m from the edge of the mudmat, a distance of ∼2 mudmat diameters. 3.2.3 Analysis parameters The 3D analyses were performed using the base case numerical analysis parameters presented in Tables 3 and 4. 3.2.4 Numerical prediction of foundation stiffness The results of the base case 3D numerical analyses are shown in Table 8 and are used in the structural modelling (Section 4). 4 4.1

General

Structural modelling was performed in Oasys GSA (Oasys, 2009) to investigate the effects of the restraint conditions on the computed reactions and the structural period. The assumed jacket structure described in Section 2.2 was modelled in the package and all the loads presented in Section 2.2.1 were applied. Three restraint conditions were applied to the ends of the four main legs, these being (i) isolated pile, (ii) pile and mudmat and (iii) fully fixed. The spring values relevant to restraint conditions (i) and (ii) are presented in Table 8. A 3D view of the jacket structure is shown in Figure 1.

4.2 Figure 3. Vertical stiffness from 2D axisymmetric modelling.

STRUCTURAL MODELLING

Results

4.2.1 Reactions The (significant) reactions computed for both main legs making up one of the frames in the Oasys GSA model for the three alternate restraint conditions considered are summarised in Table 9. 4.2.2 Structural period The structural period computed in the Oasys GSA model for the three alternate restraint conditions are summarised in Table 10. The two sets of results presented in Table 10 are for sway of the structure in two orthogonal directions. Table 8.

Figure 4. Plaxis 3D analysis model.

© 2011 by Taylor & Francis Group, LLC

Foundation stiffness results.

Restraint

Isolated Pile

Pile and Mudmat

Horizontal (kN/mm) Vertical (kN/mm) Rotational (kNm/rad)

512 3339 4.3 × 106

3718 3832 58.3 × 106

528

Table 9.

Main leg reactions.

Restraint Condition Fx (kN)1.

5.3

Fz (kN)2.

The frame analysis performed using the spring values presented in Table 8 and also fully fixed restraints shows that the moment generated at the connection of the main jacket leg to the foundation system is dependent on the stiffness thereof. It can be seen that the greatest reaction moment was computed for the isolated pile case, with the pile and mudmat system and fully fixed cases having reductions in the peak reaction moment of 3% and 14% respectively. In addition to the above, the structural period of the jacket supported on isolated piles was a maximum of 2.28 s. The corresponding structural periods for the pile and mudmat system and fully fixed cases were 1.94 s and 1.91 s respectively, reductions of 15% and 16% over the isolated pile condition. The relevance of estimating the contribution of the mudmat to the jacket response stems from the fact that the structural period corresponds directly to the dynamic amplification factor applied to the environmental loads. A reduced structural period corresponds to lower multiplication factors and therefore smaller environmental loads for use in structural and stability analyses.

Myy (kNm)3.

Isolated +3282 −6175 17760 25000 −6342 −236 Pile Pile and +3666 −6554 17850 24920 −6186 −2867 Mudmat Fully +3796 −6682 17860 24910 −5441 −3688 Fixed Notes: 1. Horizontal reactions in direction of lateral loading 2. Vertical reactions 3. Moment reactions in plane of lateral loading

Table 10.

Structural period.

Restraint Condition

Period (s) x-dirn y-dirn

% change x-dirn y-dirn

Isolated Pile Pile and Mudmat Fully Fixed

1.31 1.13 1.11

– 14 16

5

2.28 1.94 1.91

– 15 16

6

DISCUSSION OF RESULTS

5.1 Vertical stiffness analyses The vertical stiffness results presented in Figure 3 show that, as expected, an increase in the cohesion (and the associated deformation modulus) of the cemented calcareous sand yields an increase in the axial stiffness of the three foundation systems analysed. It is however noted that the axial stiffness of the isolated pile does not change significantly with increasing calcareous sand modulus, due primarily to the dominance of the rock socket in controlling the load-settlement response of the pile. The isolated mudmat on the other hand experiences a large increase in axial stiffness over the range of soil cohesion (and hence deformation modulus) values adopted, becoming 17 times stiffer. The isolated pile became only 1.5 times stiffer over the same range. Furthermore, the axial stiffness of the combined pile and mudmat foundation stiffness (Figure 3) shows that the axial stiffness of the isolated pile and mudmat elements may be summed to provide an estimate of the composite foundation behaviour, although the stiffness values adopted in the structural modelling used values generated from the 3D model.

CONCLUSIONS

The Plaxis 2D, Plaxis 3D Foundation and Oasys GSA modelling performed has shown that the inclusion of mudmats in the analysis of a piled foundation system’s load-settlement response (in appropriate ground conditions) reduces both the computed moment reactions and the amount of dynamic amplification required to be applied to environmental loads. These reductions may lead to economies in the design of the structural elements forming the jacket structure and any connecting elements to the foundation system. The encouraging results presented above suggest further research is required for real offshore platforms. ACKNOWLEDGEMENTS The contribution of Arup’s Design and Technical Fund for providing the funding to complete the work is gratefully acknowledged.

5.2 Generalised stiffness analyses The generalised stiffness analyses results calculated using a 3D numerical model presented inTable 9 shows good general agreement with the vertical stiffness values presented in Figure 3 (for base case parameters) as assessed from 2D numerical modelling.

© 2011 by Taylor & Francis Group, LLC

Frame analysis

529

REFERENCES Oasys. 2009 GSA User’s Manual Plaxis. 2008 Plaxis 2D User’s Manual Plaxis. 2007 Plaxis 3D Foundation User’s Manual Ismail, M.A., Sharma, S.S. & Fahey, M., 2005. Detection of slight cementation in offshore carbonate deposits from laboratory testing. International Symposium Frontiers in Offshore Geotechnics, Gourvenec and Cassidy (eds), pp 1033–1037.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Simplified numerical model for analysis of offshore piles under cyclic lateral loading M.M. Memarpour Iran University of Science and Technology, Tehran, Iran

M. Kimiaei Centre for Offshore Foundation systems, University of Western Australia, Perth

M. Shayanfar Iran University of Science and Technology, Tehran, Iran

ABSTRACT: Offshore pile foundations are often subjected to lateral cyclic loads due to environmental forces applied to the supported structure. Pile foundation analysis commonly includes the use of a Beam on Nonlinear Winkler Foundation (BNWF) model in which the soil reaction is related to lateral deflection through prescribed curves however this model is very sensitive to the way one considers the p-y curves for the analysis. This paper outlines a new robust BNWF model, considering nonlinear cyclic pile-soil interaction behaviour, based on the API recommended p-y cyclic curves. This model is developed using ABAQUS software and it can be easily implemented in comprehensive models for ultimate strength analysis of fixed offshore platforms. In addition to soil elasto-plastic behaviour, this model is able to capture soil strength degradation, gapping phenomenon and drag forces on the offshore piles under cyclic loadings. In a series of numerical simulations, results of this model are compared well with existing models in cyclic behaviour of offshore piles.

1

INTRODUCTION

Pile supported offshore and coastal structures in marine soil deposits are subjected to large static and cyclic lateral loads. Usually, critical lateral forces on the piles used in coastal structures are due to berthing and mooring forces, whereas in offshore jacket platforms, piles are subjected to the cyclic lateral loads due to waves. Nonlinear pile-soil interaction (PSI) is one of the main parameters which can deeply affect the overall response of the supported structures. Finite element and boundary element methods are direct numerical approaches for solving the pile-soil interaction problems in which soil, pile and the pilesoil-interfaces are modeled all together in one integrated model. For example, Trochanis et al. (1988) used finite element method whereas Kaynia & Kausel (1982) implemented boundary element method for response analysis of piles. These methods treat the soil as a continuum medium and provide powerful tools for conducting soil-pile-structure interaction analyses. Main advantage of such approaches is the capability of performing the pile-soil-interaction analysis in a fully coupled manner. However, these two methods are not commonly used in design offices mainly due to their presumed excessive computational costs and their complexity for common pile dynamic response analysis (El Naggar et al. 2005). © 2011 by Taylor & Francis Group, LLC

Beam on Nonlinear Winkler Foundation (BNWF) method is a simplified approach that can account for nonlinear PSI and is commonly used in professional geotechnical engineering and research practices as well. In the BNWF method, pile is modeled as a series of discrete beam-column elements resting on a series of springs and dashpots representing the stiffness and the energy dissipation through the soil layers (Kimiaei et al. 2004). Generally, BNWF models are efficient and reasonably precise methods that can account for various complicated conditions in a simple manner. For example, Novak et al. (1978) and El Naggar & Bentley (2000) have used BNWF models for the piles subjected to lateral dynamic loads. Some BNWF models use the p-y curves approach (unit load transfer curves) in which the soil lateral stiffness is represented using nonlinear p-y (soil resistance versus pile lateral deflection) curves. Boulanger et al. (1999) and Kimiaei et al. (2004) used the BNWF models based on the p-y curve approach for seismic soil pile structure interaction problems. The nonlinear p-y curves can be used in BNWF models for showing nonlinear lateral, axial and end bearing behavior of piles in soil layers. However results of such BNWF models are very sensitive how these nonlinear p-y curves are implemented in the models. For lateral behavior of piles, different p-y curves have been proposed for soft clay (Matlock 1970), stiff

531

Figure 2. Static and Cyclic p-y curves as per API recommendations (API 2000).

plastic behavior (Fig. 1b). Third part of this model shows the soil behaviour in the far field. It is modeled by an elastic spring (showing the soil stiffness) in parallel combination with a dashpot (showing energy dissipation through the system under cyclic loadings). Combining these three components, it has been shown that the seismic soil pile interaction (incorporating gap development and drag forces on the piles) can be reasonably modelled (Boulanger et al. 1999).

Figure 1. Characteristics of Nonlinear p-y Element: (a) Components; (b) Behaviour of Components. (Boulanger et al. 1999).

clay (Reese and Welch 1975) and sand (Cox et al. 1974, Reese et al. 1974) layers. These curves are adopted by API (American Petroleum Institute) for routine use in design of foundations for offshore platforms (API 2000). In this paper a new simplified and robust BNWF model based on the Boulanger model, is presented that can be used for cyclic response analysis of laterally loaded piles. A general finite element analysis software, ABAQUS (Simulia, 2004) is used for development of this model. Effects of drag forces and soil strength degradation on cyclic response of piles are investigated in this study. 2

BNWF MODELS FOR PSI ANALYSIS

Boulanger et al. (1999) presented a BNWF model for pile soil interaction analysis which incorporates three main parts connected in series as it is shown in Figure 1a. First part in this model is a gap element which is a combination of drag and closure springs. These two springs are connected in parallel and together they model the soil behavior in the gap region around the pile. Force-displacement behavior of these springs is illustrated in Figure 1b. In the gap region, closure spring takes no load and the total soil resistance will only be provided through the drag spring. Reaching the end of the gap region, parallel combination of the drag and the closure springs represent a rigid behaviour for the gap element (infinite total stiffness). Second part in the Boulanger model is the near field element or the plastic spring. The plastic spring has an initial range of rigid behavior between −0.35Pult < P < 0.35Pult (where Pult is ultimate lateral soil capacity) and out of this range it shows a © 2011 by Taylor & Francis Group, LLC

3

MODEL DESCRIPTION

Soil hysteretic behaviour under lateral cyclic loading shows significant changes not only in soil strength but also in soil stiffness (Matlock 1970). These sequential degradations in soil characteristics can influence the pile responses in sequential loading cycles. For the laterally loaded offshore piles, API recommends two different types of static and cyclic p-y curves which are both schematically shown in Figure 2. It is seen that by increasing the pile lateral deflections, the cyclic p-y curve incorporates a notable reduction in the soil strength. The BNWF model developed in this study, namely CPSI (Cyclic Pile Soil Interaction) hereafter, is based on the concepts in the Boulanger model (Boulanger et al. 1999). Main advantage of the CPSI model comparing with the BNWF models by Boulanger et al. (1999) and Kimiaei et al. (2004) is the capability of this model for soil strength degradation under cyclic loads. Using the CPSI model, during each iteration in the nonlinear pile response analysis, ultimate resistance of the soil (Pult ) will be updated based on the total lateral deflection of the pile as per API recommendations. Typical soil hysteric curves for the Boulanger and the CPSI models comparing with the API static and cyclic p-y curves are shown in Figures 3–4 respectively. In the present study user element (UEL) feature of the general finite element program, ABAQUS (Simulia 2004), is used to incorporate the CPSI approach into finite element models for cyclic response analysis of offshore piles. The developed UEL is a robust single node element which can be switched between the static or the cyclic responses and all soil input parameters can also be easily changed by the user.

532

Table 1.

Pile model main characteristics.

Pile length Outside Diameter Pile thickness

Elasticity modulus Density Yield stress Figure 3. Static and cyclic API p-y curves and Boulanger model behaviour.

Table 2.

80 m 1371 mm 60 mm (0 to −7.5 m) 65 mm (−7.5 to −20 m) 45 mm (−20 to −28 m) 30 mm (−28 to −80 m) 210 GPa 7800 kg/m3 360 MPa

Dimensions of pile segments.

Element size

Depth below sea bed

0.5 m 1.0 m 2.0 m 5.0 m

0.0 to −20 m −20 to −35 m −35 to −55 m −55 to −80 m

Figure 4. Static and cyclic API p-y curves and CPSI model behavior.

In the ABAQUS models using this CPSI user element, pile and surrounding soil are subdivided into a number of discrete layers. The piles will be represented by two-node beams and the stiffness of each pile segment is modeled using the standard stiffness matrix of beam column elements. The single-node CPSI user element, representing the cyclic pile soil interaction, will be attached to each node on the pile elements. Nonlinear stiffness of the soil layer around the pile segment at each node will be updated based on total lateral deflection of the pile. Stiffness matrices of the pile segments and the surrounding soil layers are then assembled, through ABAQUS nonlinear procedures, to form the global structural stiffness matrix of the system. 4

CASE STUDY

Figure 5. Soil layers and CPSI model.

To investigate response of offshore piles under cyclic lateral loads, using the features of this CPSI model, a steel pile from a sample 4-legged fixed platform in the Persian Gulf is studied here. This pile has a circular cross section with 137.1 cm outside diameter and 80 m penetration below seabed. Detailed information about this pile is given in Table 1. The soil profile in this field consists of 3 horizontal layers and the pile is loaded with a constant vertical load of 14 MN and 10 cycles of sinusoidal lateral loads of 2 MN. At soil top layers, where there are higher levels of the pile lateral deflections than the other parts of the pile, a finer mesh for the pile segments is used. Moving © 2011 by Taylor & Francis Group, LLC

533

down towards the pile tip along the pile shaft, the pile deflections and the internal forces will decrease rapidly and hence a more coarse mesh can be used in this area. Table 2 presents a summary of the pile segmentations used in this study. These segments are selected reasonably small in all layers. ABAQUS software and the CPSI user elements are employed to model and analyze this pile under described lateral cyclic loadings. This model includes 70 nodes, 69 standard beam elements and 69 CPSI user elements. A General view of this pile, the CPSI structural model, the soil layers and the physical properties of the soil layers are shown in Figure 5.

Table 3. Results of pile response analysis for different drag coefficients. Drag coefficient (α)

0%

30%

50%

Deflection (m)

1st cyc. 10th cyc. difference

−0.378 −0.441 16.40%

−0.378 −0.439 16.10%

−0.378 −0.437 15.70%

Shear Force (MN) Bending Moment (MN.m)

1st cyc. 10th cyc. difference 1st cyc. 10th cyc. difference

−1.72 −1.88 9.30% 19.5 21.3 9.20%

−1.72 −1.87 8.70% 19.5 21.2 8.70%

−1.72 −1.87 8.70% 19.5 21.2 8.70%

5

NUMERICAL RESULTS & DISCUSSION

Main objective of this part of the study is to investigate the effects of the following input parameters on the lateral cyclic response of the pile:

Figure 6. Pile deflection along the pile shaft (drag coefficient of 30%).

• Drag force on the pile in the gap region • Soil strength degradation

Using the CPSI model with no soil strength degradation, the pile response analysis is carried out for a drag force of αPult (where α and Pult represent the drag coefficient and the lateral capacity of the soil at each stage). Three different α coefficients are used in these sensitivity analyses: α = 0% (no drag effect), α = 30% (used by Boulanger et al. (1999)) and α = 50% (used by Wallace et al. (2002)). Results of the peak pile deflections (lateral displacement at the seabed level), the peak shear force and the peak bending moments along the pile shaft for the first and the last loading cycles, are presented in Table 3. It is seen that the drag coefficient, α, has no important effect on any of the calculated pile responses either for the first or for the last loading cycle. It is also observed that the pile deflection, the shear force and the bending moment for all three different cases (α = 0%, 30% and 50%) for the 10th cycle are about 16%, 9% and 9% respectively higher than the results for the 1st loading cycle. The pile deflections, the shear forces and the bending moments along the pile shaft for the drag coefficient of 30% at the first and the last loading cycles are shown in Figures 6–8. In the second part of this study, effects of the soil strength degradation on overall response of the pile under the lateral cyclic loading are assessed. Results of the pile deflections, the shear forces and the bending moments for the last loading cycle (i.e. 10th cycle) along the pile shaft for drag coefficient of 30% using the CPSI model (incorporating the cyclic p-y curve) and the Boulanger model (incorporating the static p-y curve) are shown in Figures 9–11 respectively. All these results show that the soil strength degradation (using the CPSI model) will lead to significant increase in the pile responses. Peak pile responses for the CPSI and the Boulanger models are summarized in Table 4. It is seen that the peak pile deflection, the peak shear force and © 2011 by Taylor & Francis Group, LLC

Figure 7. Shear force distribution along the pile shaft (drag coefficient of 30%).

the peak bending moment for the CPSI model are approximately 137%, 57% and 59% higher than the results for the Boulanger model. It can also be observed that critical section of the pile for the shear force and the bending moment in the CSPI model is lower (about 4 m) than the Boulanger model. 6

CONCLUSIONS

A simplified and robust BNWF model for cyclic pile soil interaction analysis of offshore piles was introduced. This model was incorporated as a user element in ABAQUS software and it was used for series of response analysis of offshore piles under lateral cyclic loads. Effects of different drag coefficients and soil strength degradations on the pile responses (peak

534

Figure 10. Pile shear force for CPSI and Boulanger model along the pile shaft for the 10th cycle (drag coefficient 30%).

Figure 8. Bending moment distribution along the pile shaft (drag coefficient of 30%).

Figure 11. Pile shear force for CPSI and Boulanger model along the pile shaft for the 10th cycle (drag coefficient 30%). Figure 9. Pile deflection for cyclic (CPSI) and static model (Boulanger) along the pile shaft for the 10th cycle (drag coefficient 30%).

Table 4. Result of pile response analysis for CPSI and Boulanger models.

pile deflections, peak shear forces & peak bending moments) were studied and it was concluded that: • Drag Coefficient has no important effect on overall

pile peak response results. • Incorporating soil strength degradation (using the CPSI model) will lead to significant increase in pile responses comparing with the static p-y curves (using the Boulanger model)

Pile response Deflection (m) Shear Force (MN) Bending Moment (MN.m)

ACKNOWLEDGMENTS This work forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established © 2011 by Taylor & Francis Group, LLC

535

CPSI model

Boulanger model

Difference (%)

−0.967 −2.94 33.8

−0.408 −1.87 21.2

137 57.2 59.4

under the Australian Research Council’s Research Centres Program and now supported by the State Government of Western Australia as a Centre of Excellence.The authors would like to thank IUST (Iran University of Science and Technology) and COFS for their financial & technical supports during this study.

REFERENCES American Petroleum Institute (API) 2000. Recommended Practice for Planning, Designing, and Constructing Fixed Offshore Platforms – Working Stress Design. Report RP2A-WSD. 20th Edition. Boulanger, R., Curras, C., Kutter, B., Wilson, D. & Abghari, A. 1999. Seismic soil-pile structure interaction experiments and analyses. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 125(9): 750–759. Cox, W., Reese, L. & Grubbs, B. 1974. Field testing of laterally loaded piles in sand. Proc. 6th Offshore Technology Conference, Vol. 1, Houston, TX. OTC 2079: 459–472. El Naggar, M.H., Shayanfar, M.A., Kimiaei, M., & Aghakouchak, A.A. 2005. Simplified BNWF model for nonlinear seismic response analysis of offshore piles with nonlinear input ground motion analysis. Canadian Geotechnical Journal, 42: 365–380. El Naggar, M.H., & Bentley, K.J. 2000. Dynamic analysis for laterally loaded piles and dynamic p–y curves. Canadian Geotechnical Journal, 37: 1166–1183. Kaynia, A., & Kausel, E. 1982. Dynamic stiffness and seismic response of pile groups. Massachusetts Institute of Technology, Cambridge, Mass. Report R82–03. Kimiaei, M., Shayanfar, M.A., El Naggar, M.H., & Aghakoochak, A.A. 2004. Non linear seismic pile soil structure interaction analysis of piles in offshore platforms. Proc. 23rd International Conference on Offshore

© 2011 by Taylor & Francis Group, LLC

Mechanics and Arctic Engineering. 20–25 June 2004, Vancouver, Canada. Matlock, H. 1970. Correlations for design of laterally loaded piles in soft clay. Proc. 2nd Offshore Technology Conference, Vol. 1, Houston, TX. OTC 1204: 577–594. Nogami, T., Otani, J., Konagai, K. & Chen, H.L. 1992. Nonlinear soil-pile interaction model for dynamic lateral motion. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 118(1): 89–106. Novak, M., Nogami, T., & Abul-Ella, F. 1978. Dynamic soil reaction for plane strain case. Journal of the Engineering Mechanics Division, ASCE, 104: 953–959. Reese, L., Cox, W. & Koop, F. 1974. Analysis of laterally loaded piles in sand. Proc. 6th Offshore Technology Conference, Vol. 1, Houston, TX. OTC 2080: 473–483. Reese, L. & Welch, R. 1975. Lateral loading of deep foundations in stiff clay. Journal of the Geotechnical Engineering ony icf-sw55 Division, ASCE, 101(7): 633–649. Simulia Inc., 2004, Abaqus Analysis User’s Manual version 6.5.1, http://www.simulia.com/support/documentation. html Online Documentation. Trochanis, A., Bielak, J., & Christiano, P. 1988. A threedimensional nonlinear study of piles leading to the development of a simplified model. Department of Civil Engineering, Carnegie Institute of Technology. Report R-88–176. Wallace, J.W., Fox, P.J., Stewart, J.P., Janoyan, K., Qiu, T. & Lermitte, S.P. 2002. Cyclic large deflection testing of shaft bridges. Part II:Analytical studies. Report from California Dept. of Transportation: Los Angeles.

536

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Centrifuge modelling of rapid load tests with piles in silt and sand C.T. Nguyen Civil Engineering & Geosciences Faculty, Delft University of Technology, The Netherlands

H. van Lottum & P. Hölscher Deltares, The Netherlands

A.F. van Tol Civil Engineering & Geosciences Faculty, Delft University of Technology and Deltares, The Netherlands

ABSTRACT: Rapid and static load tests were conducted on open-end and closed-end piles in the Deltares GeoCentrifuge. On flight, a pile was driven into the soil. Both fine-grained sand and silt samples were tested. In both the rapid and static load tests, the soil resistance of a closed-end pile was higher than the soil resistance of an open-end pile in both sand and silt. The ratio for maximum soil resistance between a rapid load test and static load test does not depend on pile type but on soil type: 1.0 for sand and 1.2 for silt. The results show that plugging during a rapid load test differs from plugging during a static load test. Silt must be used for the proper scaling of the test for open-end piles in the geocentrifuge.

1

INTRODUCTION

Pile load tests are a standard procedure for the verification of pile load-displacement behavior. The methods used are the static load test (SLT), the dynamic load test (DLT) and the rapid load test (RLT). The tests vary in terms of the dimensionless wavelength Nw = (Tf × cp )/L in which Tf is the loading duration, cp is the pile wave velocity and L is the pile length. Nw < 10 for the DLT, 10 < Nw < 1000 for RLT and Nw > 1000 for the SLT (Hölscher et al., 2008). Although the SLT is the most reliable method, it is often too expensive and time consuming to apply routinely. The RLT is increasingly used because it is better in terms of execution, elaboration and quality assurance than the DLT (Middendorp et al., 1992) and is more suitable for use in offshore foundation engineering than the SLT. Open-end piles generally behave as though fully plugged during static loading but they can behave in a partially plugged way during rapid or dynamic loading, especially when loading rates are high (Bruno & Randolph, 1999). The degree of plugging depends not only on the loading rate but also on the type of soil. Different degrees of plugging are expected to result in different levels of soil resistance. An understanding of plugging during an RLT is important for the application of RLTs to open-end piles: if a pile plugs during an SLT but does not plug during an RLT, the RLT will be unreliable and may underestimate pile capacity. Scale modelling pile load tests offers a good possibility for this research. It avoids the high costs of field testing and offers additional possibilities compared © 2011 by Taylor & Francis Group, LLC

with field testing. Centrifuge modelling is considered to be a reliable method due to the accurate representation of the stress state, especially the self-weight stress gradient, around and inside the model pile at a reduced scale. An experimental study of RLTs and SLTs with open-end piles was performed with different soil types to examine plugging behaviour in silt and sand, especially during RLTs, and to compare soil resistance in rapid and static conditions. This paper presents the results from four test series comprising several RLTs and SLTs.

2 2.1

DESCRIPTION OF RESEARCH Centrifuge modelling

Given the requirement of stress similarity between the model (with the centrifuge length Lmodel and the centrifuge acceleration of amodel ) and the prototype (with the length Lprototype and the earth’s gravity aprototype ), the scale factor is defined as:

Table 1 shows the scale factors of some parameters on the basis of dimensional analysis (Taylor 2005): The experimental study was carried out in the GeoCentrifuge at Deltares (The Netherlands). Figure 1 shows the facility. It was described in detail by Huy et al. (2008).

537

2.2

Model piles

The model pile was made from steel with a length of 300 mm, a diameter of 11.3 mm (D), wall thickness of 0.5 mm and mass of 875 gram (M ); this mass includes the pile mass and the mounting gear on the pile head. Table 1.

Scale factors in centrifuge test.

A load cell was mounted on the pile head to measure the applied force. 2.3

Model materials

Baskarp sand (d50 = 130 µm) and silt (d50 = 58 µm) were chosen for the tests. Table 2 lists the basic parameters for the soils (the quoted values for friction angle

Parameters

Model

Prototype

Table 2.

Length/Displacement Acceleration Time (dynamics) Mass Velocity Force Stress Strain

1 N 1 1 1 1 1 1

N 1 N N3 1 N2 1 1

Parameters

Dimension

Sand

Silt

Grain vol. mass d50 Min. porosity Max. porosity Friction angle Permeability

kg/m3 µm % % degree m/s

2647 130 34 46.9 40◦ 12 × 10−5

2650 58 42.2 53.9 38◦ 1.5 × 10−5

Figure 1. Centrifuge test setup (Huy, 2008).

© 2011 by Taylor & Francis Group, LLC

538

Properties of soils.

and permeability are at 65% relative density) and Figure 2 shows the grain size distribution curves. To minimise the scale effects, the ratio of pile wall thickness to the mean grain size d50 needs to be larger than 10 and the ratio of the inner diameter of pipe pile to d50 must be larger than 200 (Nicola & Randolph, 1997). The silt almost satisfies this condition (8.6 and 178). In the sand, the ratios are 3.9 and 79. In prototype terms, the test with silt corresponds to the normal use of open-end piles in sea-bed sand, while the test with sand is an extreme case in a fine gravel layer which is sometimes to be found in reality. The soil sample was prepared by drizzling sand into water, followed by densification using impact loading (Rietdijk et al., 2010). This method made it possible to achieve a reasonably homogeneous and reproducible sample of 65% relative density (for these types of soils). Water was selected as the model pore fluid for all tests. It is therefore reasonable to assume drained behaviour. Based on the results of Huy (2008), the response of the pile under rapid loading will be drained, with water as the pore fluid in both cases (Baskarp sand and silt). The effects of excess pore pressure can be ignored. Furthermore, it is very difficult to saturate the silt with viscous fluid and the silt and sand can be used again easily after the tests if water is used.

in sand (Huy et al., 2008) are also shown here for the purposes of comparison. 3

RESULTS OF THE CENTRIFUGE TEST

Figure 3 shows two typical results for measured pile head force and applied pile displacement. The pile head forces have been corrected for the self-weight of the pile. During an RLT, the pile can be seen as a rigid body. In that case, the force on the pile head (Fmeasured ) is equal to the sum of the soil resistance (Fsoil ) and the inertia force (Finertia ) of the pile (Middendorp et al., 1992). The soil resistance can therefore be calculated from:

where M is the pile mass and a is the pile acceleration. The acceleration is calculated numerically as the second derivative of the measured pile displacement at all time steps.

2.4 Test programme Three tests were performed at the gravity level N = 40 with the same loading programme: two tests in silt (one with an open-end pile (OEP) and one with a closedend pile (CEP)) and one test with an OEP in sand. During the tests, the pile was first pushed from the pre-embedded depth of 10D to a depth of 20D using the large hydraulic actuator. Two RLTs were then performed with displacements of 1% D (Rapid u = 1%) and 10% D (Rapid u = 10%) respectively (duration 10 ms) and, finally, an SLT with a displacement of 10% D (Static) was performed. The results from one test conducted previously (also at Deltares) with a CEP

Figure 2. Grain size distribution curves.

© 2011 by Taylor & Francis Group, LLC

Figure 3. Measured Load-Displacement curves, OEP.

539

Figure 4 shows an example of the measured pile head force, inertia force and resulting soil resistance and prescribed pile displacement from the RLT with silt. This soil resistance still includes velocity effects due to rapid loading. The applied force can be considered rapid, even though, compared with a field test (e.g.Matsumoto & Nishimura, 1996), the generated force has very steep flanges and a long duration of maximum force. 4

and Figure 7 shows the results of the tests in sand. Part a) shows the results for the OEP and part b) the results for the CEP. The test for the CEP in sand can be found in Huy et al. (2008). Generally, the soil resistance-displacement curves of RLTs have quite similar patterns: the force first rises quickly to its maximum value, then stays high at about the maximum value before finally falling rapidly.

DESCRIPTIONS AND DISCUSSION OF THE MODEL PILE TEST RESULTS

This section describes the comparison of SLTs and RLTs in silt and sand in detail. It should be pointed out that, from this point on, the soil resistance force during the RLT will be the calculated pile head force after eliminating the inertia force of the pile, and that all the numbers and quantities are in terms of model scale (N = 40 g). 4.1 Pile installation As described above, the model piles were pushed into the soil medium with the large hydraulic actuator from the initial depth of 10D to the final depth of 20D with a driving velocity of 10 mm/min.At this very low driving speed, the installation process can be considered as static jacking. Figure 5 shows the pushing records from the installation phase. It is clear that the installation of the model pile in sand requires about 30% more force than in silt. A possible explanation is the grain size of sand, which is quite large compared to the thickness of the pile wall. 4.2

Figure 5. Load-Displacement curve for installation phase.

Soil resistance

Figures 6 and 7 show the soil resistance-displacement curves for different maximum displacement values. Since the duration of the loading was the same in all tests, the loading speed also varies between these tests. Figure 6 shows the results of the tests in silt

Figure 4. Example of measured and calculated signals.

© 2011 by Taylor & Francis Group, LLC

Figure 6. Load-Displacement curve for pile in silt.

540

This loading pattern deviates from the loading pattern observed in field tests with a shallower increase to the maximum load and a shallower decrease to zero. This is a limitation of the hydraulic loading system, as seen in Figure 4. With the sand sample, the maximum soil resistance during the RLT is comparable with the maximum soil resistance during the SLT (RM = 1); with the silt sample, the maximum soil resistance during the RLT is 20% higher than the maximum soil resistance during the SLT (RM = 1.2). These differences apply to both the closed-end and open-end piles.

The maximum soil resistance of the closed-end pile is higher than the maximum soil resistance of the openend pile in both the RLT and SLT: about 30% for the sand sample and 10% for the silt sample. The ratios of soil resistance at the unloading point during the RLTs to maximum soil resistance during the SLTs (RUP in Table 3) were quite different in all tests. For the CEP in sand, the ratio was 0.9 while, in the other tests, the ratio was significantly less than 1. Figures 6 and 7 suggest that the unloading point method is not altogether appropriate for the RLTs in this study. This may be due to the steep loading pattern or the high inertia forces during these RLTs. The soil resistance observed during the SLTs in sand was higher than in silt: soil resistance with the OEP was 1.5 times higher; a factor 2 was found for the CEP. These differences could possibly be explained by the properties of the soil materials. Firstly, the friction angle of Baskarp sand is 1–2◦ higher than the friction angle of silt (at a relative density of 65%). Secondly, the d50 of the sand is 2.5 times larger than the d50 of the silt. The d50 governs the thickness of the shear band along the pile shaft, at the outer surface for the CEP pile and at the outer and inner surface for the OEP pile (Wolf et al., 2003; Wood, 2002), and at the pile tip. 4.3

Stiffness

Figure 7 also shows clearly that the stiffness of rapid loading is higher than that of static loading. This concurs with the numerical results for dynamically loaded piles in saturated soil of Hölscher and Barends (1992). 4.4

Figure 7. Load-Displacement curve for pile in sand. Table 3.

Plugging

After installation and all loading phases, the pile was dug out. The final plugging length of the soil inside the model piles was 55 mm (5D) with silt and 22 mm (2D) with sand. The total displacement of each pile was 122 mm (10.8D), with the total embedded length of each pile being 241 mm (20.8D). Plugging length as a percentage of the total embedded length of pile was about 23% for silt and 9% for sand. These are relatively extreme values for plugging length when compared to those generally observed in reality (10–20% of the embedded length of the pile) (Randolph et al., 1991). Since the measured plugging length is highly dependent on the material, it is important to use a correctly scaled material; in this case of N = 40, silt must be used.

Soil resistance in RLT and SLT at displacement of 10% D. Rapid Load

Test

Max Load [kN]

UP Load [kN]

Static Load [kN]

RM =

Closed-end Sand Open-end Sand Closed-end Silt Open-end Silt

1.33 1.00 0.81 0.73

1.21 0.38 0.19 0.26

1.35 0.95 0.66 0.60

0.99 1.05 1.22 1.21

© 2011 by Taylor & Francis Group, LLC

541

FMax Load FStatic

RUP = 0.90 0.40 0.28 0.43

FUP Load FStatic

A close inspection of Figure 6 shows that the SLTs for the OEP and the CEP are almost identical. The RLTs for all piles show that the force declines after reaching the maximum. With the OEP, the force decreases slightly more than for the CEP and is slightly more perturbed. The soil column inside the pile may have slipped during the RLTs. However, the differences are small and the soil resistance of the open-end pile was quite comparable to the soil resistance of the closed-end pile. This suggests that the piles plug during both SLTs and RLTs. The motion of the plug would have to be measured directly to obtain more accurate information. 5

CONCLUSIONS

This paper described experimental work investigating soil plugs in open-end pipe piles in a geotechnical centrifuge. Both static and rapid load tests were studied in two types of soil: fine-grained sand and silt. The results of the model tests show that centrifuge testing is a feasible and efficient approach to studying the behaviour of open-end piles. The conclusions can summarised as follows: 1. The plugging length in sand is about 2.5 times less than the plugging length in silt. 2. The soil resistance of a closed-end pile is about 30% higher in sand and 10% higher in silt than the soil resistance of an open-end pile in both RLTs and SLTs. 3. The ratio of the maximum soil resistance in the RLTs to that in the SLTs depends on the soil type: 1.0 for sand and 1.2 for silt. 4. In the test in silt, the soil resistance at the unloading point in the RLTs seems to be unrelated to the soil resistance in the SLTs. The reason for this is not fully understood but it may be due to the steep loading pattern, which deviates from the smooth pattern assumed in the unloading point method, or due to the high inertia forces. 5. Rapid stiffness is significantly higher than static stiffness. 6. The proper scaling of an open-end pile requires proper scaling of the grain size. Silt must be used for a 1:40 scale. The research is still ongoing. To improve out understanding of plugging behaviour and the impact of plugging on open-end pile capacity during RLTs, the preliminary tests can be improved by:

2. Varying the loading rate to investigate its impact on both plugging and pile capacity. REFERENCES Bruno, D. & Randolph, M.F. 1999. Dynamic and static load testing of model piles driven into dense sand. Journal of Geotechnical and Geoenvironmental Engineering 125 (11) p. 988–998 De Nicola, A. and Randolph, M.F. 1997. The plugging behavior of driven and jacked piles in sand. Geotechnique 47 (4) p. 841–856 De Nicola, A. and Randolph, M.F. 1999. Centrifuge modeling of pipe piles in sand under axial loads. Geotecnique 49 (3) p. 295–318 Hölscher, P. & Barends, F.B.J. 1992. The relation between soil-parameters and one-dimensional toe-model. Proc. 4th Int. Conf. Application of Stress Wave Theory to Piles p. 413–419 Huy, N.Q., van Tol, A.F. and Holscher, P. 2008. Rapid model pile load tests in the geotechnical centrifuge. Rapid Load Testing on Piles. p. 103–127 Huy, N.Q. 2008. Rapid load testing of pile in sand. PhD thesis, Delft University of Technology Matsumoto, T. and Nishimura, S. 1996. Wave propagation phenomena in statnamic test of a steel pipe pile. Proc. 5th Int. Conf. Application of Stress Wave Theory to Piles p. 1015–1030 Middendorp, P., Bermingham, P. & Kuiper, B. 1992. Statnamic load testing of foundation piles. Proc. 4th Int. Conf. Application of Stress Wave Theory to Piles p. 581–588 Paik, K., Lee, J., Salgado, R. and Kim, B. 2003. Behavior of open- and close-ended piles driven into sand. Journal of Geotechnical and Geoenvironmental Engineering 129 (4) p. 296–306 Randolph, M.F., Leong, E.C. and Houlsby, G.T. 1991. One-dimensional analysis of soil plugs in pipe piles. Geotechnique 41 (4) p. 587–598 Randolph, M.F., May, M., Leong, E.C., Hyden, A.M. and Murff, J.D. 1992. Soil plug response in open-ended pipe piles. Journal of Geotechnical Engineering 118 (5) p. 743–759 Rietdijk, J., Schenkeveld, F.M., Scahminée, P.E.L. and Bezuijen, A. 2010. The drizzle method for sand sample preparation. Accepted in Proceedings of the International Conference on Physical Modelling in Geotechnics. Taylor, R.N. 2005. Centrifuges in modeling: principles and scale effects. Geotechnical Centrifuge Technology, Blackie Academic & Professional p. 20–34 Wolf, H., Konig, D. and Triantafyllidis, T. 2003. Experimental investigation of shear band patterns in granular material. Journal of Structural Geology 25 p. 1229–1240 Wood, D.M. 2002. Some observations of volumetric instabilities in soils. International Journal of Solids and Structures 39 p. 3429–3449

1. Measuring the plugging length during installation and all successive static and rapid loading steps;

© 2011 by Taylor & Francis Group, LLC

542

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Field measurements on monopile Dolphins A. Sadeghi-Hokmabadi & A. Fakher School of Civil Engineering, University of Tehran, Iran

ABSTRACT: A monopile dolphin includes a single pile used for berthing and anchoring of large vessels in offshore or onshore terminals. Several methods have been developed to analyze piles under lateral loading. One of the most effective methods is the Strain Wedge Model (SWM) which has a number of advantages in comparison with traditional p-y curves. In the Pars Special Economic Energy Zone (Asalouyeh) in the south of Iran, a number of single piles as dolphins were constructed and some full-scale lateral loading tests were conducted on them under the supervision of the second author. In the present paper, a program called Lateral Analysis of Piles (LAP), which has been developed by the authors, is used to examine the Strain Wedge Model for pile analysis using the results of these full-scale loading tests. The research shows that the SWM calculates a greater pile head displacement than the test data, and illustrates the need for local calibration.

1

INTRODUCTION

In general, there are two types of berth structure; quay and jetty. A quay (or wharf) is a landing place parallel to a navigable waterway that provides access to ships and boats (Figure 1.a). Because of its high lateral resistance, the fenders must be well-designed to absorb the berthing energy of a ship. A jetty (or pier) extends out into the water from the shore. It is in the perpendicular direction to the shoreline serving as a landing place and where loading equipment allows the

use of a lighter structure. Ships can berth directly at the structure, but usually require separate structures, such as dolphins, to absorb the high energy of the ship (Figure 1.b). In some cases, dolphins consist of a number of piles. This type has low lateral deformation and, therefore, a reduced ability to absorb energy. A monopile comprises a single large-diameter pile which is embedded in the soil and behaves as a console. The ability of monopiles to absorb a high amount of energy, their low cost, and simple construction method has made

Figure 1. a) Schematic picture of a quay, b) Schematic picture of a Jetty with two berthing dolphins in middle and four mooring dolphins in sides, c) construction of Jetty with its Monopiles in Asalouyeh.

© 2011 by Taylor & Francis Group, LLC

543

them common alternatives for offshore structures such as wind turbines and mooring or berthing dolphins (Quinn 1972). In the analysis of monopiles, lateral behavior is important and the interaction between the pile and soil should be modeled accurately. A number of researchers have investigated laterally loaded pile behavior, providing a number of different approaches. These methods can be classified in to the following categories: (a) Continuum-base approaches; (b) Load-transfer (or subgrade reaction) approaches. In the first category, the soil has been modeled as a continuum media, requiring several soil properties inputs for analysis (Fleming et al. 1992). The complexity and unavailability of soil properties of this first approach make it less attractive. The load-transfer approach is more commonly used and was selected for this study. The load-transfer method models the pile as an elastic member and the soil as series of nonlinear springs (p-y curves). The nonlinear soil springs describe the local variation of lateral soil–pile interacting resistance with lateral displacement. Traditional p-y models were initially developed by Matlock (1970) and Reese et al. (1974). Later, a number of p-y curves were developed by different researchers (like Murchinson & O’Neill (1984) & Scott (1980)). Traditional p-y curves do not consider pile properties such as pile bending stiffness, pile cross-sectional shape, pile head restraint, and pile installation method (Ashour et al. 2004). SWM is an advanced method in comparison with traditional p-y curves. It can consider three-dimensional behavior of soil, the effect of piles dimension and shape, and the piles head conditions. However, SWM, as like as traditional p-y curves, is a semi-empirical method. In the other words, the main drawback to these approaches is that they are based on empirical parameters (i.e. the modulus of subgrade reaction) which can only be back figured from the results of pile load tests (Basile 2003). The aim of this study is to assess the accuracy of SWM by using the results of some full-scale tests in the Pars special economic energy zone area (Asalouyeh) in Iran. In the present paper, at first the characteristics of SWM are briefly discussed. Details and the results of the undertaken full-scale tests are shown and specs of developed computer program (LAP) has been describes. Later on, the tested monopiles are analyzed with LAP and the results are compared with the tests’ data and general conclusions are made.

2

STRAIN WEDGE MODEL

The Strain Wedge Model (SWM) is an approach that has been developed to predict the response of a flexible pile under lateral loading (Norris 1986). In the Strain Wedge Model (SWM), the soil resistance against the lateral loading is determined by the three-dimensional © 2011 by Taylor & Francis Group, LLC

Figure 2. Basic Strain Wedge in Uniform Soil (Ashour et al. 1998).

Figure 3. Soil-Pile interaction in Multisublayer Strain Wedge Model (Ashour et al. 1998).

passive wedge of soil that develops in front of the pile (Figure 2). As shown in Figure 2, this passive wedge is characterized by base angles, θm and βm , the current passive wedge depth h, and the spread of the fan angle, ϕm (the mobilized friction angle). The horizontal stress changes at the passive wedge face, σh , and the side shear τ, act. Indeed, SWM allows the assessment of the nonlinear p-y curve response of a laterally loaded pile based on the envisioned relationship between the three-dimensional responses of a flexible pile in the soil to its one-dimensional beam on elastic foundation parameters (Ashour et al. 1998) as in Figure 3. The main objective behind the development of the SWM is to solve the beam on elastic foundation (BEF) problem of a laterally loaded pile based on the envisioned soil-pile interaction and its dependence on both soil and pile properties. Compared to other approaches, the SWM depends on well known on accepted principles of soil mechanics (the stressstrain-strength relationship) and an effective stress soil analysis. For more information about SWM refer to Ashour et al. (1998 & 2004). This method is used in the present research to analyze the full-scale tested monopiles.

544

3

UNDERTAKEN FULL-SCALE TESTS ON MONOPILES

Table 1.

Experimental researches conducted on the behavior of laterally loaded piles could be divided into two basic types, namely full-scale and small-scale or model testing. Full-scale tests are generally believed to provide the most accurate results, but they are rare because of the large costs required and difficulties involved. Therefore, the results of full scale tests are valuable. In the presented research, a number of full scale tests were performed on monopile dolphins.

Section

Monopile No. 2 1 ST52 2 ST52 3 ST60 4 ST70 5 ST70 6 ST70 Monopile No. 3 1 ST52 2 ST60 3 ST70 4 ST70 5 ST70 6 ST60 Monopile No. 4 1 ST52 2 ST60 3 ST70 4 ST70 5 ST70 6 ST70 7 ST70

Asalouyeh is located in southern Iran on the Persian Gulf. It is 300 km east of the city of Bushehr on the coast of Iran. Pars Petrochemical Port in Asalouyeh has 15 berths. At piers 5 and 15, monopiles are used as berthing and anchoring dolphins (Fig. 1.c). Four monopiles were tested. Monopiles No. 1 and 2 are the inner and outer piles of Berth 15 at a water depth of 14 m. Monopiles No. 3 and 4 are the inner and outer piles of Berth 5 at a water depth of 26 m. The final elevation of the monopile heads after installation was 5 m above mean sea level. These monopiles have a cylindrical shape and were made from three types of steel. The thickness and types of steel used are variable in depth and are shown in Table 1. Details of the monopiles are shown in Figure 4. The soil parameters in the field were obtained for each layer using borings. Because of the high soil stiffness, it was not possible to perform in-situ tests such as the standard penetration test. The geotechnical properties of the soil are shown in Table 2. These parameters were obtained by describing the disturbed samples and laboratory tests. For instance the internal friction angle is determined from laboratory shear box. Table 2 presents the drained density (γd ), wet density (γt ), estimated value of standard penetration test (Nspt ), effective cohesion (C
), internal friction angle in degrees (ϕ
) and undrained cohesion (C u ).

Outer diameter (m)

Thickness (mm)

Yielding stress (kN/m2 )

1.778 1.778 1.778 1.778 1.778

25.40 25.40 28.58 31.75 34.93

360000 420000 490000 490000 490000

1.905 1.905 1.905 1.905 1.905 1.905

25.40 28.58 28.58 34.93 41.28 44.45

360000 360000 420000 490000 490000 490000

1.778 1.778 1.778 1.778 1.778 1.778

25.40 25.40 28.58 31.75 34.93 34.93

360000 420000 490000 490000 490000 420000

1.905 1.905 1.905 1.905 1.905 1.905 1.905

25.40 25.40 28.58 31.75 34.93 44.45 41.28

360000 420000 490000 490000 490000 490000 490000

applied load and may be disregarded. Also, since the spacing between the piles (21.5 m) is more than eight times the diameter of the piles, there is no pile group effect (Fleming et al. 1992). Four monopiles were tested under lateral static loading. Monopiles No. 1 and 2 were loaded in five steps. Monopiles No. 3 and 4 were loaded in three steps to accommodate the displacement limitation of the jacking system. At each step, the displacement of each pile was measured using Total Station. The loading steps increased and, for each step, the load was applied for 15 min for small loads and 30 min for large loads. Figure 5 shows the results. As it mentioned, the loads are applied at the head of the monopile dolphins and the displacement is measured in their head as well.

3.2 Tests method and results

© 2011 by Taylor & Francis Group, LLC

Type of steel

Monopile No. 1 1 ST52 2 ST60 3 ST70 4 ST70 5 ST70

3.1 Tests location

A heavy duty tension system was designed and constructed that uses a hydraulic jack to provide force and a cable to transfer tension force from one monopile to another. The testing followed ASTM D3966-81, item 24 (ASTM 1995). The tension system sat on one monopile and pulled the other one. Bolts placed in the head of the monopiles for a quick release system were used for the temporary installation of the tension system on one monopile and a pulley on the other. Cables were installed between the tension system on one monopile and support on the other with a 56 in diameter pipe between them to support the weight of the cables and avoid any initial force from them. This pipe is allowed to have axial displacement. Analysis shows that the maximum friction between the cable and pipe was less than 3% of the

Details of monopile sections.

4

LATERAL ANALYSIS OF PILES (LAP) PROGRAM

A program was developed to analyze the monopiles. The Lateral Analysis of Piles (LAP) program was written in FORTRAN programming language to solve the governing equation for a beam on an elastic foundation (Equation 1) by Hetenyi (1946),

545

Figure 4. Details of monopiles in Asalouyeh. Table 2.

Geotechnical properties of the soil in the field.

Layer description

Sand

Sand and gravel

Sand stone

Depth (m) classification γd (ton/m3 ) γt (ton/m3 ) NSPT C
(ton/m2 ) ϕ
(◦ ) Cu (ton/m2 )

0.0–8.0 SP 1.7 2 >50 0 38 0

8.0–21.0 GP 1.95 2.1 >50 0 40 0

21.0–30.0 — 1.8 2.1 — — — —

where EI = bending stiffness of the pile; Px = axial load on the pile; y = lateral deflection of the pile at point x along the length of the pile; and Es = soil subgrade reaction (spring stiffness). LAP uses the finite difference method proposed by Matlock and Reese (1961) to solve Equation 1. It considers four sets of boundary conditions at the top of the pile, such as free-head or fixed-head pile. Also, LAP can use different types of spring stiffness (Sadeghi-Hokmabadi et al. 2009) like linear springs, Non-linear p-y curves, and SWM. In addition, LAP can assess pile group behavior under lateral and dynamic lateral loading such as earthquake loads (Sadeghi-Hokmabadi 2009). In the © 2011 by Taylor & Francis Group, LLC

546

Figure 5. Results of lateral loading tests on monopile dolphins.

present research, it was used as a means to analyze the dolphins at Asalouyeh and examine the accuracy of SWM. 5

COMPARISON AND DISCUSION

In the present research, LAP is used to analyze the tested monopiles using SWM. In addition, the mentioned monopiles are analyzed using COM624

(Resse & Sullivan 1980) as well. COM624 is a program for analyzing single piles under lateral loads and uses the p-y curves suggested by Reese et al. (1974). Figure 6 presents the results of analyses for the four monopiles in the term of head displacement versus lateral load at the head of monopiles. As Figure 6 shows, both the SWM and COM624 calculate a greater pile head deflection than the measured data. In comparison, the SWM gives closer answers with measurements undertaken in the presented case study than COM624. The SWM receives force at the pile head as an input and gives the pile head displacement as output (Ashour 1998). This method calculates p-y curves during the computation. In other words, the SWM does not use pre-defined p-y curves like the traditional p-y method (Fakher et al. 2009), and it is not possible to define a certain modification factor for this method like the p-y method. The average of ratio between the calculated pile head displacement and the observed one for monopiles number 1 to 4 is calculated as 0.82, 0.85, 0.87, and 0.98 respectively. Also, the total average of this ratio for these 4 set of monopiles is 0.88. It means that the data of performed tests are 12 percent less than predicted pile head displacement using the SWM. The real behaviour of pile head displacement is nonlinear. The p-y curves and SWM have difference with real situation in the flexure of pile head-displacement curves. Indeed, the total lateral stiffness in the real situation declines sooner, but in these methods it decline later and have approximately linear behaviour in the tests loads. The difference between proposed p-y curves and SWM with the real situation is occurred because of the development of plastic region near the soil surface. In fact, in the real situation under the testing loads the near surface soil has a plastic manner and yields, but p-y curves and SWM do not show this behaviour under the tests loads level. It should be noted that the total behaviour of pile-spring system is very sensitive for the near surface soils, and these soils should be modeled carefully (Fakher et al. 2009).

6

CONCLUSION

The results of full-scale tests on large diameter piles showed that the monopile dolphins behave like long piles. The LAP program was developed to analyze piles under lateral loading. This program has the ability to consider different boundary conditions and types of spring stiffness like p-y curves and Strain Wedge Model. According to the results of full-scale in-situ test, the accuracy of Strain Wedge Model has been investigated. In granular marine soils, the traditional p-y curves and SWM calculate the pile head displacement as being greater than the test data from the present study. This means that real piles withstood large amounts © 2011 by Taylor & Francis Group, LLC

547

Figure 6. Pile head horizontal displacement versus lateral load according to test data and LAP analysis for monopiles No.1, 2, 3, and 4.

of force for the specified displacements. Thus, using these curves without calibration leads to overestimating the piles displacements and demonstrates the need for local calibration.

In comparison, the SWM gives the closest answers to the measurements undertaken in the presented case study than COM624. The shape of the pile head displacement under real conditions declines sooner than in the calculated results because the analytical models do not show the soil plasticity near ground depth. To modify, the ultimate resistance of the non-linear springs should be decreased and the primary stiffness should be increased. REFERENCES Ashour, M., Norris, G. & Pilling, P. 1998. Lateral loading of a pile in layered soil using the strain wedge model. Journal of Geotechnical and Geoenviromental Engineering, Vol. 124, No.4. Ashour, M., Pilling, P., & Norris, G. 2004. Lateral Behavior of Pile Groups in Layered Soils. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 130 (6), pp. 580–592. Basile, F. 2003. Analysis and design of pile groups, In Numerical Analysis and Modeling in Geomechanics (eds J. W. Bull). Spon Press, London, Chapter 10, 278–315. Fakher A., Sadeghi-Hokmabadi A., & Saeedi-Azizkandi A. 2009. Assessment of lateral load-transfer methods of piles by full scale in-situ tests. Proceeding of the 2nd International Conference on New Developments in Soil Mechanics and Geotechnical Engineering, Nicosia, Cyprus; pp. 230–238. Fleming, W.G.K., Weltman, A.J., Randolph, M.F., & Elson, W.K. 1992. Piling Engineering. 2nd Edition, Blackie Academic & Professional, Glasgow, UK.

© 2011 by Taylor & Francis Group, LLC

548

Hetenyi, M. 1946. Beams on elastic foundation. The University of Michigan Press, Ann Arbor. Murchison, J. M. & O’Neill, M. W. 1984. Evaluation of p-y relationships in cohesion less soil. Analysis and design of pile foundations, ASCE, New York, 174–191. Matlock, H. 1970. Correlations for design of laterally loaded piles in soft clay. Proc., 2nd Annual Offshore Technol. Conference, Houston, Texas. Norris, G.M. 1986. Theoretically based BEF Laterally Loaded Pile Analysis. Proceedings, Third International Conference on Numerical Methods in Offshore Piling, Nantes, France, pp. 361–386. Quinn, A.D. 1972. Design and construction of ports and marine structures. McGraw-Hill Inc., USA. Reese, L.C. and Cox, W.R., & Koop, F.D. 1974. Analysis of laterally loaded piles in sand. 6th Annual Offshore Technology Conference, Austin Texas, 2(OTC2080): 473–485. Reese, L.C., & Sullivan, W.R. 1980. Documentation of computer program COM624 parts 1 and 2: analysis of stresses and deflections for laterally loaded piles including generation of p-y curves. Geotech. Eng. Ctr., Bureau of Eng. Res., Uni. of Texas, Austin. Texas. Sadeghi-Hokmabadi, A. 2009. Development of a computer program for the analysis of single piles and pile groups under lateral loads. Post-graduate research thesis, University of Tehran, Iran. Sadeghi-Hokmabadi, A., Seyfi, H. & Fakher, A. 2009. Analysis of single piles under lateral loading using the Strain Wedge Model. 8th International Congress of Civil Engineering (8ICCE), Shiraz University, Shiraz, Iran (In Farsi Language). Scott, R.F. 1980. Analysis of centrifuge pile tests: Simulation of pile driving. Research Rep. OSAPR Project 13, American Petroleum Institute, Washington, DC.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Behaviour of driven tubular steel piles in calcarenite for a marine jetty in Fujairah, United Arab Emirates J. Thomas WorleyParsons Services Pty Ltd, Perth, Western Australia

M. van den Berg Delta Marine Consultants, The Netherlands

F. Chow WorleyParsons Services Pty Ltd, Perth, Western Australia

N. Maas Vopak Horizon Fujairah Ltd, Fujairah, United Arab Emirates

ABSTRACT: Vopak Horizon Fujairah Limited is the biggest oil storage and bunkering company in the Emirate of Fujairah with a sizeable terminal that has recently undergone its fifth expansion. This includes an extension to the existing marine jetty with the provision of two additional berths in 15m water depth for vessels with a capacity up to 110,000 DWT. The jetty structures are founded on tubular steel piles driven into shallow variably cemented very weak to weak sedimentary rocks with varying calcium carbonate content. A static uplift load test to failure followed by a static compression load and numerous dynamic End of Initial Drives (EoID) and restrike tests with signal matching were carried out for a better understanding of the pile behaviour. This paper describes the pile tests conducted, results obtained and their interpretations. The suitability of the API RP2A method for clays in estimating capacity of piles driven into the calcarenite at this site is investigated. The pile response from the static load test due to reversals in loading directions is also discussed.

1

INTRODUCTION

2

The jetty structures at Vopak Horizon Fujairah’s terminal are founded on tubular steel piles driven into variably cemented very weak to weak sedimentary rocks with varying calcium carbonate content. To verify pile design, a static uplift load test to failure followed by a static compression load test was carried out. During production piling, numerous dynamic tests with signal matching were carried out to verify the axial pile capacities. The static load test was carried out on a sacrificial pile. The pile was driven to a penetration of 18.41 m from the seabed level. An average shaft friction of 81 kPa was estimated from the uplift load test. During production piling, dynamic EoID and restrike test were carried out. VHFL asked WorleyParsons to review the dynamic test results and confirm that the tests were carried out in accordance to generally accepted quality standards. As part of the review WorleyParsons also assessed the pile capacity based on the API RP2A WSD (2007) clay recommendations, utilising borehole specific design UCS profiles. © 2011 by Taylor & Francis Group, LLC

PROJECT DESCRIPTION

The jetty expansion is part of the Phase V expansion project of VHFL in the Emirate of Fujairah in the United Arab Emirates. For this project DMC was designer for the Civil Marine works. The new expansion involves two new berths in the shape of a finger pier. The new berths are required to increase the ship handling capacity of the terminal due to the increase of the storage capacity on land. The berths are designed for vessels with a capacity from 5,000 to 110,000 DWT and are situated in 15 m water depth in an area sensitive to seismic activities (Zone 2b according to the Uniform Building Code). Construction started in September 2007 and was completed in December 2009. The finger pier consists of: • An access trestle with a length of approximately 300 m. • Two loading platforms • Eight breasting dolphins • Six mooring dolphins • Interconnecting walkways in between the dolphins.

549

Figure 1. Overview of new berths. Table 1.

Pile working loads (kN).

Loading

Min

Max

Tension Compression

0 400

1500 3000 Figure 2. Carbonate content.

In total 218 piles have been driven. Rock coring was performed using PQ double tube core barrel producing a nominal diameter hole of 121 mm and a nominal core diameter of 84 mm. Standard Penetration Tests (SPTs) in accordance with BS 1377 (1990) were conducted in uncemented to slightly cemented soils. Laboratory tests including particle size distribution, Uniaxial Compressive Strength (UCS) and carbonate content were carried out on recovered SPT and core samples.

• 200 OD 914 mm 19 mm WT • 18 OD 1016 mm 22 mm and 25 mm WT

The structural design is carried out in accordance to British Standards. The design of the foundation is executed based on working loads using an overall factor of safety on the pile bearing capacity. All the piles driven have been subjected to dynamic testing. For this reason the factors of safety for the final assessment after driving have been taken as 2.0 for tension and 1.75 for compression. The range of working loads for the piles is given in Table 1 above. 3

SITE GEOLOGY

The jetty site is located in Fujairah, situated along the east coast of the United Arab Emirates along the Gulf of Oman. The near-surface geology of the coastal Fujairah region is dominated by alluvial “wadi” gravel plains. The alluvial deposits are generally derived from predominantly ophiolitic, igneous rock forming the nearby Hajjar Mountain Range to the west.The alluvial gravel deposits typically extend to the coastline, where intercalation and reworking with carbonate marine sediments is commonly evident. 4

4.1

The soil and rock were classified in accordance to the Clark and Walker Carbonate Sediment Classification System. The total carbonate content was estimated in accordance with the rapid titration method of BS 1377 assuming all of the carbonate present was calcium carbonate. Figure 2 shows the carbonate content profile at the site. The rock materials encountered were generally logged as siliceous calcarenite to a depth of −27 m CD and underlain by calcareous sandstone to the remaining investigated depth. 4.2

© 2011 by Taylor & Francis Group, LLC

Uniaxial compressive strength

UCS tests were carried out on selected rock core samples in accordance with the methods outlined in ASTM D2938 (2002). The measured UCS values are provided in Figure 3.

SITE INVESTIGATION

4.3

Thirteen boreholes at the project site were sunk generally to a depth of 20 m below existing seabed. It is understood that an Edeco T30 rotary drive drill rig was used mounted on a jack-up drilling platform. The boreholes were advanced through variably cemented sand using rotary wash boring equipment and techniques.

Carbonate content

Design UCS profile

A UCS strength profile is required for pile design and was assessed from a combination of laboratory testing, field strength estimation and core photographs. Reliable UCS testing of calcarenite tends to be difficult as pre-existing but visually unidentifiable defects and/or anisotropic cementation results in premature

550

5

Figure 3. Design UCS profile.

failure of the UCS specimens and gives results that are generally lower than the true intact rock strength. For drilled and grouted piles, a strength profile chosen from inaccurate laboratory test results might only cause a conservative pile design, whereas a conservatively chosen design UCS profile based on inaccurate test results may result premature refusal of driven piles. It is therefore important to assess an optimum UCS design profile both for the sake of capacity assessment and pile driveability. The design UCS profile was selected based on the strength description in the borehole log, core photographs, measured UCS values from the recovered samples and experience with previous pile driving operations in similar geological formations. If the strength is highly variable with depth (generally observed for cemented sedimentary rocks), a certain averaging of the UCS strength profile relying on experience and judgment is required. At this site, Point Load Index tests (Is50 ) were not carried out. If Is50 values were available, then these values would have been added to the strength data base by converting to inferred UCS values. The inferred UCS values may be obtained by linearly correlating the Is50 values to the adjacent UCS results, after filtering out unreliable tests. Field/laboratory Point Load Index tests are generally numerous and likely to provide better representation of the inherent vertical strength variability in calcarenite than UCS tests. The pile capacity assessment should be based on reasonable lower bound strength parameters, while pile driveability should be based on reasonable upper bound strength parameters. The Is50 values are more likely to provide an upper band of strength data and are therefore valuable for a better prediction of driveability. © 2011 by Taylor & Francis Group, LLC

PILE DESIGN APPROACH

The required pile penetrations to carry the loading platforms access trestle and dolphin loads were verified by a static load test carried out on a test pile (TS38) driven to a penetration of 18.41 m from the seabed level. Based on the uplift load test, an average shaft friction of 81 kPa for the test pile was estimated (excluding the pile and the plug weight). The estimated pile penetration based on this value was found to be too optimistic for other piles, indicated by dynamic tests during production piling. In this case it appears that the longer test pile had a higher average shaft capacity than the shorter working piles. This contrasts with the normal expectation of skin friction degradation, i.e. a lower average shaft capacity for longer driven piles in sands and clays, e.g. Jardine et al. (2005). Rock strength variability may have contributed to this observation. The effects of variations in rock strength can be reduced by using a design method based on sound engineering principles that takes account of strength variability. Beaumont and Thomas (2007) used the clay design method provided in API RP2A WSD (2007) to predict pile capacity of steel tubular piles driven into similar geological formations in the Pilbara coast of the northwest Australia. In their approach, the very weak to weak rock was considered as a cohesive material and the undrained shear strength (su ) of the material was taken as 0.5 × UCS for the purpose of estimating pile capacity. The suitability of this approach for the Fujairah site was assessed and is discussed below. The test pile was installed 10 m from BH8 and 30 m from BH9. The predicted pile capacity based on the API RP2A clay method is provided in Figures 4 and 5 for boreholes BH8 and BH9 respectively. The design UCS profiles provided in Figure 3 were used for estimating pile capacity. The skin friction in tension was assumed to be 0.75 times the skin friction in compression. The end bearing capacity was estimated by applying unit end bearing resistance of 4.5 × UCS over the gross area of the pile, multiplied by a factor of 0.77 to account for the compressibility of the plug, as suggested by Bruno and Randolph (1999). At strata boundaries the full end bearing resistance of the stronger material was assumed to mobilize after a pile penetration of one pile diameter into the stronger material.

6

STATIC LOAD TESTS

A 914 mm OD × 19 mm WT steel tubular test pile without a driving shoe was installed on 25 January 2008 using a Delmag D100-13 diesel hammer. The test pile TS38 is part of the access trestle. The seabed level at the pile location is −14.5 m CD and the pile penetrated to about 18.41 m (pile toe level at about −32.91 m CD).

551

Table 2.

Pile stiffness (kN/mm).

Test direction

Tension

Compression

Unload-reload loop Initial/reload Unload

178 300 250

218 135 253

Figure 4. Ultimate pile capacity – BH8.

Figure 6. Static load test – Pile TS38 (tension negative).

Figure 5. Ultimate pile capacity – BH9.

The test pile was surrounded by four permanent piles in a square grid with a spacing of 5.5 m, which served as reaction piles during the load test. The reaction piles were located 4.2 pile diameters away from the test pile. A static uplift test was carried out 46 days after pile installation (42 days after the dynamic restrike test) followed by a static compression test 3 days after the completion of the static uplift test. The uplift pile capacity was about 4610 kN and appears to have fully mobilized during testing. The compression test was stopped at a pile load of about 5000 kN before full © 2011 by Taylor & Francis Group, LLC

mobilization of pile capacity to avoid overloading of the reaction piles. The load test data is shown in Figure 6. The pile toe movement was assessed to be about 49.4 mm at the maximum uplift load (pile head movement of 66.5 mm at maximum uplift test load with elastic pile elongation of 17 mm). One unload-reload load cycling in both tension and compression loading was carried out during testing. The pile stiffnesses are provided in Table 2. During the uplift load test, a reversal of loading direction was experienced by the pile. The pile was subjected to another reversal of loading direction during the subsequent compression loading. The residual stresses present in the pile after uplift unloading, elastic pile compression and the reversal in principal stress direction are probably responsible for the reduced stiffness apparent during the first compression cycle at a compression load of about 1500 kN. The increased stiffness at the initial stage of the reloading phase of the uplift test following the unload-reload loop is also likely to be due to presence of residual stresses. The compression load cycle indicated a shaft friction reduction to about 1500 kN. This is probably due to the reversal in principal stress directions and Poisson loading effects (reduction or increase in pile diameter as the pile is elongated in tension or compressed in compression). Similar but smaller (maximum 26%) reductions in shaft capacity were recorded by the relatively stiff Imperial College instrumented pile in medium dense silica sand due to a change in loading

552

direction (Chow, 1997). Restrike tests on some piles at Fujairah also showed apparent reductions in shaft friction as the piles were driven deeper. In some cases the magnitude of the reduction was greater than the increase in shaft friction observed during initial set-up (the latter being based on the comparison of the EOID blowcount and the first few blows during restrike). It is generally expected that driven or drilled and grouted piles in siliceous carbonate or carbonate material show post peak friction degradation behaviour. Thomas (1998) among others observed strain softening response in piles jacked in/driven into overconsolidated stiff clays upon a reversal of loading direction. However, in this particular case a reversal of loading direction did not apparently cause any overall strain softening behaviour. It is possible that localized strain softening was masked by progressive shaft failure and/or the gradual mobilization of base capacity with pile toe displacement.

7

Figure 7. Results of dynamic pile tests.

CAPWAP RESULTS

Dynamic testing was carried out on all of the piles in the pile group where TS38 test pile was present. The EoID soil resistance of these piles is provided in Figure 7. The borehole logs and the core photographs indicate a high degree of strength variability in the vertical and lateral directions. During pile driving operations, this variability was evident as the observed blowcounts are quite different for similar adjacent piles driven with same hammer. Dynamic tests carried out on other piles in the same group as TS38 clearly shows this variability. A variation of about 100% in EoID soil resistance to driving was observed between the least resistant and most resistant piles. A larger variability was observed for the end bearing component of the soil resistance. This indicates the presence of bands of stronger rock where a pile can refuse and may result in inadequate uplift capacity if the chosen hammer is not adequately sized.

8

Figure 8. Results of driveability analysis.

DRIVEABILITY ANALYSIS

A driveability analysis was carried out using the GRLWEAP (2005) computer program. The soil resistance to driving was assumed equivalent to the ultimate pile capacity estimated using the API RP2A clay method for a coring pile.This assumption may be warranted for driveability due to the high level of strength variability in calcarenite. The default parameters provided in the GRLWEAP program for the Delmag D100-13 diesel hammer with a hammer efficiency of 65% was used in the analysis. In this analysis, a skin quake of 2 mm, a toe quake of 1 mm, a skin damping of 0.5 sec/m and a toe damping of 0.35 sec/m were used. Driveability analyses were carried out for borehole profiles BH8 and BH9. The predicted blowcounts for these two borehole profiles are provided in Figure 8. © 2011 by Taylor & Francis Group, LLC

Also included in the figure are the observed blowcounts of test pile TS38. The predicted and observed blowcount profiles show excellent comparison and this is another indication of the applicability of the API RP2A clay method for capacity prediction for the siliceous calcarenite present in this site. The observed blowcounts of another pile in the group is also provided as an indication of the variability that was observed. Variability in observed blowcounts during production piling were found to be quite high, some of the piles in a particular group were driven to target penetration effortlessly whilst some other piles in the same group met refusal prior to reaching target penetration. There was no apparent link between driving blowcount and the order that the piles were driven in.

553

• Reversals in pile test direction caused apparent

reductions in shaft capacity due to changes in principal stress direction and Poisson loading effects. • Significant pile set-up may be expected for piles driven into siliceous calcarenite and/or calcareous sandstone. • Pile uplift capacity is about 75% of the compression shaft friction obtained from a reliably conducted dynamic test.

REFERENCES

Figure 9. Measured setup for shaft friction and end bearing.

9

PILE SET-UP

A restrike test was carried out on TS38, 4 days after pile installation. Comparing the results of EoID and restrike tests indicates that the shaft friction increased by about 18% and the end bearing increased by about 16%. Restrike tests were carried out on 14 additional piles of individual pile groups during production piling to assess the setup factor for the pile group. The shaft friction set-up for the other piles was generally observed to be higher than the test pile as shown on Figure 9. The reduction in end bearing resistance was due to the under-mobilisation of end bearing, probably due to inadequate delivery of hammer energy and/or insufficient hammer size. 10

CONCLUSIONS

The following conclusions are drawn from the static and dynamic tests carried out on a 914 mm diameter driven steel tubular pile:

American Petroleum Institute Recommended Practice 2AWSD (2007), Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design. ASTM D2938 (2002), Standard Test Method for Unconfined Compressive Strength of Intact Rock Core Specimens. Beaumont, D. and Thomas, J. (2007), Driving Tubular Steel Piles into Weak Rock – Western Australian Experience, Proceedings, 10th Australia New Zealand Conference on Geomechanics, pp. 430–435. British Standards Institute (1990). BS 1377 Methods of test for Soils for civil engineering purposes. Bruno, D. and Randolph, M.F. (1999), Dynamic and Static Load Testing of Model Piles Driven into Dense Sand, Journal of Geotechnical Engineering Division, ASCE. Chow, F.C. (1997). Investigations into displacement pile behaviour for offshore foundations, Ph.D Thesis, University of London (Imperial College). Clark, A.R. and Walker, B.F. (1977), A proposed scheme for the classification and nomenclature for use in the engineering description of Middle Eastern sedimentary rocks, Geotechnique, Vol. 27, pp. 93–99. Jardine, R.J., Chow, F.C., Overy, R.F. & Standing, J.R. (2005). ICP design methods for driven piles in sands and clays. Thomas Telford (publ). Thomas, J. (1998), Performance of piles and pile groups in clay, PhD Thesis, The University of Western Australia. Uniform Building Code (1997), Structural Engineering Design Provisions, Volume 2, Table 16 I.

• The capacity of a steel tubular pile driven into

rock at this site may be estimated using the API RP2A clay method. However, this approach may not be applicable at all calcarenite sites. Its use should be undertaken with care until proven through driveability trials and pile load tests.

© 2011 by Taylor & Francis Group, LLC

554

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

CPT-Based design method for axial capacity of offshore piles in clays B.F.J. Van Dijk & H.J. Kolk Fugro, Leidschendam, The Netherlands

ABSTRACT: A statistically reliable design method is proposed for large pipe piles driven into clays. The method relies on Cone Penetration Test (CPT) input parameters and a database containing high quality data on axial pile load tests in clays. Factors such as soil plasticity, overconsolidation ratio, pile length, pile slenderness and time between pile driving and loading were considered.

1

INTRODUCTION

Most offshore design methods use peak undrained shear strength as the primary input parameter for determining axial capacity of pipe piles driven into clays, e.g. API (2000) Main Text and Kolk & Van der Velde (1996). However, undrained shear strength is not a unique parameter. A particular “reference” laboratory undrained shear strength must be considered, usually defined by some combination of soil sampling method and laboratory test method. For example, API (2000) is largely based on tube push sampling and unconsolidated undrained triaxial compression and miniature vane testing. In practice, “reference” values are commonly inferred from correlations with non-reference values, such as Cone Penetration Test (CPT) inferred values, supplemented by engineering judgement. This practice introduces significant uncertainties in axial pile design (Van der Wal et al. 2010). The (Piezo) Cone Penetration Test (CPTu) is a more robust method to assess in-situ soil strength. Thus, it is no surprise that many CPT-based design methods are available for determining axial capacity of piles driven into clays. These semi-empirical methods have typically been calibrated for pile geometries and pile types common in onshore practice. They may not be appropriate for relatively long and large diameter pipe piles typically used offshore. This paper proposes a statistically reliable CPT based design method for large pipe piles driven into clays, based on a database containing results of high quality pile load tests in clays.

2

DATABASE

2.1 Selection of pile load tests Thirty-three high quality pile load tests on driven piles at fifteen different locations were selected from public literature (included in the reference list) for this study. High quality refers to both the pile load test and © 2011 by Taylor & Francis Group, LLC

the soil data. Typical reasons for excluding other pile load tests include: test pile pushed rather than driven, other non-representative installation methods, implausible differences in repeat pile load tests, uncertainty on representative CPT data, and significant amounts of sand inclusions within clay layers. Starting point for the database was an earlier study by Kolk & Van der Velde (1996). The Kolk & Van der Velde database (KV-database) was extended with results from eighteen other high quality pile load tests. The database contains results of both axial compression and tension pile load tests. The piles selected for the database are predominately friction piles in clay for which sufficient high quality soil and load test data were readily available. Details of the pile load tests are presented in Table 1.

2.2

Soil parameters

For the database, CPT cone resistance qc , CPT pore pressure u2 and equilibrium in-situ pore pressure u0 estimates were determined at 0.1 m depth spacing. Additionally, average values of soil unit weight, plasticity index PI, liquid limit LL and undrained shear strength cu were determined. The net cone resistance qn was derived using:

where σvo = total in-situ overburden pressure; and a = the cone dependent net area ratio of the crosssectional steel area at the gap between cone and friction sleeve to the cone base area. No values for the net cone resistance qn and/or the pore pressure (u2 ) were available for 12 pile load tests. For these instances, qn was derived from qc by estimating pore pressure u2 according to correlations such as presented by Mayne (1990) and Robertson (1990). For 15 pile load tests the cone dependent net area ratio a was not available, in which case a value of 0.75 was selected.

555

Table 1.

Database of selected pile load tests.

Location

Load Length Diameter Pile tip Set-up test** m m – days –

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33

22.6 4.5 3.7 3.6 9.1 9.1 9.0 9.0 48.0 48.0 11.6 11.6 11.6 13.1 13.1 10.8 18.6 21.9 21.9 9.4 50.0 10.0 10.0 10.0 40.0 7.5 9.6 7.6 14.5 14.5 30.0 71.4 18.3

Aquatic Park* Canons Park Canons Park Canons Park Cowden Cowden Cowden Cowden Garigliano Garigliano* Hamilton Hamilton* Hamilton* Houston* Houston* Lulu Island Kontich* Kontich* Kontich Merville Mortaiolo* Onsoy Onsoy Onsoy Pentre* Pentre* Tilbrook* Tilbrook* Tilbrook Tilbrook Tilbrook* West Delta West Sole*

0.762 0.168 0.102 0.102 0.457 0.457 0.305 0.203 0.381 0.381 0.114 0.114 0.114 0.273 0.273 0.324 0.61 0.61 0.61 0.508 0.477 0.219 0.219 0.812 0.762 0.219 0.219 0.219 0.219 0.273 0.762 0.762 0.762

Open Closed Closed Open Open Closed Open Open Closed Closed Closed Open Open Closed Closed Closed Open Open Open Open Closed Closed Closed Open Open Closed Closed Closed Closed Open Open Open Open

60 60 2 2 30 30 – – – – 18 18 18 105 105 82 12 21 26 41 25 40 40 40 44 31 60 60 60 60 134 133 0.25

T C T T C C T T C C C C C C C C C C T C C T T T C T T T T T C T C

Figure 1. Predicted versus measured capacity.

Bustamante et al. (1994) reported that the tip of pile 1 at Garigliano (no. 9) was installed in a sand layer with a high end-bearing capacity. The end bearing was measured during the test. The measured ultimate pile capacity for this pile was corrected for this measured end bearing value. Thus, only the shaft friction of the pile was used in the study.

* Derived from Kolk & Van der Velde (1996) database. ** T = Tension loading; C = Compressive loading test.

3

The normalised cone resistance Qt was used as a measure for the overconsolidation ratio OCR:


where σvo = effective in-situ overburden pressure. This links with Chen & Mayne (1996) who found OCR = 0.32 Qt for 205 clay sites around the world.

2.3

Pile test results

The ultimate resistance Q of a pile was defined as the maximum resistance measured during the test, corrected for the effective weight of the pile (total weight of the pile minus the weight of the displaced soil), giving consideration to whether testing was in compression or tension. Unit shaft friction f in occasional sand layers was calculated using API (2000) Main Text recommen
dations for dense sand: f = min (βσvo , fmax ) where β = shaft friction factor; and fmax = limiting unit shaft friction value. © 2011 by Taylor & Francis Group, LLC

EXISTING DESIGN METHODS

Three existing CPT-based design methods to determine axial capacity of piles in clays were considered (Fig. 1): Bustamante & Gianeselli (1982), Almeida et al. (1996) and CUR (2001). Table 2 shows statistics for the ratio of predicted to measured pile capacity, R. The CUR (2001) method provides the better predictions with R = 0.98. However, 39% of the tests are either under or over predicting by more than 20%. Only the Almeida et al. (1996) method considers the overconsolidation ratio. Although their study did not explicitly reveal any length effects, they recommend to reduce shaft friction as suggested by Semple & Rigden (1984) or Randolph & Murphy (1985) in case the ratio embedded pile length L to diameter D exceeds 60. The Almeida et al. (1996) method also considers a more conservative relation in case PI < 20%. 4 4.1

DEVELOPMENT OF NEW DESIGN METHOD General equation

A new design method was developed, considering possible dependency of shaft friction on plasticity and

556

overconsolidation of the soil, and pile characteristics such as length, diameter and wall thickness WT. The general equation for pile capacity in compression was used, assuming a fully plugged pile tip condition:

where q is unit end bearing and z is depth below seafloor. In case of tension, this study considered reversed end bearing where applicable. However, reversed end bearing may not always occur (e.g. when thin sand layers are present). For design purposes, following standard engineering practice for tension piles, the authors suggest not to rely on pile tip resistance for tension piles (i.e. no suction) and thus to omit the first term in Equation (3). It is noted that offshore piles are usually long and slender steel pipe piles with a relatively large diameter. Usually the pile tip condition of these piles is plugged under static loading.

Figure 2. R versus Plasticity Index PI.

where z = depth below seafloor; h = distance between pile tip level and z. A basic equation for computing unit shaft friction using a CPT method can be written as:

The following mathematical expression allows exploring dependency of ks on a range of parameters:

4.2 End bearing The following equation was adopted for calculating unit end bearing q in clays:

where qn,av = average qn value. This equation corresponds with the commonly used relation q = 9 cu , with a cone factor Nk = qn /cu of 13 (NGI 2006). For qn,av the authors suggest using the average qn value between 1.5D above and 1.5D below pile tip level. This is according to principles for CPT-based methods as given by API (2000) for sandy soils. Equation (4) could not be verified because of lack of reliable data. However the end bearing of long offshore piles in clay is usually small compared to the shaft friction. Any prediction errors thus have limited effect on Q. 4.3 Outer shaft friction The Kolk & Van der Velde (1996) method was considered as starting point for prediction of the outer unit shaft friction f . This method accounts for pile length effects and soil overconsolidation. The method is an α-method. It uses the peak intact undrained shear strength cu (based on unconsolidated undrained triaxial compression tests) as the primary soil parameter, according to:

and

© 2011 by Taylor & Francis Group, LLC

where Ar = 1-((D-2WT )/D)2 = pile tip displacement ratio; j, k, l, m, n and o are dimensionless parameters; and uL = unit length to render the expression dimensionless = 1.0 m = 3.3 feet. Parameters j, k, l, m, n and o were determined using regression analysis, exploring lowest coefficient of variation on the ratio of predicted to measured pile capacity, R. Using the most suitable values considering only the pile load tests of the KV-database, Equation (8) becomes:

A value of 0.08 was selected for kmax . This value corresponds with a unit shaft friction equal to cu when Nk is 13 (NGI 2006). For the pile load tests of KV-database the new method (combining Equations (3), (4), (7) and (9)) results in an average of 0.99 and a coefficient of variation of 0.13 for R. 4.3.1 Soil plasticity Figure 2 confirms that the pile friction as described by Equation (9) is not significantly affected by the plasticity of clay. This conclusion was also drawn by Kolk & Van der Velde (1996). 4.3.2 Pile length Previous studies (Randolph & Murphy 1985; Kolk & Van der Velde 1996; Jardine et al. 2005) revealed that pile unit shaft fiction reduces in relation to pile length. This may be attributed to remoulding of clay and large soil strains during pile installation and subsequent pile loading.

557

Figure 3. R versus pile length L.

Figure 6. R versus pile displacement ratio Ar .

Figure 4. R versus pile length L/D.

Figure 7. R versus time factor T .

Figure 5. R versus normalised cone resistance Qt .

4.3.5 Time between pile driving and loading Enough time should be allowed between pile driving and pile loading, to allow for dissipation of excess pore pressures and set-up of the soil. According to Randolph & Wroth (1979) 90% consolidation around solid driven piles occurs at times t between 5 D2 /c and 15 D2 /c, where c = coefficient of consolidation. Figure 7 shows a time factor T = ct/D2 > 5 to 15 for most tests, based on an empirical relation between c and LL according to c = 10−LL/40 , derived from NAVFAC (1981). No clear change of R with time is observed suggesting that both the predicted and the measured capacities in the database are representative for long-term capacity. It should be noted that this assessment is crude and therefore no design specific recommendations can be given with respect to set-up behaviour of piles in clay.

It was expected that pile slenderness L/D would affect pile capacity. Surprisingly, Figures 3 and 4 confirmed that pile frictional capacity as described by empirical Equation (9) was affected by pile length rather than by pile slenderness. 4.3.3 Overconsolidation Both API (2000) and Kolk & Van der Velde (1996) methods indicate that the pile friction to soil strength ratio, α, reduces as the OCR increases. Figure 5 confirms that the pile friction to cone resistance ratio, ks , decreases with increasing Qt as described by Equation (9). Here, Qt serves as measure of overconsolidation. 4.3.4 Soil displacement during installation It was explored if pile frictional capacity would be related to the amount of soil displaced by the penetrating pile. Pile tip displacement ratio Ar was taken as a measure for soil displacement. No direct relation between the pile displacement ratio and the measured shaft friction was found as can be observed in Figure 6. The new design method is applicable for both open ended and closed ended piles. © 2011 by Taylor & Francis Group, LLC

5 VALIDATION OF NEW DESIGN METHOD The entire database was used to validate the new method. Table 2 shows statistics for the ratio of predicted to measured pile capacity, R. Figure 8 presents results for the new design method. Table 2 and Figure 8 show improved predictions by the new CPT-based design compared to the selected existing design methods. Only 12% of the load tests is either under or over predicted by more than 20%.

6

LIMITATIONS

The proposed method is subject to limitations, most of which are common also to other design methods. Comments are as follows.

558

The proposed design method considers unit shaft friction affected by overconsolidation and pile length, but not by load direction, soil plasticity, pile diameter and pile tip displacement ratio.

Table 2. The average and coefficient of variation of R for existing design methods and the new design method.

Average of R

Coefficient of Variation of R

Database:

Database:

Method

full

KV

full

KV

ACKNOWLEDGEMENTS

Bustamante Almeida CUR New method

0.55 0.87 0.98 1.00

0.52 0.86 1.03 0.99

0.19 0.19 0.22 0.14

0.17 0.22 0.20 0.13

The authors gratefully acknowledge Fugro’s commitment and support to improving geotechnical practice. REFERENCES

Figure 8. Measured versus predicted capacity for the new design method.

Database limitations include interpretational issues on pile failure load and representative soil profiles (Jeanjean et al. 2010). Measurement uncertainties also apply. Influence of soil sensitivity was not studied. The bearing capacity may be adversely affected for sensitive soils. Screening of CPT data will be necessary for intermediate soils where cone penetration may be partially drained, giving high qn values compared to undrained conditions. Pile loading may be undrained for intermediate soils.

7

CONCLUSIONS

A simple and reliable CPT-based design method is proposed for prediction of the axial capacity of long steel pipe friction piles installed in clay, typically used offshore. The method is based on regression analysis, exploring lowest coefficient of variation. It was initially developed from an existing pile load test database and subsequently verified using an augmented database. © 2011 by Taylor & Francis Group, LLC

APIAmerican Petroleum Institute 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design, API Recommended Practice 2A-WSD, 21st Edition. (With Errata and Supplement 1, December 2002, Errata and Supplement 2, October 2005, and Errata and Supplement 3, March 2008). Almeida, M.S.S., Danziger, F.A.B. & Lunne, T. 1996. Use of the piezocone test to predict the axial capacity of driven and jacked piles in clay, Canadian Geotechnical Journal 33 (1), 23–41. Bogard, D. & Matlock, H. (1998). Static and Cyclic Load Testing of a 30-inch-diameter Pile over a 2.5-year Period, Proceedings 30th Annual Offshore Technology Conference, 4–7 May, 1998, Houston, Texas, U.S.A, Vol. 1: 455–468 Bond, A.J. & Jardine, R.J. 1990. Research on the behaviour of displacement piles in overconsolidated clay.Technology Report OTH 89296, The imperial College London. BRE 1985. Comparison of piezocones in overconsolidated clays. Building Research Establishment and Norwegian Geotechnical Institute. Report No. 84223-1. Bustamante, M. & Gianeselli, L. 1982. Pile Bearing Capacity Predictions by Means of Static Penetrometer CPT, Proceedings of the Second European Symposium on Penetration Testing, ESOPT-II, Amsterdam, The Netherlands, Vol. 2: 493–500. Bustamante, M., Gianeselli, L., Mandolini, A. & Viggiani, C., 1994. Loading Tests on Slender Driven Piles in Clay, Proceedings Thirteenth International Conference on Soil Mechanics and Foundation Engineering, New Delhi, Vol. 2: pp. 685–688. Chen, B.S.Y. & Mayne, P.W. 1996. Statistical Relationships between Piezocone Measurements and Stress History of Clays, Canadian Geotechnical Journal, Vol. 33, No. 3: 488–498. Clarke, J., Rigden, W.J. & Senner, D.W.F. 1985 Reinterpretation of the West Sole Platform ‘WC’ Pile Load Tests, Géotechnique, Vol. 35, No. 4, pp. 393–412. CUR 2001. Bearing Capacity of Steel Pipe Piles,CUR-report 2001-8, July 2001, CUR, Gouda, Netherlands. Gibbs, C.E., McAuley, J., Mirza, U.A. & Cox, W.R. 1992. Reduction of Field Data and Interpretation of Results for Axial Load Tests of Two 762 mm Diameter Pipe Piles in Clays, Proceedings of the Conference ‘Recent Large-scale Fully Instrumented Pile Tests in Clay’: 285–345. Heerema, E.P. 1979. Pile Driving and Static Load Tests on Piles in Stiff Clay, in EleventhAnnual OffshoreTechnology Conference, Vol. 2: 1135–1145. ISSMGE 1999. International Reference Test Procedure for the Cone Penetration Test (CPT) and the Cone Penetration Test with Pore Pressure (CPTU): Proceedings of the Twelfth European Conference on Soil Mechanics and Geotechnical Engineering, Amsterdam, Netherlands,

559

7–10 June 1999, Vol. 3, A.A. Balkema, Rotterdam: 2195–2222. Jardine, R., Chow, F.C., Overy, R.F. & Standing, J.R., 2005. ICP Design Methods for Driven Piles in Sands and Clays, Thomas Telford Ltd., London. Jeanjean, P. Watson, P.G., Kolk, H.J. and Lacasse, S., 2010. RP 2GEO: The New API Recommended Practice for Geotechnical Engineering, Proceedings Offshore Technology Conference, 3–6 May 1996, Houston, Texas, U.S.A Karlsrud, K., Hansen, S.B., Dyvik, R. & Kalsnes, B., 1993. NGI’s Pile Tests at Tilbrook and Pentre – Review of Testing Procedures and Results, Proceedings of the Conference ‘Recent Large-scale Fully Instrumented Pile Tests in Clay’: 405-429 Karlsrud, K., Kalsnes, B. & Nowacki, F., 1993. Response of Piles in Soft Clay and Silt Deposits to Static and Cyclic Axial Loading Based on Recent Instrumented Pile Load Tests, Advances in Underwater Technology, Ocean Science and Offshore Engineering, Vol. 28, pp. 549–583. Kirby, R.C. & Roussel, G. 1979. ESACC Project Field Model Pile Load Test, Hamilton Air Force Base Test Site Norato, California, Amoco Production Company. Kolk, H.J. and Van der Velde, E. 1996. A Reliable Method to Determine Friction Capacity of Driven Piles in Clay, Proceedings 28th Annual Offshore Technology Conference, 6–9 May 1996, Houston, Texas, U.S.A., Vol. 1: 337–346. Lambson, M.D., Clare, D.G. & Semple, R.M. 1992. Investigation and Interpretation of Pentre and Tilbrook Grange Soil Conditions, Proceedings of the Conference ‘Recent Large-scale Fully Instrumented Pile Tests in Clay’: 134– 196. Mayne, P. W. Kulhawy, F. H. and Kay, J.N. 1990. Observations on the development of pore-water stresses during piezocone penetration in clays, Canadian Geotechnical Journal, Vol. 27: 418–428. McClelland Engineers, inc. 1982. Geotechnical Investigation Borings 4, 5 & 6, Block 58, West Delta Area, Gulf of Mexico, Report No. 0181–0217. NAVFAC DM-71 1981. Soil Mechanics, Design Manual 7.1, Department of the Navy, Naval Facilities Engineering Command, 7.1–144. Norwegian Geotechnical Institute, 2006. Shear Strength Parameters Determined by In situ Tests for Deep Water Soft Soils, NGI Report No. 20041618-6. Olson, R.E. 1984. Analysis of Pile Response under Axial Loads, Final Report on Project 83–42B to API

© 2011 by Taylor & Francis Group, LLC

O’Neill, M. W., Hawkins, R. A., & Mahar, L. J., 1981. Field Study of Pile Group Action. Final Report, Report No. FHWA/RD-81/002 and Appendix C, FHWA/RD-81/005. Pelletier, J.H. and Doyle, E.H. 1982. Tension Capacity in Silty Clays - Beta Pile Test, in Proceedings 2nd International Conference on Numerical Methods in Offshore Piling, April 29–30, 1982, Austin, Texas: 163–181. Price, G. & Wardle, I.F. 1982. A comparison between cone penetration test results and the performance of small diameter instrumented piles in stiff clay. Proceedings of the Second European Symposium on Penetration Testing, Amsterdam. Randolph, M.F. and Murphy, B.S. 1985. Shaft Capacity of Driven Piles in Clay, Proceedings 17th Annual Offshore Technology Conference, Vol. 1, 371–378. Randolph, M.F. & Wroth, C.P. 1979. An Analytical Solution for the Consolidation around a Driven Pile, International Journal for Numerical and Analytical Methods in Geomechanics, Vol. 3, No. 3: 217–229. Rigden, W.J., Pettit, J.J., St. John, H.D. & Poskitt, T.J. 1979. Developments in Piling for Offshore Structures, Proceedings of the Second International Conference on Behaviour of Offshore Structures, London: 1177–1182. Robertson, P.K. 1990. Soil Classification Using the Cone Penetration Test, Canadian Geotechnical Journal, Vol. 27, No. 1, 151–158. Robertson, P.K., Campanella R.G., Davies M.P. & Sy, A, 1988. Axial Capacity of Driven Piles in Deltaic Soils using CPT, Penetration Testing, 1988, ISOPT-1: 919–928 Semple, R.M. & Rigden, W.J. 1984. Shaft Capacity of Driven Pipe Piles in Clay, Proceedings, Symposium on Analysis and Design of Pile Foundations at ASCE National Convention, San Francisco, California, American Society of Civil Engineers, New York, 59–79. Totani, G., Marchetti, S., Calabrese, M. & Monaco, P. 1994. Field Studies of an Instrumented Full-scale Pile Driven in Clay, in Proceedings Thirteenth International Conference on Soil Mechanics and Foundation Engineering, New Delhi, 5–10 January 1994, Vol. 2, Oxford & IBH Publishing, New Delhi: 695–698. Togliani, G. 2008. Pile capacity predictions from in situ tests. Proceedings of the 3rd International Conference site Characterization, Taipei, Taiwan: 1187–1192. Wal, T. van der, Goedemoed, S., & Peuchen, J. 2010. Bias reduction on CPT-based correlations, Proceedings CPT10, Huntington Beach, California, May.

560

7 Foundations for renewable energy

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Evaluation of pile capacity approaches with respect to piles for wind energy foundations in the North Sea M. Achmus & M. Müller Institute of Soil Mechanics, Foundation Engineering and Waterpower Engineering, Leibniz University of Hannover, Germany

ABSTRACT: The design of axially loaded piles in non-cohesive soils is usually based on the β-approach. Recently, alternative CPT-based approaches have been developed by different research groups. However, in general the different approaches give different results, and it is not clear which approach is the best suited, i.e. most realistic, for special conditions. For the conditions of offshore wind energy foundations in the North Sea and with regard to tensile axial loading, a parametric study was carried out to ascertain the differences in pile capacity between the different approaches. It was found that in most cases the β-approach yields smaller pile capacities than the new CPT-based approaches. However, a calibration of the CPT-based methods for the considered conditions of piles with large diameter in dense to very dense sand has not yet been carried out. The new methods must therefore be applied with due caution until verified by experiments or sufficient experience. 1

INTRODUCTION

Offshore wind energy converters (OWECs) offer huge potential for the expansion of renewable energy in the North Sea in Germany. For water depths between 25 and 50 m tripod or jacket structures are suitable. These structures are supported by open steel pipe piles with diameters between 1.5 and 3 m. Under extreme loads induced by wind and wave loading, large axial tension forces have to be transferred to the ground, which in most cases is design-driving with regard to the required pile length. Since the erection of thousands of wind energy converters is planned for the North Sea, the optimization of the design and thus an accurate prediction of tensile pile capacities is very important. In the API regulations (API 2000, API 2007) usually applied in offshore engineering, the β-approach for calculating the axial pile resistance of offshore piles in non-cohesive soils is given. This method has been used successfully in offshore engineering for a long time. However, it is now known to possibly either underestimate or in some cases overestimate the actual pile capacity significantly. To overcome the disadvantages of the current design method, CPT-based approaches have been developed by different research groups. These empirical approaches are already included in current regulations for offshore structures. However, in general the different approaches give different results, and it is yet not clear which approach is the best suited, i.e. most realistic, for special boundary conditions. The soil conditions in the North Sea are mainly characterized by sandy soils, whose relative densities range from at least medium dense to dense and often very © 2011 by Taylor & Francis Group, LLC

dense. For these conditions a parametric study was carried out to ascertain the differences in tensile pile capacity between the different CPT-based approaches and the β-approach. The experimental data base used for the approaches was evaluated in order to identify the approach best suited to the particular geometric, loading and subsoil conditions considered here.

2

DESIGN METHODS

In water depths of more than 20 m, tripod or jacket foundation structures can be favourably used for the support of the wind tower. Usually open-ended steel pipe piles with a constant outer diameter between 1.5 m and 3.0 m are used to transfer the loads into the ground. Since the required depth of the driven piles in most cases depends on the tension loading case, only tensile pile capacities are considered in this paper. In general the tensile bearing capacity of piles consists of the pile’s weight and outer and inner skin friction. In the case of plugging, the latter is limited to the total weight of the soil plug inside the pile.

where fto = outer unit skin friction for tension; fti = inner unit skin friction for tension; Ao = outer pile shaft area; Ai = inner pile shaft area; Gs
= effective steel weight of pile; and Gp
= effective weight of soil plug inside the pile. The design values for skin friction f may be established either on the basis of test results or according to empirical approaches. In offshore technology pile

563

Toolan et al. (1990) published the results of pile tests which showed an overestimation of pile capacity in the case of long piles in loose to medium dense sand. Lehane (2005) reported that the β-method is conservative for relatively short offshore piles (<45 m embedded length) in dense to very dense sands but may overestimate the pile capacity in all other conditions. The following influencing factors mainly affect the axial pile capacity (e.g. Schneider 2007, Lammertz 2008): – Subsoil type and state (relative density); – pile type (with respect to soil displacement and/or plugging); – interface shearing angle (friction between pile shaft and soil); – method of installation (friction fatigue); – soil behavior under shearing (dilatancy) and – type of load (interaction between pile tip and shaft resistance in case of compression load).

Figure 1. Forces contributing to tensile pile capacity in the unplugged (left) and in the plugged case (right). Table 1. Design parameters for predicting shaft friction in cohesionless soil (API 2007).

Relative density

soil

Medium dense Medium dense Dense Dense Very Dense Very Dense

Sand-Silt Sand Sand-Silt Sand Sand-Silt Sand

β [-]

The use of cone penetration test (CPT) results potentially allows a more precise reflection of soil density, compressibility and stress level than the consideration of the subsoil only with regard to relative density in the β-method. In 2007 the API published an Errata and Supplement 3 to the guideline API RP 2A, including new “CPT-based methods”. These approaches consider all the influencing factors given above and should thus allow a more accurate calculation of the pile capacity for a wide range of non-cohesive soils. However, offshore experience with the application of these CPT methods is still limited and therefore more experience is needed before they can be recommended for routine design, to replace the API β-method. The CPT-based methods which were introduced in the API (2007) are simplified versions of the full versions published by different research groups (Jardine et al. 2005, Kolk et al. 2005, Lehane et al. 2005, Clausen et al. 2005). These simplified methods can yield slightly different results than the full versions of these methods, but for the case of offshore piles these differences are assumed to be small. The CPT-method results discussed in this paper were derived from the simplified versions given in API (2007):

ft,max [kPa]

0.29 0.37

67 81

0.46

96

0.56

115

testing is very expensive, so empirical methods are necessary to establish the design values. The common method of calculating tensile capacities of offshore piles is the β-method given in the API RP 2A regulation (API 2000). Here, skin friction values are given dependent on the relative density of the soil and on the vertical effective overburden stress at the considered depth. In API (2000) skin friction values for tension and compression load are not distinguished. However, according to regulations of the Germanische Lloyd (GL 2005) usually applied for structures in the German North Sea, the skin friction under tension load should be assumed to be considerably smaller than the skin friction under compression load with ft = 2/3 fc . Taking this into account, the skin friction under tension load can be formulated as given in Equation (2).

1. 2. 3. 4.

where σv
= effective overburden stress; ft,max = limiting value of shaft friction to be taken from Table 1; and β = shaft friction factor to be taken from Table 1. This method was used in offshore engineering for a long time. However, it is now known to possibly either underestimate or overestimate the actual pile capacity, since the effects of the pile installation process on the axial capacity are not adequately considered. © 2011 by Taylor & Francis Group, LLC

564

Simplified ICP-05 Offshore UWA-05 Fugro-05 NGI-05

To determine skin friction according to methods 1, 2 and 3 (in the following simply termed ICP, UWA, FUGRO), the following general formula can be used (API 2007).

where qc,z = CPT cone tip resistance at depth
= effective vertical in-situ soil stress; pa = z; σv0 atmospheric pressure = 100 kPa; Ar = effective area

Table 2. Unit skin friction parameters for tension loading for the methods ICP, UWA and FUGRO (API 2007). Method

a

b

c

d

u

v

1 (ICP) 2 (UWA) 3 (FUGRO)

0.1 0.0 0.15

0.2 0.3 0.42

0.4 0.5 0.85

1 1 0

0.016 0.022 0.025

√ 4 Ar 2√ 2 Ar

ratio Ar = 1 − (Di /Do )2 ; Do = pile outer diameter; Di = pile inner diameter; L = embedded pile length; δcv = critical interface friction angle; a, b, c, d, u and v = empirical parameters to be taken from Table 2 (given for tension loading). The fourth method for estimating the skin friction is the NGI approach (Eqs. 4 and 5).

where Dr = relative density of the soil. Research results indicate that the interface shearing angle δcv for steel piles in sand is dependent on the soil characteristics, i.e. the mean grain size D50 (Jardine et al. 2005, Lehane et al. 2005, Fig. 2). For driven piles, a maximum value of δcv = 28.8◦ was found, since ramming in dense sands leads to abrasion of the steel pile surface and to particle crushing of the sand, both decreasing the friction coefficient. Both UWA and simplified ICP methods use this approach for the interface shearing angle. In the original ICP method larger shearing angles are considered, but in API (2007) the general recommendation to limit tan δcv to 0.55, i.e. δcv = 28.8◦ , is given. The FUGRO and NGI approaches both apply a constant friction coefficient. For open steel pipe piles the pile capacity is generally affected by possible plugging, described by the incremental filling ratio IFR. However, for offshore piles with a relatively large diameter it can usually be assumed that no plugging occurs during installation, i.e. IFR = 1.0 (API 2007). This means that the effective area ratio Ar used in Eq. (3) can be determined dependent only on the inner and outer pile diameters by Ar = 1−(Di /Do )2 . In all CPT-based methods the effect of friction fatigue is considered. During pile driving, the soil in contact with the pile is permanently sheared until the pile reaches its final penetration depth. This leads to a reduction of the skin friction to be considered for pile capacity determination. Obviously, the fatigue effect must be the greater, the larger the pile depth below the point of consideration is. Thus, at the pile head the skin friction is more influenced by friction fatigue than at the pile tip. In Eq. (3) the friction fatigue is accounted for by the term (D/(L − z))c . Thus, a non-linear dependence on pile depth is considered in the methods ICP, UWA and © 2011 by Taylor & Francis Group, LLC

Figure 2. Dependence of interface shearing angle on sand coarseness considered in ICP and UWA approaches.

Fugro, whereas in the NGI method a linear approach is chosen with the term z/L. In the case of shear loading, sandy soil shows a dilatant behavior, which, for the case of axially loaded piles, implies an increase in lateral stress during pile loading. However, the influence of this effect decreases with increasing pile diameter. In offshore applications piles with relatively large diameters are usually used. In the Eqs. (3) to (5) given in API (2007) the terms reflecting lateral stress increase due to dilatancy are thus left out. Finally, the interaction of base resistance and skin friction for piles under compressive loading is considered in the CPT-based methods, in contrast to the API β-approach. This is not reflected here, since only tensile pile capacity is considered.

3

DATABASE EVALUATION

The CPT-based methods described above are all semiempirical approaches, which were calibrated against a database of pile test results. Although the different databases had a large number of tests in common, they were in general different. Tests in very different soils and with different pile systems (open-ended and closed-ended steel piles, rectangular concrete piles) are included. Most of the tested piles had diameters smaller than 1.0 m. The subsoil in the North Sea typically consists mainly of sandy soils, which are at least in a medium dense and often dense to very dense state. Intermediate cohesive layers occur, but normally with limited thicknesses. The piles to be used for tripod or jacket foundations are usually open-ended steel pipe piles with diameters between 1.5 and 3 m and slenderness ratios (embedded length to diameter) between L/D = 10 and L/D = 40. Pile tests which are relevant to these conditions are very scarce in the databases. The database of the NGI approach consists of 85 test results. Most of them were carried out in loose to medium dense sands (Clausen et al. 2005). Only a few open-ended steel pipe piles in dense sand are part

565

Table 3.

Mean value and standard deviation of Qc /Qm .

Qc /Qm mean COV

© 2011 by Taylor & Francis Group, LLC

ICP

UWA

FUGRO

NGI

0.60 0.29

0.88 0.15

0.82 0.14

1.03 0.28

1.15 0.21

to the test results used, the mean values and the standard deviations (coefficient of variation, COV) of the Qc /Qm -values were calculated, as shown in Table 3. The mean value of the FUGRO approach gives the best results. However, the relatively large standard deviation also shows a large scatter, i.e. large deviations with either overprediction or underprediction of pile capacity are possible. The NGI method tends to overpredict the axial pile capacity under tension loads for the conditions considered. The ICP and UWA approaches seem to be most suitable here. On average, a slight underprediction of pile capacity is to be expected. Moreover, due to the relatively small standard deviations, the scatter is limited and the prediction thus relatively certain.

Figure 3. Comparison of calculated and measured tensile pile capacities with respect to pile slenderness ratio L/D.

of the database, whereas the diameter of these piles is smaller than 1.0 m and the slenderness ratio is quite large. Only 7% of the test results – the EURIPIDES tests – in the database are relevant to the boundary conditions in the German North Sea. ICP used a database of 39 test results for openended steel piles. Several of these tests were carried out in Dunkirk in France (Jardine et al. 2005). The soil in Dunkirk was dense sand and the piles tested had the relevant small slenderness ratios. Only the dimensions were relatively small. In this database about 26% of the tests are relevant to the boundary conditions considered here. The FUGRO database is largely the same as the ICP database. It consists of 45 test results, of which 25% are representative of piles for offshore wind energy structures in the German North Sea. The database of UWA consists of 89 tests on steel piles and is thus the largest of the databases. However, the biggest part of the database consists of tests on piles with quite large slenderness ratios. 12% of the tests are relevant for the boundary conditions considered here. These tests are also part of the NGI, ICP and FUGRO databases. All in all, only 6 test results included in the different databases, mainly stemming from the CLAROM, GOPAL, HOOGZAND and EURIPIDES test series, are relevant to open-ended steel pipe piles in sandy soil. For these tests the pile capacities were determined with the different approaches, and the results are shown in Fig. 3 in terms of ratio of calculated to measured pile capacity Qc /Qm over slenderness ratio L/D. For application of the API method, the relative densities of the soil layers were derived from the CPT diagrams given using the method proposed by Jamiolkowski et al. (2001). Due to the limited number of tests, a final assessment of the approaches based on them is difficult. It can however be stated that theAPI β -approach greately underestimates the axial pile capacity for piles with slenderness ratios of less than 20. In order to compare the quality of the different approaches with respect

API

4

PARAMETRIC STUDY

The above considerations show that none of the different design approaches are sufficiently reliable with regard to open-ended steel pipe piles with small slenderness ratios in sandy soil. A parametric study was carried out to elucidate the sensitivity of the calculated tension pile capacities on the most important system parameters. Here, the pile geometry (embedded length L and diameter D) and the relative density Dr of the sandy soil are varied. Driven steel pipe piles with lengths between 20 and 50 m and diameters between 1 m and 3 m, as to be expected for tripod and jacket foundations in the North Sea, are considered.The wall thickness of the steel pipe was set to t = D/40. Homogeneous sand was assumed, and the relative density Dr was varied between 0.5 and 1.0, i.e. between medium dense to dense and very dense state. The CPT cone resistances qc to be used in the CPT-based methods were determined depending on relative density and overburden stress using the approach of Jamiolkowski et al. (2001). For all parameter combinations, the tensile pile capacities were calculated with all of the CPT-based methods and with the common API β-approach as described above. The pile capacity according to API QAPI was used as a reference value. The ratios of the pile capacities from the CPT-based methods to the reference value Q/QAPI are presented for three different relative densities in Fig. 4 dependent on the pile slenderness ratio L/D. Fig. 4 clearly shows that distinct differences between the pile capacities can occur. The CPT-based

566

Figure 5. Comparison of tension pile capacities derived from the different approaches with respect to the pile diameter.

However, the application of the CPT-based methods must be done very cautiously, since the results are actually based only on the extrapolation of experimental test results. For instance, the large pile capacities according to the FUGRO approach for very dense sand and short piles are more of a theoretical nature, since the approach was not calibrated on measurement results for such conditions. In Fig. 5 the dependence of the ratio Q/QAPI on the absolute pile diameter is presented. The trend lines given represent the mean values of all calculations with different pile lengths and relative densities. For the approaches ICP, UWA and FUGRO the Q/QAPI values are the greater, the larger the pile diameter is. A very strong increase is predicted by the FUGRO approach. In contrast, the NGI approach does not predict an increase of Q/QAPI with the pile diameter at all. However, it must be noted that for such large pile diameters no experimental evidence exists. This means that the factors Q/QAPI between 1.4 and 2.3 obtained for a pile diameter of 3 m are in no way secured. 5 Figure 4. Comparison of tension pile capacities derived from the different approaches with respect to slenderness ratio and relative density.

approaches mainly yield larger pile capacities than the API approach, in particular for very dense sands and piles with small slenderness ratios. Regarding the effect of pile slenderness, all CPTbased methods show an increase in the Q/QAPI -value with decreasing slenderness ratio. The difference is of minor importance with the NGI approach, whereas with the FUGRO approach an extreme increase in pile capacity results for very dense sand. The API β-approach seems to considerably underestimate pile capacities in dense or very dense sands. © 2011 by Taylor & Francis Group, LLC

567

CONCLUSIONS

For the erection of thousands of wind energy converters in the North Sea an enormous number of piles must be installed in future. For economic reasons, an accurate prediction of pile capacities and in particular tensile pile capacities is very important. Against this background, the tensile pile capacities predicted by the β-approach and by the recently developed CPT-based methods were compared and assessed with respect to open-ended steel pipe piles with small slenderness ratios in typical North Sea soil conditions, i.e. dense to very dense sands. The evaluation of the databases against which the new CPT-based methods were calibrated showed that only a very limited number of pile test results were

carried out under conditions similar to the abovementioned conditions. Thus, none of the existing design methods can be classed as sufficiently reliable for the special conditions considered. In particular, no experimental evidence exists regarding the relatively large capacities predicted by the CPT-based methods for piles with large diameters of 2 or 3 m. The sensitivity analysis carried out in a parametric study clearly showed that the relative density of the sand subsoil and the absolute pile diameter strongly affect the results of the different methods. Based on the comparison with a very limited number of relevant pile test results, the ICP and UWA approaches seem to be the most suitable for the application case considered. However, at least for very dense sands and very large diameters, these approaches should be applied very cautiously, i.e. introducing conservative assumptions. More experience and experimental pile tests are necessary to improve the accuracy of pile capacity prediction.

REFERENCES API 2000. Recommended Practice for Planning Designing and Constructing Fixed Offshore Platforms – Working Stress Design. API 2007. Errata and Supplement 3 – API Recommended Practice 2A-WSD, Recommended Practice for Planning, Designing, Constructing Fixed Offshore Platforms – Working Stress Design. Clausen, C.J.F., Aas, P.M. & Karlsrud, K. 2005. Bearing Capacity of Driven Piles in Sand, the NGI Approach.

© 2011 by Taylor & Francis Group, LLC

In Taylor & Francis (Eds.), 1st International Symposium Frontiers in Offshore Geotechnics: ISFOG 2005. GL 2005. Germanischer Lloyd Rules and Guidelines, IV Industrial Services, Guideline for the Certification of Offshore Wind Turbines. Germanischer Lloyd Wind Energie GmbH, Hamburg/Germany, Edition 2005. Jamiolkowski, M., Lo Presti, D.C.F. & Manssero, M. 2001. Evaluation of relative density and shear strength of sands from CPT and DMT. In American Society of Civil Engineers (Ed.), Soil Behavior and Soft Ground Construction, Geotechnical Special Publication No. 119. Jardine, R., Chow, F.C., Overy, R. & Standing, J. 2005. ICP Design Methods for Driven Piles in Sands and Clays. Tomas Telford. London. Kolk, H.J. & Baaijens, A.E. 2005. Design criteria for pipe piles in silica sands. Taylor & Francis (Eds.), 1st International Symposium Frontiers in Offshore Geotechnics: ISFOG 2005. Lammertz, P. 2008. Ermittlung der Tragfähigkeit vibrierter Stahlrohrpfähle in nichtbindigen Boden. Essen: VGE Verlag GmbH. Lehane, B.M., Schneider, J.A. & Xu, X. 2005a. A review of design methods for offshore driven piles in siliceous sand. Lehane, B.M., Schneider, J.A. & Xu, X. 2005b. CPT Based Design of Driven Piles in Sand for Offshore Structures. Schneider, J.A. 2007. Analysis of Piezocone Data for Displacement Pile Design. Schneider, J.A., Xu, X. & Lehane, B.M. 2008. Database Assessment of CPT-Based Design Methods for Axial Capacity of Driven Piles in Silicious Sands. American Society of Civil Engineers (Ed.), Journal of Geotechnical and Geoenvironmental Engineering. Toolan, F.E., Lings, M.L. & Mirza U.A., 1990. An appraisal of API RP2A recommendations for determining skin friction of piles in Sand. Houston, Texas: Proceedings of the 22nd Annual Offshore Technology Conference.

568

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Installation of suction caissons for offshore renewable energy structures O.J. Cotter, B.W. Byrne & G.T. Houlsby Department of Engineering Science, University of Oxford, UK

ABSTRACT: Suction caisson foundations are a promising option for a range of offshore renewable energy structures. Currently they are being considered for offshore wind turbines either as a mono-foundation or as the foundations for more conventional multi-footing jacket structures. Installation of the caissons is important and so a thorough understanding of the installation mechanisms in a range of soils and the consequent effect on capacity must be obtained. This paper explores the important installation issues to consider for a range of sites around the UK and summarises some of the findings of a recent research project looking at installation processes (Cotter, 2009). This includes experimental results ranging from installation into homogeneous sand, homogeneous clay, layered soil profiles, the effect of pressure grouting on caisson response and the use of skirt tip injection to aid installation.

1

2

INTRODUCTION

Offshore renewable energy structures, such as wind turbines or tidal turbines, might use suction caissons as foundations. One of the key processes is the design for the installation as this is where substantial cost savings could be made. A number of different approaches for assessing the installation suction, and other associated installation parameters, are available but no common standard has yet been agreed. For the technology to be widely employed, robust installation behaviour must be demonstrated, and must be shown to agree with the different approaches proposed. The soil conditions around the UK vary significantly, ranging from homogeneous sand and clay, through to complex layered materials. A recent research project, carried out at Oxford University, has explored suction caisson installation behaviour for a range of these different site conditions (Cotter, 2009). This will enable designers to assess where caisson foundations are appropriate and also where they are not. This paper provides a summary of the outcomes from that project and more details can be found in Cotter (2009). Initially, dimensional analysis was carried out to determine the key dimensionless groups controlling the problem. The experimental work was designed, using these dimensionless groups, to be representative of full scale behaviour at a range of sites around the UK coastline. The work described in this paper covers model caisson installation in homogeneous soil profiles before describing the effect of skirt tip injection on installation suction and to control caisson verticality in sand. The paper then reports work undertaken in layered soils examining sand over clay, sand with a thin clay lense and sand over inclined clay, before considering clay over sand. Finally the paper will report the effect of pressure grouting the underside of the caisson lid. © 2011 by Taylor & Francis Group, LLC

EQUIPMENT

A range of apparatus was developed for this project. Pumping was achieved using high quality gear-pumps, manufactured by Ismatec. The pump speed was computer controlled via an RS232 link allowing accurate flow rate control. Pressure measurements were made using Druck pressure transducers from which suction could be calculated, and displacements were measured with LVDTs. The computer used to control the gearpump(s) was also used to log the data via a 16 bit A-D card and a program written in Labview. This program incorporated feed-back control loops to vary the pumping speed to maintain a target installation rate. However, the pumps could also be manually controlled from the program when necessary. Saturated soil samples were prepared by consolidation from slurry in the case of clay (standard laboratory kaolin), and by the application of an upward hydraulic gradient and vibration for sand (a fine grained silica sand). Site investigation data was gathered in the form of undrained shear strengths (obtained from miniature vane tests) for the clay and relative densities for sand. The caisson self weight was controlled with a counterbalance system. The mass of water pumped from the caissons was recorded to measure the flow rates present during the experiment. During installation, the caisson was initially installed under self-weight (i.e. with no suction), after which installation proceeded only by the application of suction. Care was taken when transitioning from the self weight installation phase to the suction installation phase in order to minimise any disturbance to the soil sample. For all experiments, caissons were free to move vertically with lateral restraint, apart from the steering tests, for which lateral movements were possible. Caisson sizes varied from 150 to 570 mm diameter and wall

569

Figure 1. A typical suction installation curve, in sand, that shows a close match with theory.

Figure 2. Suction installation curves for penetration with varying degrees of skirt tip injection.

thickness to diameter ratios varied between 0.2% and 1%. Further details of the experimental set-up for each set of experiments, including soil data, can be found in Cotter (2009).

skirt tip pressure to be measured. The recorded data was then compared to published estimates (Houlsby and Byrne, 2005b) and it was found that the value of a varied with installation rate. Higher penetration rates caused a lower reduction of a with depth than slower installation rates. The discharged water volumes were also measured during installation. From these data, and the recorded suction, it was concluded that the overall plug permeability increased with installation depth. The overall plug permeability was also greater during faster rates of penetration, which supported the conclusions made in relation to the measurement of a. Despite these variations, the influence of penetration rate on suction required was not significant over the range of rates used. Installations were undertaken at rates in the range of 0.02 and 1.0 mm/s. For caisson installation in homogeneous clay it was found that the clay resistance could be accurately calculated and it was noted that negligible seepage occurred during suction installation.

3

HOMOGENEOUS SOIL

The initial experiments were conducted on homogeneous materials in order to assess known calculation approaches such as those presented by Houlsby and Byrne (2005a and b) or Andersen et al. (2008). Two types of tests were carried out. The first consisted of jacking the caisson into the soil to measure the overall resistance to penetration in the absence of suction. From these tests it was possible, in combination with information derived from the site investigation, to deduce the range of parameters to be used in the calculation methods. The second consisted of installation using suction to assess whether the parameters describing the process may need to be modified. Figure 1 shows a typical suction penetration curve for a caisson in sand. The curve shows an initial penetration under self-weight (i.e. no suction) followed by further penetration under the application of suction. Also plotted is an estimate of the suction required using the approach of Houlsby and Byrne (2005b). This shows that it is possible to obtain a very close agreement: the match shown makes use of estimated soil properties, consistent with the site investigation data. One of the key parameters in the test is the ratio of suction at the skirt tip to that under the base (denoted here as the parameter a). This may be affected by relative density changes of the soil within the caisson skirt. A modified caisson was used which allowed the © 2011 by Taylor & Francis Group, LLC

3.1

Sand – Skirt tip injection and steering

It has been suggested that water injection at the skirt tip may aid installation in sand. In particular by varying the water pressure at the different injection zones, it is thought that caisson verticality can be controlled during installation. To study these ideas, two sets of tests were undertaken. The first test series explored the effect of a constant increase of water pressure by injection around the skirt tip (also described in Cotter et al. (2008)). Two configurations were assessed, one with 8 nozzles equally spaced around the skirt perimeter and a second with 16 nozzles. Figure 2 shows a plot of a selection of the results for the 16 nozzle tests. The injection pressure at all of the nozzles was

570

Figure 3. Installation suction and caisson rotation plotted against time showing effect of steering.

controlled by varying the rate of pumping. The suction required for installation reduced as the injection pressure increased, with the magnitude of the reduction appearing to be constant with depth. Tests using the caisson with 16 nozzles required less suction for installation than those with 8 nozzles when the same injection pressures were applied. The use of injection pressure was found to cause piping at shallower penetration depths, and this also limited the maximum injection pressure throughout the test. It became clear that the injection pressure measured rapidly reduced with radial distance away from the source. The injection points were therefore approximated as point sources and the hydraulic gradients resulting from these point sources were used to assess the total pressure changes in the sand during installation. The main limitation of the testing was the size of the pipe-work required, which dictated that the water pressure at the skirt tip could not be modified at a greater number of locations. For larger caissons this scale issue will be less significant so that injection could be undertaken at more locations, and the pore pressure change around the skirt tip will become more uniform. This would enhance the effect of skirt tip injection. Caisson steering was attempted by changing the water pressures at the skirt tip. For example it is likely that a differential pressure across the caisson skirt will lead to caisson rotation. This was explored by supplying an injection pressure to nozzles on one side of the caisson and monitoring the caisson response. A typical test is shown in Figure 3 where θy and suction is plotted on a common time axis. The periods where there is no suction applied correspond to the phases where the caisson did not move and θy remained constant. Therefore, as the caisson was stable before the start of a steering phase, any change in angle resulted from the application of steering and was not due to any prior © 2011 by Taylor & Francis Group, LLC

Figure 4. Plot of results from uplift experiments. V
is the applied effective vertical load and α is the adhesion factor.

instability. Figure 3 shows that it does appear possible to steer the caisson in a controllable fashion. The tests demonstrated that this method was most effective at the shallow depths. The angle through which the caisson could be steered was found to be proportional to the injection pressure applied at the nozzles.

4

LAYERED SOILS

For many sites, the soil profile will consist of sand interspersed with clay or vice versa. There is very limited information in the literature about caisson performance in layered soils.

4.1

Clay overlying sand

Clay overlying sand is a soil condition identified as being present around the coast of the UK. To explore this profile, experiments were carried out in samples of clay over sand, some of which have been reported by Cotter et al. (2008). The key issue identified from these tests was that during suction installation, plug uplift can occur, while the skirt tip is still within the clay layer. Dimensional analysis was carried out to assess typical sites around the UK and potential wind turbine foundation designs. These are plotted in the space defined by su /γ
D and V
/γ
D3 shown as the shaded area on Figure 4. A number of experiments were then carried out to explore this area and the results are plotted confirming the uplift boundary derived from theoretical considerations. The variation of the uplift boundary due to changes of α is also shown. Ding et al. (2001) described the installation of a 9 m diameter caisson into clay over sand. Sufficient information was provided to determine the non-dimensional installation conditions, which are

571

Figure 5. Installation suction plotted against displacement indicating plug uplift.

also included in Figure 4. It can be observed that the installation plots beneath the boundary for uplift. As the caisson was installed without plug lift, the result appears to agree with the expected outcome from the uplift model. Figure 5 presents data for an experiment which ended with plug lift. The recorded suction and an estimate of the suction required for both installation and for plug uplift are shown. It is observed that the suction measured followed the estimate until the suction for uplift was approached, at which point the suction exceeded that estimated for installation. When the suction reached a value at which uplift was predicted, plug lift occurred and installation stalled. For experiments ending with plug lift, lower suction pressures were observed while the plug was drawn up into the caisson. In these experiments the penetration of the caisson ceased, and it was not possible to install the caisson any further using suction. 4.2

Sand overlying clay

A thin layer of sand overlying clay should generally not present any substantial installation problems. The seepage gradients required for the sand installation will be available at the start and the force required for jacking into the clay (once through the sand) can usually be developed from the supplied under pressure. 4.2.1 Sand with a thin clay layer Problems may develop where otherwise homogeneous sand contains a relatively thin clay layer. The caisson is likely to penetrate the sand and into the clay layer, but © 2011 by Taylor & Francis Group, LLC

572

Figure 6. Installation suction plotted against displacement indicating plug uplift.

might experience difficulty installing into the bottom sand material. It is thought that the thin clay layer may prevent the seepage flows from developing. Figure 6 shows a typical result from a test exploring this issue. As expected the skirt tip moved through the top sand layer and the installation suction increased with depth. When the skirt tip began penetration in clay, the suction increased substantially to the level required to force the skirt through the clay. Whilst the skirt tip was in the clay layer, at about the midpoint in this case, the suction decreased rapidly. After the rapid loss of pressure, installation was still possible but the suction was much reduced. The installation continued to the base of the clay layer, and then into the sand beneath. Also shown on Figure 6 are estimates of the jacking force required to install the caisson, the suction pressure required during the sand only phase and the suction that would cause clay plug uplift. It was clear that uplift of the plug had occurred whilst the skirt tip was in the clay and in all the experiments undertaken evidence of uplift was observed. Although uplift occurred it was clear that the clay plug did not remain intact within the caisson as water flow from the underlying sand layer was measured. This water flow allowed seepage gradients to develop, as the caisson penetrated the bottom sand stratum. 4.2.2 Inclined clay substratum In practice, it is unlikely that sand will overlie a perfectly flat clay horizon. An inclined clay layer therefore presents an interesting installation issue, particularly if the inclination is steep. As the skirt tip approaches the highest point of the clay layer, seepage will gradually

Figure 8. Moment load and rotation behaviour for a caisson grouted using high grout pressures.

Figure 7. Installation suction plotted against displacement for a test on an inclined clay sub-stratum.

become cut off. At deep penetration depths, if seepage can be significantly reduced, the sand resistances are anticipated to become relatively high. On the other hand, if seepage flows can still be maintained throughout the plug, despite the entry of water being partly blocked by the clay, installation may still be possible. The suction measured during installation in a typical test exploring this problem is shown in Figure 7. The final depth of clay is also shown: this depth corresponds to the point at which the skirt tip becomes completely installed in clay (the maximum penetration distance of the skirt tip in sand). As the clay inclination was 33.2◦ , the transition depth was 119 mm. At shallow depths, suction was similar to that which would have been estimated for homogeneous sand installation. Deviations from the homogeneous sand estimate occurred, progressively, as the skirt tip approached the sand/clay boundary. Full installation was possible even when conditions had been chosen to cause piping failure in the sand before the skirt tip had fully entered the clay.

5

EFFECT OF GROUTING

After installation, it is likely that under-base pressuregrouting will be carried out in order to ensure a uniform contact stress across the base of the caisson. The effect of under-base grouting, at low and high pressures, on the moment load response was investigated during this project for installations in sand. Figure 8 presents some typical cyclic moment load responses, under a constant vertical load, of a suction installed caisson that was pressure grouted using a high grout © 2011 by Taylor & Francis Group, LLC

573

Figure 9. Moment stiffness response for different underbase conditions.

pressure. The results of jacked and suction installed (un-grouted) caissons are also shown for comparison. It can be seen that at large rotations the high pressure grouted caisson was significantly stiffer than the suction installed or jacked caissons. Figure 9 shows a plot of normalised unloading stiffness for a range of caisson installation conditions. For the tests shown, the unloading stiffness of the suction installed caisson was lower than the jacked in place caisson at all rotations. At large rotations, grouted caissons were stiffer than the jacked caisson, but at small rotations the stiffness was recorded to lie between the suction installed caisson and the jacked caisson. The vertical displacements during cyclic moment loading were also compared for the different installation processes. For installations where grouting had been undertaken, the accumulated vertical settlement

(d) Tests in clay over sand demonstrated that, under certain conditions, plug uplift can occur and that caisson refusal is a definite risk and should be considered at the design stage. (e) Installation into sand containing a thin clay layer was shown to be possible even though plug uplift was observed during the clay penetration phase. (f) Installations into sand overlying an inclined clay substratum were successful. (g) High-pressure grouting improved the moment stiffness of caissons under cyclic loading. (h) Pressure grouting, using any pressure, was found to reduce the vertical settlement of the caisson under cyclic moment loading. The results of this work demonstrate conditions where caissons can be used successfully and also areas where the engineer must consider the design much more carefully. Further details of the work can be found in Cotter (2009). Figure 10. Vertical displacement response for different underbase conditions.

ACKNOWLEDGEMENTS was significantly reduced irrespective of the pressure used to grout the caisson. A plot of accumulated displacement during cyclic loading is shown in Figure 10 for caissons grouted at low pressures as well as for un-grouted suction installed and jacked caissons. 6

CONCLUSIONS

This paper has summarised the results of a project on the installation of suction caissons for renewable energy problems. The project was mainly experimental though dimensional analysis was carried out to ensure that the experimental work was relevant to conditions around the UK coastline. Some of the conclusions were: (a) The response in homogeneous soil was consistent with previous work, including that of Houlsby and Byrne (2005a and b) (b) Skirt tip injection was shown to reduce the suction pressure required for installation. (c) Differential skirt tip injection was shown to be effective at controlling the verticality of the caisson during installation.

© 2011 by Taylor & Francis Group, LLC

The first author wishes to acknowledge support provided by Magdalen College, The Department of Engineering Science, Fugro Ltd and the EPSRC. REFERENCES Andersen, K. and Jostad, H. P. and Dyvik, R. (2008). Penetration Resistance of Offshore Skirted Foundations and Anchors in Dense Sand. Journal of Geotechnical and Geoenvironmental Engineering 134 1: 106–116 Cotter, O.J., Houlsby, G.T. and Byrne, B.W. (2008). Installation of Suction Caisson Foundations into Clay over Sand. Proceedings of the BGA International Conference on Foundations, Dundee, Scotland, 24–27 June. Cotter, O.J. (2009). The installation of suction caisson foundations for offshore renewable energy structures. DPhil Thesis, The University of Oxford. Ding, H., Qi, L. and Xu, J. (2001). Bucket foundation platforms installed in shallow and ice-drifting area. Journal of Cold Regions Engineering 15 4: 211–218. Houlsby, G.T. and Byrne, B.W. (2005a). Calculation procedures for installation of suction caissons in clay and other soils. Proc ICE – Geotechnical Engineering 158 2: 75–82. Houlsby, G.T. and Byrne, B.W. (2005b). Calculation procedures for installation of suction caissons in sand. Proc ICE – Geotechnical Engineering 158 3: 135–144.

574

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Lateral behaviour of large diameter monopiles at Sheringham Shoal Wind Farm L. Hamre & S. Feizi Khankandi Det Norske Veritas (DNV), Norway

P.J. Strøm Statoil ASA, Norway

C. Athanasiu Norwegian University of Science and Technology (NTNU), Norway

ABSTRACT: There are very sound reasons for the growth of the offshore wind industry. Both the UK and Europe as a whole possess considerable wind resources as measured in mean wind speed and consistency. Most of the existing wind energy turbines in the North Sea are founded on monopiles, and the most common geotechnical design method for these piles is the p-y concept according to the API regulations. However, the validity of this method is not proved for large diameter piles with predominantly lateral loading. In this paper the relation between the API model and a more advanced Finite Difference method is investigated for one location at the Sheringham Shoal Offshore Wind Farm site. The analyses were carried out using the computer programs SPLICE and FLAC 3D. The analyses show that based on the same geotechnical properties, the traditional API method gives larger deformation than the Finite Difference code using the softening-hardening soil model. 1

INTRODUCTION

1.1 General Energy consumption and environment pollution is one the most concerns and discussions these days. However, most part of the world energy consumption is based on the fossil fuels such as oil, gas and coal. Replacement of these power supplies to renewable energy resources such as sun, wind and water can reduce dangerous greenhouse gases. Modern wind turbines range from around 600 kW to 6 MW of rated power, although turbines with rated output of 1.5–3 MW have until now been the most common for commercial use. Areas where winds are stronger and more constant, such as offshore and high altitude sites are preferred locations for wind farms. Globally, the long-term technical potential of wind energy is believed to be five times total current global energy production, or 40 times current electricity demand. This could require large amounts of land to be used for wind turbines, particularly in areas of higher wind resources, or increased development of offshore sites. As might be expected, offshore wind farms are more expensive to construct, and optimised design with respect to material use, construction and installation is therefore sought after. But they have certain distinct advantages over land. The mean wind speed at sea is for example proportionally higher than on land and roughness and turbulence are less. © 2011 by Taylor & Francis Group, LLC

Most of the existing WTGs (Wind Turbine Generators) in North Sea are founded on monopiles, as are the foundations for the Sheringham Shoal Offshore Wind Farm. The site is located 20 km north of the Norfolk coastline 15–23 m water depth. A total of 88 WTGs will be installed, starting spring 2010. The Sheringham Shoal monopiles have diameters of 4.7 m to 5.7 m, and penetrations between 23 m and 37 m. 1.2

Background

The most commonly used method for calculating lateral resistance of monopiles intended for WTGs is the API model, as described in several design codes (e.g. API RP 2A, DNV-OS-J101). This API procedure is an empirical method based on back analyses of long, slender jacket piles in soft soil (for instance Gulf of Mexico). The piles that are included in the API test database are predominantly governed by vertical failure (skin friction/tip resistance), and the p-y curves are intended to model horizontal support to the pile and thus reduce pile stresses. But they are not intended for use in the soil capacity evaluation, especially when small deformations are a concern, as is the case for WTG’s. Also the diameter/pile length relation for the API-piles is much smaller than for commonly used monopiles for WTG’s. Therefore alternative analyses solutions may be justified. This study is intended to be an example of how advanced analyses can support a more optimistic

575

Table 1.

Summary of key soil parameters.

Soil unit –

Depth interval m

Unit weight kN/m3

IP %

Dr %

BCT BDK

0–5.0 0–7 7–12 7–18 15–40 40–62 10–70

17 21.3

30 17

– –

20.3 21.3 20 19.5

– 15 25 6

80–100 –

– 200–1100

– 40.5 36.0 44 33.5

–

220–950

35.5–41

EG SBK CK

choice of soil parameters/model assumptions in the design of large diameter monopiles.

Undrained shear strength (Su) kPa 20–60 0–250

Table 2. The small strain shear modulus, Gmax . Soil unit –

Depth (m)

Lower boundary Gmax /sCu

Upper boundary Gmax /sCu

BDK SBK CK

0–10 18–34 12–18 44

200 200–1200 200 3000

800 600–2600 800 4000

1.3 Analyses approach The performed analyses consist of the following steps (see also Figure 6): 1. Beam model with non linear API (2000) soil springs. (as included in the computer program SPLICE, DNV, 1994) 2. Beam model with non-linear springs according to DSPY curves (Achmus, 2005 & 2007) 3. Beam model with non linear API soil springs including large diameter effects according to Stevens et al. (1997) 4. 3D Finite Difference analyses (computer program FLAC3D, Itasca 2009) with Mohr-Coulomb model and G-modulus corresponding to stress level at half the maximum stress. 5. 3D Finite Difference analyses with softeninghardening soil model with G-modulus degradation according to Seed & Idriss adjusted to site specific soil test results at the Sheringham Shoal field (Kramer, S.L, 1996, Gardline 2009). 6. 3D Finite Difference analyses including dynamic effects and cyclic degradation properties (Gardline 2009, Achmus, 2005 & 2007). This paper presents a study of one monopole intended for position D7 at the Sheringham Shoal Offshore wind farm. 2

SOIL DATA

2.1 Sheringham Shoal soil conditions The soil conditions at the Sheringham Shoal Offshore Wind Farm site can generally be described to comprise a veneer of Holocene gravelly sands overlying firm to stiff, sandy clays of the Bolders Bank Formation (BDK). The normally consolidated Botney Cut Formation (BCT), comprising laminated clays and silts, peats and fine sands, is interpreted to fill two north-south trending channels and a smaller northwest-southeast trending channel cut into the underlying Bolders Bank Formation. The Egmond Ground Formation (EG) consisting of dense to very dense sand overlies the Swarte © 2011 by Taylor & Francis Group, LLC


deg

576

Bank Formation (SBK), which typically comprises stiff to hard, sandy-gravelly clay, and fills valleys that have been cut into the underlying Chalk Group (CK). Descriptions for the five main units mentioned above are given in Gardline report to Statoil (Gardline 2009), and a summary of key parameters for each soil unit is given in Table 1. The UU tests carried out at the field reveal strain values of strain at failure in the order of 2.5–4%. This was in design considered to be unrealisicity high, and a value of 1% was selected for the design. This value was based on the triaxial tests (Rambøll 2009). The basis for the empirical method suggested by the API recommended practice is UU tests, hence any other choice of test method needs documentation. One of the tasks of the independent analyses carried out by DNV and described in this paper is to validate the use of ε50 from CAUC tests. The small strain shear modulus, Gmax , has been investigated in-situ using PS-logging techniques and in the laboratory using Bender Elements. Following lines presents the normalized parameter Gmax /sCu giving recommended low and high estimate profiles for the BDK, SBK and CK units (Table 2). 2.2

Soil models used in the analyses

In order to be able to compare the results from the different analyses approaches, the same soil background was used in all the analyses. • In the API model, the CAUC tests were used instead

of the UU test results. One of the governing parameters for this model is the ε50 parameter, found to be 1% for relevant triaxial CAUC tests in soil layers.

Figure 2. Applied moment time history. Extract from the 600 s series used in the analyses.

load cycles during the design life of the structure is very high. In the analyses, the dynamic loading was taken as one 600 s period that includes at least one of the highest load amplitudes in the time series. The maximum was taken as for the static loads, as indicated in Figure 2. However, for the cyclic degradation of the soil, a longer time period must be applied. The composition of the design cyclic loading governing for geotechnical evaluation including corresponding pore pressure build up and dissipation is not investigated in detail, but based on an engineering judgment, a soil degradation relevant for a load situation with Neq = 10 cycles with maximum load level was included in the design.

Figure 1. Deformation contour of the monopile. • The FLAC 3D Mohr Coulomb model was based

on the same triaxial tests, including a linear E50 modulus established from ε50 = 1%. • The FLAC 3D Hardening – Softening soil model was based on the same triaxial tests as described above, with the real stress strain behavior from these tests. • The cyclic degradation was evaluated based on advanced cyclic testing carried out for relevant soil layers (Gardline 2009). The test results were used in order to establish relevant parameters for use in Achmus (2005 & 2007). 2.3 Detailed soil conditions for position D7

3.3 Monopile make up The design support structure for position D7 consist of a cylindrical monopile with a diameter of 5.7 m and a penetration depth of 35 m with varied pile wall thickness with depth. In the FLAC 3D model, the most efficient way of modeling the pile is with solid elements. Since moment loading will govern the pile behavior, a pile model with equal rotational stiffness (EI) to the real pile was used in the design.

The analyses were carried out for the monopole intended for the D7 position. The soils at this position consist of different layers presented in Figure 1.

4

METHODOLOGY

4.1 General 3

LOADS AND MONOPILE GEOMETRY

3.1 Static loading The analyses were based on an example load case as given below: Horizontal load: H = 9 500 kN Overturning moment: M = 370 000 kNm The loads give quite high utilization of the soil; hence the non-linear effect will be clear in the analyses. For a real design situation, the loads may be lower, ant thus the non-linear effect less evident. 3.2

Dynamic loading

The dynamic loading on a WTG is governed by both wind and wave components, and the total number of © 2011 by Taylor & Francis Group, LLC

Two different computer programs, SPLICE and FLAC 3D are used for the static and dynamic analysis. In the following section the description of type of analysis and material modeling is given. 4.2 SPLICE analysis In the first part of the study, analyses were carried out based on a beam model including closed form solutions for p-y curves as described in the API recommended practice (2000). However, the strain at half the maximum stress was taken as ε50 to 1%, based on CAUC tests, not the UU tests. The effect of cyclic degradation of soil as described by Achmus (2005 & 2007) as included in the so-called DSPY model was investigated. The results

577

from this evaluation were compared with the results from cyclic laboratory tests for Sheringham Shoal soils (Gardline 2009), and found to compare well when the parameters used in the Achmus model, (pN = p1 N (α−1)t , yN = y1 N (α)t ), of α = 0.2 and t = 0.17 were applied. Effects of the large pile diameter were included using the relationship established by Stevens et al. (1997). The relationship between strain and deformation for a large diameter pile, yc = 1.4 ∗ε50 ∗ D0.5 (This formula is only valid for diameters given in meters). This gives an increase in p-y stiffness with a factor of approximately 4 for the given pile geometry compared to the traditional API-approch (yc = 2.5 ∗ ε50 ∗ D). Since the beam model with non-linear springs as included in the SPLICE computer program (DNV 1994) includes several empirical assumptions, additional more advanced analyses were carried out in order to document that the ε50 = 1% from CAUC tests can be used in the design. The loads given in Section 3 were used in the comparative analyses.

Figure 3. A view of monopile model in FLAC 3D.

degradation factor. A 600 s time period used in the analyses is given in Section 3 (Figure 2). The pile is modeled from the tip elevation to the top elevation at 1m above seabed. A view of half the modeled pile and the surrounding soil is shown in Figure 3.

4.3 FLAC 3D analysis In the second part of analysis, the monopile was modeled in FLAC 3D. FLAC 3D is an acknowledged program in the geotechnical engineering field. It is a finite difference code consisting of different constitutive soil behavior models. The approach based on FD analysis takes into account the initial conditions, nonlinear pile-soil interaction and nonlinear soil behavior. In the FE and FD programs such as FLAC 3D, definition of a suitable interface around the pile and its tip and the soil degradation during the analysis gives a good picture of the theoretical pile/soil behavior. The aim of these analyses was to investigate the behavior of a large monopile in principle and to check whether the API method can be used including strain relations from triaxial CAUC tests rather than the UU tests. The material properties for the soil are based on the offshore geotechnical investigation carried out by Statoil (Gardline 2009) as summarized earlier. In the analyses, material properties of the surrounding soil were modified based on the degradation from shear mobilization. Two constitutive models were used; Mohr-Coulomb and Softening-Hardening models. First, a simple Mohr-Coulomb model was performed with static loads as given in Section 3. Advanced analyses were then performed with softening-hardening model of FLAC 3D. The strain- softening/hardening model allows representation of nonlinear material softening and hardening behavior (cohesion, friction, dilation) as functions of the deviatoric plastic strain. Degradation from cyclic loading was included in the dynamic load condition. Number of cycles, one or two way cycling and the loading amplitude and frequency are the main parameters for calculation of this © 2011 by Taylor & Francis Group, LLC

5 5.1

RESULTS Splice analysis results

The Splice analyses, including closed form solutions based on API (2000) for load deflection behavior give a lateral deflection at pile top of 16 cm. When a more recent model (Achmus 2005 & 2007), based on cyclic pile performance was applied, a reduction to 14 cm in pile head displacement was found. This is considered to be because the model is based on a different relation between strain and deformation, and not due to cyclic loading effects as such. The Splice model was also used with an approach where the effect of large diameter was included. The same p-y curves as for the first approach was used, but with a reduction coefficient on the lateral deflection, y, of 0.25 (Stevens 1997). For this situation, the pile top deflection was reduced to 8 cm. The pile deformation with depth is given in Figure 6.

578

5.2

FLAC analysis results

Figure 4 and Figure 5 shows how the model typically deforms under the applied loading. With the MohrColoumb model, the maximum horizontal deformation at pile head level is calculated to 1 cm. A maximum of 12 cm deformation was found from the softening-hardening constitutive model. In the cyclic load case (see Figure 2), time history of the horizontal load and moment is applied to the FLAC 3D model. This time history of applied moment has been described in section 3.2. The deformation response of the monopile top under cyclic loading was found to be 15 cm as given in Figure 6 and 7.

Figure 7. Displacement time history of the monopile top under cyclic loading.

Figure 4. Deformation contour of the monopile.

6

SUMMARY AND CONCLUSIONS

The main reason for this study was to document that the ε50 – value from the triaxial CAUC tests can be used in the design of large diameter monopiles. Different approaches were used, including a dynamic FD model including soil degradation from cyclic loading. The following main results were found from the analyses: • The API closed form p-y curve concept gives the

Figure 5. Displacement vectors. •

•

•

•

•

Figure 6. Summary of monopile deformation.

© 2011 by Taylor & Francis Group, LLC

largest deformation even when ε50 – values from the triaxial CAUC tests were applied. A more resent model including cyclic degradation of soil give less deformation than the API model with ε50 – values from the triaxial CAUC tests. The FD model with Mohr-Coloumb soil model and E50 values from triaxial CAUC tests show very small pile top displacements. The reason for this is that the stresses exceed half the maximum stress in the top layers, and E50 may therefore not be representative in these layers. The FD analyses with softening/hardening model give pile top deflections similar to the API model including “large diameter effects”, however the rotation is larger due to stiffer soil response in the deeper layers in the FD model. Including cyclic soil degradation effects in the dynamic FD analyses with softening/hardening soil give pile top reactions similar to the API model. A more detailed evaluation of both dynamic loads and cyclic degredation of soil should be carried out before a conclusion on this effect is drawn.

The study is based on several analyses for one WTG position at the Sheringham Shoal Wind Farm site. It shows that the pile top deflection for a large diameter monopole may be overestimated if a traditional APImodel is applied. The dynamic loading including cyclic soil degradation is not governing for this specific site since the API model with ε50 – values from the triaxial CAUC tests was used in the design. However, if more sophisticated soil models and analyses are applied further work on this issue is recommended.

579

REFERENCES Achmus M., Abdel-Rahman K. & Kuo Y., Behaviour of large diameter monopiles under cyclic horizontal loading, 12th international colloquium on structural and geotechnical engineering (ICSGE), Egypt, 2007 Achmus M. & Abdel-Rahman K., On the design of monopile foundations with respect to static and quasi-static cyclic loading, Appelstrabe 9 A, D-30167, Germany, 2005 API, ‘Recommended Rractice 2A-WSD (RP-2A-WSD)’, ERRATA and supplements, 2007 DNV Classification Notes, No.30.4, Foundations, 1992 DNV-OS-J101: ‘Design of Offshore Wind Turbine Structures’, 2004 & 2007

© 2011 by Taylor & Francis Group, LLC

580

FLAC 3D (Itasca 2009), ‘Finite Difference code’, USA Gardline report, SH Doc. no. SC-00-NN-G15-00003 “Sheringham Shoal Offshore Windfarm Development. Geotechnical Investigations”, 2009 Rambøll, ‘Sheringham shoal offshore wind farm, design report, Geotechnical’, 2009 Kramer, S. L, ‘Geotechnical earthquke engineering’ Prentice Hall, 1996 SESAM package, ‘Gensod, Pilgen and Splice’, DNV, 1994 Stevenes J.B. & Audibert J.M.E., ‘Re-Examination of p-y curves formulation’, OTC 3402, USA, 1997

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Centrifuge modelling of offshore monopile foundation R.T. Klinkvort & O. Hededal Technical University of Denmark, Copenhagen, Denmark

ABSTRACT: Today one of the most used concepts for wind turbine foundation is the monopile. The foundation concepts for these monopiles on deeper water is uncertain and consequently the design needs to be conservative leading to uneconomic designs. This paper describes a total number of 6 static and 5 cyclic centrifuge tests on a laterally loaded monopile in dry sand. The prototype dimension of the piles was modelled to a diameter of 1 meter and penetration depth on 6 meter. The test series were designed in order to investigate the scaling laws in the centrifuge both for monotonic and cyclic loading. It was not possible in the tests to reproduce the same prototype response for both the monotonic and the cyclic loading. It was not clear if this scatter in prototype data was due to normal measurement uncertainties or if the response is depending on the scaling factor. 1

INTRODUCTION

Single large diameter tubular steel piles commonly denoted monopiles is today a very used foundation method for offshore wind turbines. The design of these monopiles is commonly based on the theory of laterally loaded piles which relies on empirical data originated from the oil and gas industry, Reese and Matlock (1956) & McClelland and Focht (1958). The lateral capacity is determined by modelling the pile as a beam and the soil as a system of uncoupled springs, this is known as aWinkler model. The springs are described by p-y curves defining the load-displacement relationship for the interaction between soil and pile, API (1993). The formulation of these curves was originally calibrated to slender piles, but is today even used for design of large diameter monopiles with a slenderness ratio L/D as low as 5. The monopiles used for wind turbine foundations thus act as stiff piles. Therefore it is relevant to investigate the behavior of stiff piles in more detail. The tests series presented in this paper is an initial program that intends to investigate the response of model monopiles subjected to different artificial gravities in a centrifuge. The concept called modelling of models is used to investigate the response from five different piles which are scaled to the same prototype dimensions. 2

by the angular rotation speed (ω) and the distance (R) from the rotational axis. The ratio between gravity (g) and artificial gravity is described by the gravity scale factor (N ).

In centrifuge modelling two key issues are represented, the scaling laws and the scaling errors. 2.1

To transform results from test carried out on models to prototypes the dimensional analysis can be used, Langhaar (1951). The foundation for the dimensional analysis is Buckingham’s  theorem. From this, dimensionless parameters can be determined. These dimensionless parameters have to be the same for the prototype and the model to have full similarity. If all governing laws of similitude are in place a true model is obtained. This implies that stresses and strains are scaled by a factor of 1, deflection and lengths is scaled by a factor of N , forces are scaled by a factor of N 2 and so on; see e.g. Taylor (1995).

CENTRIFUGE MODELLING

When performing centrifuge tests an artificial gravity is applied to a model test setup. This is done to ensure that the stress field in the model is similar to the stress field in the prototype. This is important in model testing due to the non-linearity of the stressstrain relations of soils. To apply the artificial gravity the model is placed at the end of a rotating arm. The acceleration in a specified point in the model is given © 2011 by Taylor & Francis Group, LLC

Scaling laws

581

2.2

Scale effects

In physical modelling it is seldom possible to produce a model where all details of the model is scaled correctly in the prototype. Therefore some approximations have to be made. These differences are called scale effects and are important to be aware of when the test results are interpreted. Model studies are not perfect and it is important to understand this. Two main effects will be presented here. The first is the stress distribution. Looking at Equation 1 on the preceding page it can be seen that the applied gravity is depending on the distance to the rotational axis.

Table 1. d [mm]

e [mm]

L [mm]

N [–]

16 22 28 34 40

40 55 70 85 100

96 132 168 204 240

62.5 45.5 35.7 29.4 25

Table 2. sand.

This distance will increase through the model. In the prototype the stresses will increase linearly due to the constant gravity field, whereas the stresses in the model will increase parabolically. To minimize this error the radius is defined from center to a depth of 2/3 of the pile penetration depth, Stuit (1995). When performing centrifuge modelling it is not possible to scale the sand grain diameter correctly, since this will imply a difference in friction angle and cohesion. Therefore, when considering bearing capacity, it is most often necessary to use the same sand in the model as in the prototype. This causes the sand grains to be scaled by a factor of N in the model. This is known as the particle size effect. The grain size effect has been investigated with “modelling of models”. Particle size effect has been tested for laterally loaded piles by Hoadley et al. (1981) and they found that a “model diameter/ grain size diameter” ratio of 50 and above gave a good agreement. Remaud et al. (1998) found that a ratio over 60 was enough to avoid particle size effects. Both of these studies were performed on long slender piles. Nunez et al. (1988) performed modelling of models on tension piles. They found that the smaller piles tested at high accelerations gave consistently higher capacity than larger piles tested at smaller accelerations.They explain this difference with installation effects and differences in wall thickness and conclude that the effect from particle size is not significant. EXPERIMENTS

As the first of a larger test series on monopiles a series of modelling of models have been performed to analyze the response of a monopile in relation to the applied gravity. The test program was performed on five solid steel piles with a diameter between 16– 40 mm and penetration depths between 96–240 mm which were all scaled to a prototype pile with a diameter of d = 1 m and penetration depth L = 6 m. In figure 1 a sketch of the test pile can be seen. In Table 1 the dimension of the five piles and the scaling © 2011 by Taylor & Francis Group, LLC

Classification parameters for the Fontainebleau

Specific gravity of particles Minimum void ratio Maximum void ratio Average grain size Coefficient of uniformity

Figure 1. Sketch of pile.

3

Dimensions and scaling factor for the piles.

Gs emin emax d50 Cu

2.646 0.548 0.859 0.18 1.6

Table 3. Void ratio for the different tests. d [mm]

16

22

28

34

40

Monotonic Cyclic

0.58/0.57 0.59

0.58 0.56

0.57 0.56

0.59 0.58

0.56 0.55

factor is shown. This should scale all the piles to the same prototype pile. All monotonic and cyclic tests were performed in dry Fontainebleau sand. Leth et al. (2008) has collected classification parameters for the Fontainebleau sand which can be seen in table 2 on the next page. The average grain size of the Fontainebleau sand is 0.18 mm. With pile diameter ranging from 16 mm to 40 mm this leads to a “model diameter/ grain size diameter” ratio ranging from 88 to 189. The centrifuge at DTU uses a spot pouring hopper (SPH) for the preparation of the sand sample. Due to the geometry of the container and pile the sand is prepared using a circular travelling loop as described in Zhao et al. (2006). The sand is installed in a container with a inner diameter of 50 cm and a height of 49 cm. A new sample is prepared for each of the tests. CPT tests have been carried out to validate the pouring method. All these CPT tests showed the soil sample has a good homogeneity in the container. After the sand is prepared, the pile is installed at 1 g. It must be expected that the sand is compacted in a higher degree around the pile, for large piles than for small piles. When the tests are carried out it must be expected that the stresses in the sand is so high that potential preconsolidated areas disappears. Installing the pile at 1 g. is therefore intended to minimize the effects from the installation. A total of 11 centrifuge tests have been performed: six monotonic and five cyclic. For all the tests the relative density was found to vary in the range 0.8–0.94. A table with the different void ratios can be seen in table 3 on the following page. The relative densities

582

Figure 3. Normalized plot bearing capacity and initial stiffness as a function of the scaling factor.

Figure 2. Normalized plot with the five static tests.

are calculated by knowing the weight and the volume of the sand sample. The average value for both the static and cyclic tests is for the relative density ID = 0.924 and a void ratio of e = 0.57 leading to a triaxial frictional angle of φ = 38◦ .

d = 16 mm it can be seen that the maximum bearing capacity is increasing with the applied gravity. This could indicate that the linear scaling which is assumed is problematic. Interpretation method 2: The maximum bearing capacity and the initial stiffness is plotted on figure 3 against the scaling factor. The maximum capacity is found as the maximum value found on figure 2 and the initial stiffness is found at the point where the applied load is P = 0.1. This is shown on figure 2 as the black markings. From figure 3 it seems to be a clear linear relationship between the maximum bearing capacity and the scaling factor. Looking at the initial stiffness of the load deflection response no clear relationship is seen. The variance of the stiffness could though indicate that a constant stiffness from the tests could be expected. Here is also plotted the initial stiffness found from the cyclic testing which support this conclusion. Four of the piles were mounted with measuring of the pile head rotation, if the pile is assumed to behave as a rigid pile, the pile movement can be described according to equation 4.

3.1 Monotonic tests The force and deflection is normalized, to compare the general pile behavior. On the y-axis the normalized force is plotted. This is found as shown in equation 2.

On the x-axis the normalized deflection is plotted. This is shown in equation 3

In figure 2 the observation of the monotonic loading can be seen. Remember that all the test is scaled to same prototype and the response from the different tests should be identical. However a variation in the results can be seen. The test performed at 62.5 g showed a significantly high bearing capacity therefore a second test on the d = 16 mm was performed to validate the response. The second test confirmed the response. Interpretation method 1: Looking at figure 2 you could say that the pile with a diameter of d = 16 mm shows a much higher capacity than the other piles and thereby indicates that the pile diameter particle diameter is too small. If this pile is neglected an acceptable scatter of the results is obtained. From this a bearing capacity for the prototype pile could be expected to be Pmax ≈ 0.32. This will be called interpretation method 1. On figure 2 the bearing capacity according to Hansen (1961) is shown for three different frictional angles. This indicates small change in frictional angle can be the reason for this scatter. On the other hand using the result from the pile with a diameter of © 2011 by Taylor & Francis Group, LLC

The assumption of the pile behaves like a rigid pile is satisfied according to Poulos and Hull (1989) if the stiffness of the sand is lesser than Es = 35 MPa. If the pile should act as a slender pile then the soil stiffness should be over Es = 3090 MPa. Even if the stiffness of the sand is larger than 35 MPa it is expected that the pile will be located close to the rigid boundary. Therefore it is assumed that the pile behaves as a stiff pile. From this assumption the point of rotation can be found knowing the deflection of the pile u and the rotation θ. The normalize point of rotation measured from pile tip is plotted in figure 4. Due to practical reasons the rotation of the 16 mm pile could not be measured. All the piles shows that the normalized point of rotation is located below the pile tip at initial deflection and the pile is therefore sheared through the sand. After some deformation the rotation point moves up and is located

583

Table 4.

Load characteristics for the cyclic tests.

N

Pmono.,1

Pmono.,2

ζb,1

ζb,2

ζc

62.6 45.3 35.7 29.4 25.0

0.32 0.32 0.32 0.32 0.32

0.39 0.34 0.34 0.32 0.29

0.53 0.52 0.44 0.41 0.41

0.44 0.49 0.44 0.42 0.44

−0.04 −0.05 −0.02 −0.02 −0.10

Figure 4. Point of rotation.

in the pile where it is stabilized until failure. From figure 4 no clear relation between normalized point of rotation and scaling factor can be seen. It seems like for all the piles that the rotation point stabilizes around a value of 0.22 except the pile with a diameter of d = 28 mm which haves a lower rotation point that the others. Using the theory of Hansen (1961) the rotation point is calculated to 0.2 which is close to the observation. It seems like all the piles is moving in the same manner. 3.2

Figure 5. Schematic illustration of average deflection and secant stiffness.

minimum values of load and the deflection is found. From this the average deflection can be calculated as shown in equation 6 and the secant stiffness can be calculated as shown in equation 7.

Cyclic tests

The cyclic tests were performed with 500 force controlled cycles. To investigate the effects from cyclic loading this paper uses a method describe in LeBlanc (2009) to described the cyclic loading. The load characteristics are denoted ζb and ζc . They are determined as shown in equation 5.

Here Pmax and Pmin are the maximum and minimum applied force in the cyclic loading. Pmonotonic is the maximum bearing capacity found from the corresponding monotonic test. The amount of the applied load depends on the interpretation of the monotonic test. The cyclic loading was performed as individual tests, with five different maximum capacities according to the monotonic tests shown on figure 2 on the preceding page. It was the intension to perform the cyclic test with a ζb = 0.40 and a ζc = 0 but due to the control system it has not been possible to perform tests with exactly the same load characteristics. However the load characteristics can also be calculated assuming a constant bearing capacity for the monotonic tests. The characteristics of the cyclic loading for the tests series for the two types of interpretation can be seen on Figure 4. For the cyclic loading the accumulation of deflection and the change in secant stiffness is calculated. This is done as showed on figure 5. For every cycle the maximum and © 2011 by Taylor & Francis Group, LLC

The best fit to the accumulation of deflection was done with a power fit as proposed by Long and Vanneste (1994), cf. equation 8.

Here u0 is the accumulated deflection at the first cycle and α is an empirical coefficient which controls the shape of the curve. n is the number of cycles. The accumulated deflection for a given cycle is defined as the average value for the cycle. The values of the coefficient to the proposed formula can be seen in Table 5. If interpretation method 1 is used the accumulation depends on the load characteristic. ζc is nearly constant for all the tests expect test on d = 40 mm. It must therefore be expected to see a relation between ζb and the coefficient to the power fit. A linear relationship is assumed which leads to the following equations.

584

Table 5. Empirical constant for accumulation of the deflection from the cyclic testing. d

u0

α

16 22 28 34 40

0.016 0.015 0.010 0.015 0.012

0.324 0.315 0.339 0.245 0.256

Figure 7. Accumulation of average deflection with interpretation method 2 prediction.

Figure 6. Accumulation of average deflection with interpretation method 1 prediction.

If on the other hand interpretation method 2 is used a linear relationship between the coefficients for the power fit and the scaling factor can be assumed. This leads to following equations.

On figure 6 the accumulation of the deflection for cyclic testing is seen. Here is also shown the prediction as proposed in equation 8 for the interpretation methods 1. The prediction for the interpretation methods 2 can be seen in Figure 7. None of the methods give good predictions, but it seems that interpretation method 2 is the best. It should again be noted that the cyclic loading is performed according to the maximum bearing capacity found from the monotonic tests. This means that the piles are not loaded to the same prototype loads. The maximum prototype load for the small pile with the large scaling factor is therefore larger than the large pile with the small scaling factor. Lin and Liao (1999) proposed a logarithmic fit to the change in secant stiffness as shown in equation

Here k0 is the secant stiffness at the first cycle and κ is an empirical coefficient which control the shape of the curve. n is the number of cycle. A formulation like © 2011 by Taylor & Francis Group, LLC

Figure 8. Change in secant stiffness.

this fits the first 100 cycles for the 5 cyclic tests, but as it can be seen in Figure 8 the secant stiffness starts to decrease or stabilize after 100 cycles. It has not been possible to fit the entire number of cycles cyclic. Looking at Figure 8 it can be seen that the secant stiffness is changing from test to test. The secant stiffness is large for the large piles and smaller for the small piles. The explanation for the difference can again be explained for interpretation method 1 as high ζb values gives small secant stiffness. Using method 2 high scaling factor gives high secant stiffness. No clear dependency is seen for the two interpretation methods.

4

CONCLUSIONS

A test series of modelling of models have been performed for both monotonic and cyclic loading. It has not been possible for the two loading types to reproduce exactly equal prototype response. The results have been analyzed in two ways; one as a normal scatter in the response, and one using a dependency of the scaling factor. It seems like the scaling factor affects the results but it is not clear. Nunez et al. (1988) reports

585

also higher capacity for the small piles tested at high g levels and more tests have to be conducted. The fact that the piles in this test series are acting as stiff piles could be an explanation of the difference from previously modelling of models tests. More tests have to be conducted in order to clarify the scaling laws for these stiff laterally loaded piles. REFERENCES API (1993). Recommended Practice For Planning, Designing and Construction Fixed Offshore Platforms – Load and Resistance Factor Design. American Petroleum Institute. Hansen, J. B. (1961). The ulitimate resistance of rigid piles against transversal forces. Danish Geotechnical Institute, Copenhagen, Denmark Bulletin NO. 12, 5–9. Hoadley, P. J.,Y. O. Barton, and R. H. G. Parry (1981). Cyclic lateral load on model pile i a centrifuge. In Proceedings of the Tenth International Conference on Soil Mechanics ans Foundation Engineering, Volume Vol. 1. Langhaar, H. L. (1951). Dimensional analysis and theory of models. Technical report, John Wiley & Sons. LeBlanc, C. (2009). Design of Offshore Wind turbine Support Structures. Ph. D. thesis, Aalborg University. Leth, C. T., A. Krogsbøll, and O. Hededal (2008). Centrifuge facilities at danish technical university. In 15th Nordic Geotechnical Meeting. Lin, S. S. and J. C. Liao (1999). Permanent strains of piles in sand due to cyclic lateral loads. Journal of Geotech. and Geoenv. Engng. 125 No. 9, 798–802.

© 2011 by Taylor & Francis Group, LLC

Long, J. H. and G. Vanneste (1994). Effects of cyclic lateral loads on piles in sand. Journal of Geotechnical Engineering 120, 225–244. McClelland, B. and J. A. Focht (1958). Soil modulus for laterally loaded piles. Journal of the soil mechanics and foundations division – Proceedings of the American Society of Civil Engineers -, 1–22. Nunez, I. L., P. J. Hoadley, M. F. Randolph, and J. M. Hulett (1988). Driving and tension loading of piles in sand on a centrifuge. In Centrifuge 88. Poulos, H. and T. Hull (1989). The role of analytical geomechanics in foundation engineering. Foundation Engineering: Current principles and Practices, ASCE, Reston 2, 1578–1606. Reese, L. C. and H. Matlock (1956). Non-dimensional solutions for laterally loaded piles with soil modulus assumed proportional to depth. Proceedings of the 8th Conference on Soil Mechanics -, 1–41. Remaud, D., J. Garnier, and R. Frank (1998). Pieux sous charges lat’erales: étude de léffect de groupe. 5. Journées Nationales Génie Civil Génie Côtier,Toulon, pp. 369–376. Stuit, H. G. (1995). Sand In The Geotehcnical Centrifuge. Ph. D. thesis, Technische Universiteit Delft. Taylor, R. N. (1995). Centrifuges in modelling: principles and scale effects. Blackie Academic & Professional. Zhao, Y., K. Gafar, M. Elshafie, A. Deeks, J. Knappett, and S. Madabhushi (2006). Calibration and use of a new automatic sand pourer. In Physical modelling in Geotechnics, 6th ICPMG ’06, pp. p. 265–270.

586

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Gravity based foundations for the Rødsand 2 offshore wind farm, Denmark L. Krogh, J.H. Lyngs & J.S. Steenfelt COWI A/S, Copenhagen, Denmark

ABSTRACT: Foundation of offshore wind turbines is a pre-eminent example of soil-water-structure interaction where loading and response are intrinsically linked to each other. Rødsand 2 is an offshore wind farm just south of Denmark where 90 new 2.3 MW wind turbines will be erected during 2009–2010 to complement the nearby Nysted offshore wind farm. The gravity based foundations are up to 16 m high prefabricated reinforced concrete structures, which are precast on barges in Swinoujscie, Poland. Each barge is towed to the offshore location where the foundations are placed on prepared gravel beds in excavations in glacial deposits. This paper describes the key geotechnical design issues together with the experiences gained from close co-operation between all project parties. Once again designers had to re-evaluate the notion of already knowing all about the strongly overconsolidated clay till, and to face discussions regarding the bearing and sliding capacities of strongly eccentrically loaded foundations.

1

INTRODUCTION

The location of the Rødsand 2 offshore wind farm is in the Baltic Sea just south of Denmark as shown in Figure 1, where 90 new 2.3 MW wind turbines will complement the neighbouring 64 Nysted wind turbines from 2002. The Owner is E.ON Vind Sverige with Grontmij Carl Bro as consultant. COWI is the marine designer for the Contractor Per Aarsleff – Bilfinger Berger Joint Venture. DNV is the certifying body. 1.1 Foundation concept The gravity foundations are designed as reinforced concrete structures, consisting of an octagonal base plate with a cylindrical shaft in the centre as illustrated in Figure 2. Outer walls are established along the edge of the base plate for containing ballast stones of heavy hyperite stone for sufficient dead load. Inner walls connect the outer walls to the shaft. The conical shape of the upper shaft part reduces the impact of ice loads. The shaft is filled with sand to protect the J-tubes inside as well as to provide dead load. This geometry applies to all foundations for construction and economic reasons with the shaft length and amount of ballast stones as the variables in the design process. The mass of the structures is up to 1300 tons. The foundations were precast on barges in Swinoujscie, Poland, and towed to their offshore location, where they were installed in an excavated pit on a prepared gravel bed by a purpose-built crane. A recess in the central part of the gravel bed with a diameter equal to the inner diameter of the shaft was established to © 2011 by Taylor & Francis Group, LLC

Figure 1. Rødsand 2 location in the Baltic Sea just south of Denmark. The circle locates wind turbine position M14.

eliminate the risk of the base plate riding on a hard spot. Ballast stones and scour protection were carefully placed inside and outside the ballast chambers, respectively, making the foundation ready for subsequent installation of the wind turbine.

2

SUBSURFACE CONDITIONS

The site of the wind farm covers an area of approximately 28 km2 and the seabed level on site varies between −5.9 m and −12.5 m. The area is composed mainly of glacial deposits, dominated by different types of clay till with boulders. Sand till occurs in subordinate amounts and meltwater deposits are quite

587

Figure 3. Undrained shear strength measured by UU and CAU triaxial testing. Lower confidence limits are shown as lines.

Figure 2. Foundation geometry. All dimensions are in metres.

rare. In a few parts of the area, the tills are covered by late- and postglacial layers. The till layers are underlain by chalk.At one location a floe of the chalk was met close to the excavation level. 2.1

Site investigations

During 2007 a soil investigation campaign of 80 geotechnical boreholes advanced to a depth of 7 to 10 m, 37 vibrocores and 183 cone penetration tests (CPTs) were conducted within or near foundation footprints to provide the basis for the geotechnical design. Furthermore, 19 CAU triaxial tests were conducted on selected soil samples as well as 200 UU tests on clay till. The latter were carried out as the Site Investigator failed to carry out the planned field vane tests due to the high strength of the clay till. Due to absence of boreholes and/or representative CPTs within the footprint of many of the final turbine locations combined with relatively low penetration depth of the CPTs conducted, supplementary investigations were carried out during winter 2008/2009. 31 boreholes with down-hole standard penetration tests (SPTs) were carried out at 28 locations to a depth between 7 m and 20 m. At one location, four boreholes were carried out to elucidate a highly varying soil profile, where the chalk deposit at places was revealed already at foundation level. Soil samples were generally retrieved for further triaxial and oedometer testing. 2.2

Soil properties

The CPTs revealed high strengths of the strongly overconsolidated clay till, with cone tip resistance up to qc = 30 MPa at shallow depths. However, since the © 2011 by Taylor & Francis Group, LLC

588

penetration depth of the CPTs was rather shallow (average 2.9±2.4 m), the soil strength at depth was derived based on the laboratory strength testing. Figure 3 shows the undrained shear strength as derived from all laboratory strength testing. The tests demonstrate large scatter. The UU tests were expected to underestimate the in-situ strength by a large margin compared to the CAU triaxial tests as reflected in Figure 3, as a result of the generic heterogeneous set-up of the clay till material. The shallow CPTs return in situ cu values between 250–2000 kPa below excavation level. The lower 5% confidence limit for the mean values (as there is no discernible trend with depth) is 535 kPa and 328 kPa for 19 valid original and supplementary CAU triaxial and 200 UU tests, respectively. The mean value and standard deviation are 634 ± 250 kPa and 345 ± 154 kPa, respectively. The ratio of average values is 1.8. The tests support a final value of the undrained shear strength of cu = 250 kPa to be adopted for design at locations with till soils, as opposed to an originally anticipated value of cu = 200 kPa. This allowed a considerable reduction in ballast material. The determination of the drained strength properties turned out to be crucial. The design was optimized to balance the drained bearing capacity with the undrained sliding capacity, with each of these conditions alternately being design governing at different water levels. Figure 4 shows the measured effective strength parameters from the CAU triaxial tests together with the characteristic effective strength parameters as adopted for design. The individual triaxial tests show a relatively large scatter in effective friction angle and effective cohesion, but since the two properties are intrinsically linked to provide the failure condition, it proved efficient to plot the failure points of the total population in order to describe the behaviour of the material. A characteristic effective friction angle of ϕ
= 33◦ and an effective cohesion of c
= 25 kPa was agreed upon to be adopted for design for clay till with cu ≥ 250 kPa. Clay tills with higher plasticity indices (17–31%) than the typical values for Danish clay till (4–8%) were found at some positions. Triaxial and oedometer tests

Figure 4. Effective strength parameters as measured and adopted for design. The MIT stresses s and t are defined as s = (σ1
+ σ3
)/2 and t = (σ1
− σ3
)/2.

Figure 5. Multibeam survey carried out by the Contractor, exemplifying the unevenness of the bottom of the excavation.

demonstrated that reduced values of the strength, stiffness and deformation parameters had to be adopted for design at these locations. 3

GENERAL DESIGN METHODOLOGY

The geotechnical design is conducted in accordance with DNV’s latest code for design of offshore wind turbine structures, DNV (2007), as well as Owner’s specific Design Basis requirements. 3.1 Bearing capacity and sliding

Figure 6. The interface between intact clay till and gravel bed after excavation and placing of gravel.

The bearing capacity was determined in accordance with the general bearing capacity formula for a foundation with horizontal base for the short-term, undrained and the long-term, drained situations, respectively. Additional bearing capacity calculations due to the extreme eccentric loadings, i.e. for e = M /V > 0.3b, were carried out as well. For the 64 neighbouring Nysted turbine foundations the sliding criterion turned out to govern the design due to the possible existence of an upper, weak layer of remoulded soil on top of the intact soil. However, careful planning between Contractor and Designer of pit excavation, cleaning equipment and processes proved it possible to avoid adverse influence from a remoulded layer in the design criteria for the Rødsand 2 foundations. The foundation level with cu ≥ 250 kPa was achieved by excavation using a backhoe with a tolerance of ±0.2 m. After excavation the foundation pit was cleaned for soft material using an air lift mounted on a screeding frame and guided by diving assistance. This method results in an uneven surface with only tens of millimetres of softer material in pockets on top of the strong, intact soil. This condition was verified by multi-beam survey, CPTs as well as diver’s knife tests and video sweeping of the pit immediately before placing the gravel bed material. A survey demonstrating the unevenness of the excavated bottom is shown in Figure 5. Figure 6 illustrates schematically the potential small amounts of settling suspended sediments and local pockets of remoulded/disturbed clay till which will be

penetrated by the gravel layer. The gravel has a mean grain size of about 20 mm and a maximum grain size of 90 mm. In principle, sliding surfaces could pass through the intact clay till, the gravel bed, the interface between concrete and gravel bed or through a combined surface of gravel bed and intact clay till. The characteristic intact clay till strength at excavation level is in excess of cu = 250 kPa, corresponding to a sliding capacity of Hd = cu,d · A
where cu,d is the undrained design shear strength and A
is the effective sliding area. For the gravel and the concrete/gravel interface the sliding capacity is Hd = Vd · tan δ
d , where δ
d is the design interface friction angle, and Vd is the vertical load. All other things considered equal the undrained shear strength increases as a result of the vertical loading by the wind turbine. Due to the initial high value of cu , sliding surfaces passing through clay till or clay till/gravel will have a significantly higher capacity compared with surfaces in the gravel and concrete/gravel interface. The non-existence of a soft soil layer was verified by the execution of five shallow CPTs immediately before placing 0.5 m thick layer of gravel bed material. The interpretation of these required due consideration of the near-surface measurement issues like cracking of the surface soil surrounding the penetrating cone, influence of geometrical conditions of the probe as well as the tip resistance depending on the vicinity to the surface. In this way strict success criteria

© 2011 by Taylor & Francis Group, LLC

589

for the measurements in the upper decimetres were defined to ensure the anticipated soil strength based on a conservative cone factor of 15. This value is higher than the commonly adopted value of 10 for Danish clay till. However, the higher value was chosen in agreement with the Owner and DNV to avoid lengthy testing to accommodate effects of cyclic degradation and because all general experience has demonstrated, that correlation between CPT and triaxial strengths in Danish clay till is difficult. Potential pockets of soft, remoulded material relative to competent clay till were easily identified by a tip resistance < 0.1 MPa, and removed. 3.2

Soil stiffness

Since the turbine loads and the natural frequencies of the turbine and foundation are mutually dependent, the design loads as well as the foundation geometry and properties are determined iteratively between the turbine supplier and foundation designer. Hence, in order to determine the natural frequencies of the turbine and foundation structures, the dynamic foundation stiffness, i.e. the dynamic structural stiffness and the dynamic soil stiffness combined, was determined. The dynamic soil stiffness is directly linked to the dynamic shear modulus of the soil, which in turn is highly dependent on the shear strains induced in the soil. The stiffness at very small strain levels is typically many times higher than the stiffness at large strains. The clay till at the wind farm site is at places so hard that it is not possible to confidently indicate a maximum strength and stiffness of the soil. Hence, only a lower bound value of the soil stiffness was determined using the structural stiffness as the upper limit of the foundation/structure stiffness range. For dynamic shear strains in the range of γ ∼ 0.1%, a lower bound value of the dynamic shear moduγ=0.1% lus was determined as Gdyn = 165cu in accordance with instructions in the Owner’s Design Basis. For the majority of the locations where the soil profile consists of clay till only, the linear-elastic method of George Gazetas, as tabulated in DNV (2007), was used to determine the soil stiffness based on a constant value of the dynamic soil modulus. To demonstrate the magnitude of the induced shear strains, finite element models were employed, i.e. Plaxis 2D v9 with soil model parameters calibrated towards results of the triaxial testing of the soil as well as previous plate loading tests on similar gravel bed material. The stiffness for first time loading as well as for reloading of the gravel bed material were determined from plate loading tests on uncompacted and compacted material, respectively. 3.3 Settlement and tilt The maximum tilt of the wind turbine shall be less than 0.5◦ in accordance with Owner’s Design Basis. This includes an installation tolerance of 0.25◦ . The © 2011 by Taylor & Francis Group, LLC

differential settlement and tilt due to the soil behaviour originate from inevitable variations in the soil below the foundation base plate as well as influence from the prevailing wind direction. The differential settlement originating from variations in the soil was determined dependent on the vertical settlement. It was based on observations of the relationship between the foundation width, vertical settlement and tilt of offshore foundations. Because of the high overconsolidation of the material, the initial and consolidation contributions of the total vertical settlement were determined in accordance with a classic linear elastic calculation based on tangent values of the constrained modulus derived from present oedometer tests. The results of these were found to be in line with general Danish clay till experience. The creep settlements were estimated in accordance with experiences retrieved from extended investigations of Danish clay till conducted by Kristensen et al. (1995), where creep settlements for piers were observed up to a value of 20% of the sum of the initial and consolidation settlements per log cycle of time. For a consolidation time of 1 month, the creep amounted to some 40–50% of the vertical settlement during the lifetime of 25 years. The differential settlement due to a prevailing wind direction was calculated for two cases: A permanent overturning moment corresponding to 0.25Mmax , where Mmax is the maximum unfactored overturning moment in the ultimate limit state, as well as for a case with a temporary, short time, fixed high value of 27 MNm in accordance with Owner’s Design Basis. Settlements and differential settlements proved not to be a design issue.

4

DESIGN AT LOCATIONS WITH LAYERED SOIL

At around 10% of the foundation positions, the soil profile proved to be non-homogeneous, i.e. layered soil consisting of both young and old meltwater deposits as well as Tertiary clay and chalk interbedded in clay till. Furthermore, the soil strength was found to be significantly decreasing with depth at some locations. The common bearing capacity formula equations were not suitable for the layered soil conditions, since introduction of very conservative assumptions would be necessary. Hence, both two- and three-dimensional numerical modelling was applied in the geotechnical design in these situations. These analyses are exemplified in the following by the design process of the foundation at position M14, cf. Figure 1. The soil profile at this location consists of clay till underlain by varying low strength meltwater deposits and low strength clay till, as listed in Table 1. The effective strength parameters turned out to be governing for the design. Where non-horizontal layering was found, the design was carried out assuming worst-case conditions.

590

Table 1.

Design soil profile at foundation location M14. Top level [m] ϕ
[◦ ] c
[kPa] cu [kPa]

Clay till Clay, rather fat Clay till Sand Clay till

−12.5 −15.6 −17.5 −19.0 −22.0

33 17 30 31.5 32

25 4.5 7.5 0 21

250 45 75 – 210

Figure 8. Abaqus calculation model. Left: Entire domain with soil layering. Right: Meshed foundation base plate.

would be strained unacceptably in case of a redesign and construction of a revised foundation geometry, a full three-dimensional modelling was established as non-3D calculation methods were assumed to underestimate the bearing capacity.

4.2 Three-dimensional modelling

Figure 7. Meshed 2D calculation domain showing deformation at failure. The eccentrically loaded foundation is equalised with a centrally loaded effective area.

4.1 Plane strain modelling The finite element software Plaxis 2D v9 was employed based on a plane strain assumption. The intension with the analyses was to reflect an extension of the code-based bearing capacity formula calculations to include soil layering. Thus, the analyses inherently use the same simplifications as the bearing capacity formula equations, e.g.: 1. The moment loading on the foundation is equalized with a centrally loaded effective area of a strip foundation. 2. The torsional moment is applied as an equivalent horizontal force in accordance with DNV (2007). The Mohr-Coulomb material model with characteristic (unfactored) soil parameters was employed for the failure analysis and the safety against failure was calculated by means of a ϕ/c reduction calculation step. Calibration tests were carried out on models with non-layered soil conditions, and an accurate match with results of the bearing capacity formula was found. Thereafter, an extension of the finite element analyses to layered soil conditions was trivial. Figure 7 illustrates the deformed calculation domain for location M14. For nearly all wind turbine locations with layered soil profiles, sufficient safety against failure was verified by the plane strain calculations. Thus, the level of safety was similar to the locations on uniform soil conditions designed in accordance with the bearing capacity formula. However, for the above location M14, the material factor of safety was calculated as Msf = 1.05, i.e. less than the code requirement of γm = 1.15. Since the time schedule and economy © 2011 by Taylor & Francis Group, LLC

A proper modelling of the soil-structure interaction was evaluated to be of highest importance in this analysis. The correct 3D octagonal foundation base plate geometry was modelled with the general-purpose finite element software Abaqus v.6.9-1 as illustrated in Figure 8. The circular hole in the foundation base plate centre models the circular recess in the gravel bed. The foundation was modelled as a rigid body, and the soil was modelled with the Mohr-Coulomb material model and second-order hexahedral solid elements. The torsional moment was applied directly. The analysis was carried out by establishing the initial stress state, applying the foundation loads, and thereafter reducing the soil strength parameters by increasing the material factors until failure was fully developed. The analysis showed an increase in the material factor of safety to Msf = 1.40 compared to Msf = 1.05 as determined by the plane strain analysis, thus verifying sufficient bearing capacity. The increased bearing capacity is attributed to proper modelling of the contact area under the foundation, direct modelling of the torsional loading, and development of a three-dimensional failure figure. In the plane strain analysis, the plastic stress distribution under the effective foundation area is assumed to be uniform, with the failure condition fulfilled everywhere in the area. However, in the 3D total area analysis, it is not necessary to adopt this assumption. Furthermore, the contact area between soil and structure is allowed to be more widespread than in the effective area approach. In the present analysis of location M14, the contact area has been found to be approximately 50% larger than the corresponding effective area. The failure figure and the contact area are depicted in Figure 9, where the failure figure is found to be limited at depths by the interbedded sand layer. A significant amount of the drastic increase in the material factor of safety is attributed to the specific soil conditions at the actual site, with a stronger layer overlying weaker deposits. Thus, the failure will have

591

Figure 9. Left: Magnitude of plastic strain indicated under failure at location M14. Right: Contact area under ULS loading.

Figure 10. A view from the construction site.

a character of punch-through, and the difference from a plane strain analysis with an effective area approach to a three-dimensional model will be increased. In this specific case, the increase in material factor of safety is approximately 35%. Comparative calculations have been carried out with the actual foundation placed on a uniform soil profile. The corresponding increase in material factor of safety for this case was calculated to around 20% depending on failure state (drained/undrained) and the actual material strength properties. 5

of load combinations, and specific advanced numerical modelling. Cast foundations are depicted in Figure 10. The geotechnical design has benefitted fruitfully from the careful planning and knowledge sharing between project parties, resulting in an optimized design already before the early commencement of activities on site. Designer’s incorporation of the contractor’s experience with the offshore equipment and working processes made it possible to close previously unsolved issues with regards to the sliding capacity. Thus, an upper remoulded layer was demonstrated not to exist with the planned excavation and cleaning processes. Furthermore, specific purpose-planned supplementary investigations with triaxial testing made it possible to re-evaluate and increase the original design soil properties with considerable ballast saving as a result. At locations with layered soil conditions, twodimensional plane strain analyses with Plaxis provided a straightforward extension of the normal, well winnowed design scheme, whereas the establishment of an Abaqus model allowed a three-dimensional modelling of the foundation geometry, loading and stress states.The three-dimensional model was found to yield minor to major increases of the bearing capacity compared with the plane strain approach, depending on the actual conditions in question. In the present case with location M14, the establishment of an advanced model allowed the generic geometry of the foundation to remain unchanged, with significant savings both for the design and construction processes as outcome. ACKNOWLEDGEMENT The authors gratefully acknowledge the permission by Owner E.ON Vind Sverige and Contractor Per Aarsleff – Bilfinger Berger Joint Venture to publish the paper.

CONCLUDING REMARKS

The Rødsand 2 project was considered an important show case in connection with the UN COP15 meeting on climate change in Copenhagen, December 2009. Thus, the foundation design process and the foundation production had to meet a very strict deadline. This was achieved with a practical and economical design due to a combination of simple, robust bearing capacity formula design schemes to handle the multitude

© 2011 by Taylor & Francis Group, LLC

592

REFERENCES Kristensen, P.S., Regtop, J. & Balstrup, T. 1995. Predicted and observed settlements and tilts of offshore bridge piers. DGF-Bulletin11. Proceedings XI ECSMFE, Copenhagen. Det Norske Veritas. 2007. Offshore standard DNV-OS-J101. Design of offshore wind turbine structures. DNV.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Geotechnics for developing offshore renewable energy in the US M. Landon Maynard University of Maine, Orono, Maine, USA

J.A. Schneider University of Wisconsin-Madison, Madison, Wisconsin, USA

ABSTRACT: With climate change and energy security issues being of concern globally, interest in renewable energy generation has greatly increased in the US, particularly with the 2005 Energy Policy Act. A major focus has been on development of offshore ocean and wind force energy conversion technologies. However, a disconnect exists between technology designers and developers and the operational environment. Little attention has been given to seabed-foundation interaction, with the exception of offshore wind, and site investigation, soil properties, and foundations are largely ignored until final development stages. Designs that incorporate mooring and foundation system responses with metocean conditions will likely lead to increased efficiency of power production and cost. This paper identifies challenges and recommendations for offshore geotechnics of developing renewable energy facilities, including costs, site investigations and foundations. 1

INTRODUCTION

The development of commercial scale renewable energy sources is growing in interest within the United States (US). Provisions within the 2005 Energy Policy Act increased funding for research and development of clean, renewable technologies such as wind, tidal, and wave energy. Since this time, government agencies such as the National Renewable Energy Laboratory (NREL) and Minerals Management Service (MMS), among others, began publishing wind and ocean energy resource documents, which have served to target technologies and locations for renewable energy development (wind: Musial et al. 2006; current: MMS 2006; wave: Bedard et al. 2005; tidal: Bedard et al. 2006; Musial 2008). Table 1 summarizes the extractable US offshore energy potential for each of the above mentioned energy resources. The estimates for wind include a 60% area of exclusion and 40% efficiency. While much attention has been given to offshore wind in the US in the last decade, particularly to Cape Table 1. Resource comparison of extractable US offshore renewable energy (after Musial 2008).

Energy Source

Extractible Capacity (GW)

Deep water offshore wind (>30 m) Shallow water offshore wind (<30 m) Wave energy Ocean current Tidal energy

375 75 30 6 2

© 2011 by Taylor & Francis Group, LLC

Wind, a Massachusetts coastal offshore farm delayed by environmental permitting and public protest, no developments currently exist. However, several developments are in planning (e.g. Delaware, Rhode Island, Ohio, etc.) for the East Coast and Great Lakes regions (Musial & Ram 2006). Additionally, research and development of tidal, wave, and current ocean energy conversion devices (ECDs) is underway. As experience with offshore wind and ocean ECDs increases, developments will tend to shift from fixed to floating structures to capture greater resource potential in deeper waters. With the exception of offshore wind, the focus of ECD development has almost exclusively been on turbine technology, where simplistic assumptions of seabed fixity are made for initial design and prototyping phases. Seabed-foundation interaction is often overlooked or is one of the last concerns for technology developers. For instance, moorings and seabed attachment ranked 17 of 18 for the top research needs for development of offshore renewable energy resources (Bedard 2008). Yet, neglect of subsurface and foundation issues has been one of the key factors that has lead to delays and cost overruns for offshore wind development in Europe (Gerdes et al. 2006). While turbine technology is paramount to energy extraction, ECD designs that optimize system performance based on seabed-foundation interaction and metocean loading will likely lead to greater power production efficiency and reduced cost. In this paper, we provide an overview of offshore energy conversion devices in use and in development. Issues related to marine geotechnics and a systems approach to efficient and sustainable offshore energy

593

Figure 1. Offshore ECDs fixed to the seabed.

development are emphasized. Lastly, we provide discussion of targeted geotechnical site investigations and laboratory testing, optimized site planning, and foundation options that can aid in efficient and sustainable offshore energy development. 2

Figure 2. Floating offshore ECDs.

METOCEAN ENERGY TECHNOLOGIES

The technologies on the forefront of development are for offshore wind, tidal, wave, current, and ocean thermal energy conversion (OTEC). OTEC is in the formative stage and will not be discussed herein. Fixed and floating technologies discussed in the following sections are illustrated in Figures 1 and 2. 2.1 Wind Wind is currently the most prominent offshore renewable energy source developed worldwide. Europe had © 2011 by Taylor & Francis Group, LLC

594

nearly 2 GW of offshore wind power installed by the end of 2009 and an additional 1 GW of power is expected to be installed by the end of 2010 (EWEA 2009). European offshore wind is installed primarily in shallow waters (<30 m), and the majority are fixed to the seabed using either gravity base or monopile foundations. Costs for shallow water wind structure foundations are approximately 10% of total turbine costs and increase with depth to about 20% of turbine costs at a water depth of 30 m. While no offshore wind is presently installed in the US, ten projects were proposed as of 2007 that totaled

over 1.8 GW of capacity (Musial & Ram 2007). An estimated 80% of US offshore wind energy potential is in water depths greater than 30 m, and over 60% is in water depths greater than 60 m (Robinson & Musial 2006). Key offshore wind development projects include: • Horns Rev & Horns Rev 2 (North Sea, offshore

western Denmark): 369 MW capacity, 171#, 2.0– 2.3 MW turbines, 6–17 m water depth. • Beatrice (North Sea, offshore eastern Scotland): 10 MW capacity, 2# 5 MW turbines supported by fixed jacket, 45 m water depth. • Hywind (North Sea, offshore southern Norway): 2.3 MW, 1# 2.3 MW turbine supported by Spar, 220 m water depth.

2.4

Ocean current energy research was primarily conducted in the late 1980’s and early 1990’s, however it is again being considered as a potential energy source. While the ocean current resource has not been fully analyzed for the US Gulf stream, a potential 6 GW resource is estimated offshore of Florida. Florida Atlantic University’s Center for Ocean Energy Technology is currently developing a deployable ocean current prototype (Driscoll et al. 2009).

2.2 Wave energy Wave energy conversion is a diverse technological industry, which includes oscillating water column (OWC) and overtopping devices (OTD) that are mainly installed in shoreline and nearshore regions which may be fixed or floating structures and wave activated bodies (WAB) that operate in the water column or on the sea surface which are moored and anchored to the seabed (Harris et al. 2004). Minimal large scale development has occurred in the US to date, but major international projects include (Bedard et al. 2005): • 2.25 MW Pelamis wave energy converter (WEC)

currently operating in the Atlantic near Aguçadoura in Northern Portugal. • 0.3 MW oscillating water column device (Oceanlinx) at Port Kembla, Australia. 2.3 Tidal energy Tidal resources in the continental US are relatively minor, but may evolve into a part of the energy balance, particularly within Alaska. Like WECs, tidal energy conversion is in various stages of development, design, prototype testing and full-scale operation. Of particular interest are tidal in-stream energy conversion (TISEC) devices, which are used to capture energy from water as it moves through the device. These devices differ from more traditional technologies that rely on impoundment and release of tidal waters (Bedard et al. 2006). Several operating or prototype TISEC devices exist worldwide: • 1 MW capacity OpenHydro seabed mounted hori-

zontal axis rim drive turbine: Bay of Fundy • 1.2 MW capacity horizontal axis (two rotors on

either side of a monopile) SeaGen: Strangford Lough, offshore Northern Ireland. • Advanced Design Cross Flow (ADCF) prototype floating turbine (Ocean Renewable Power Company) Eastport, Maine. © 2011 by Taylor & Francis Group, LLC

Ocean current energy

595

3

ENVIRONMENTAL LOADING

Substantial experience has been gained over the past 60 years in calculation of wind, wave, current, and ice loads imposed on fixed and floating offshore oil and gas (O&G) structures (API 2000). Ocean ECDs differ from traditional offshore O&G platforms. O&G structures are designed to minimize effects of environmental loading while ocean ECDs are designed to interact with at least one metocean force to generate power. This thereby requires greater engineering of both the structure and foundation to resist the resulting large lateral loads and overturning moments. Shallow water wind turbines rigidly fixed to the seabed via monopoles, GBS, or jacket structures typically experience the maximum horizontal load due to wind forces for water depths less than 10 m, but in deeper waters wind and wave forces are comparable. In the relatively low wave environment of the Great Lakes, high strength freshwater ice may be the critical design loading. Wave, tidal, and current energy devices are designed to operate on the sea surface or in the water column and will not be as affected by wind. Wave forces dominate loads on WEC devices, and tides and currents dominate respective ECD’s. Interaction with other environmental forces must still be considered. For example, wake turbulence forces for submerged ECDs may be a more significant design component due to high seawater density compared to wind and air density. Offshore ECDs are typically relatively light structures that experience large horizontal loads. Ratios of horizontal to vertical load for fixed structures are typically on the order of 20 to more than 50%, and may be significantly larger than expected for offshore O&G structures (e.g., Houlsby et al. 2005, O’Doherty et al. 2009). Floating structures are more prevalent for supporting offshore ECDs, even in relatively near shore environments. Numerical simulation software is publically available for assessing variation in 6-degrees of freedom (pitch, sway, surge, roll, yaw, heave) for floating offshore wind turbines through aeroelastic hydrodynamic modeling (Jonkman & Buhl 2007). Similar studies are not publically available for other floating offshore ECDs, although dynamic response may not be as much of a critical concern for long term performance as it is for offshore wind turbines.

4

GEOTECHNICAL CONSIDERATIONS

Geotechnical considerations for offshore renewable energy facility development will revolve around assessment of soil properties and optimized foundation design. This presents a challenge, considering that project budgets are one to two orders of magnitude lower than those for typical multi-well offshore O&G projects. Below, typical foundations options are presented, followed by a discussion of pertinent soil behavior and optimization of site investigations (SI). Table 2 summarizes foundation options for offshore O&G structures, as well as those used for typical offshore ECDs to date. 4.1

Foundation options

Offshore O&G development over the past 70 years has resulted in use of a wide range of foundation types. Application has not always been successful, although assessment of anticipated response within a rational soil mechanics framework will lead to more reliable design (Randolph et al. 2005). For fixed structures, foundation types may include shallow foundation gravity base systems (GBS), driven piles, bored piles and suction caissons. Anchoring systems for floating structures include driven piles, suction caissons, fixed fluke drag anchors, vertically loaded drag anchors (VLA), suction embedded plate anchors (SEPLA), and dynamically penetrated anchors (DPA) (i.e. rocket or torpedo) anchors for deepwater (> about 30 m water depth). Each of the embedment anchor types may be suitable for fixed or floating ECDs, however primary load direction and combined loading capacity must be checked. The largest range of foundation options to date has been used for fixed and floating offshore wind turbines. Foundation solutions for fixed shallow water wind and ocean ECDs include GBS shallow foundations and a range of driven and drilled single (monopile) and multiple pile systems. Suction caissons have been studied in the last decade as an option for tripod and monopod substructures for offshore wind in shallow and transitional (<60 m) water depths (Houlsby et al. 2005), and may represent viable foundation options for ECDs. WEC device evolution has been from shoreline gravity base fixed structures to floating ECDs operating in higher wave density environments in deeper waters with extreme loading and dynamic conditions. Efficient, reliable anchoring solutions are a priority for new generation WECs, particularly with respect to; i) anchor stability and dynamic response for installations in different seabed soils, ii) anchoring design schemes, and iii) monitoring anchor systems for operating ECDs to better understand operation, maintenance needs, and optimize future designs (Ming & Aggidis 2008). Optimization of the number of anchors required, as well as the site footprint will serve to reduce costs, particularly for site investigations. This has proven © 2011 by Taylor & Francis Group, LLC

effective for shallow water wind. However, with an increase in operational water depth or transition to floating devices, multiple anchors will be necessary. The Pelamis WEC utilizes three drag embedment anchors at the front and one at the rear for yaw control (Bedard et al. 2005) and Hywind, Statoil Hydro’s commercial scale spar floating wind turbine, is anchored using three drag embedment anchors. While these anchors provide economical solutions for small scale or prototype installations, there are some drawbacks to larger scale farms. Current practice for these anchors requires drag installation, keying and proof testing, which requires 2–3 vessels and an ROV, and it can be difficult to assure desired orientation and penetration (Randolph et al. 2005). Drawbacks for drag anchors include: i) their ineffectiveness for vertical loading (unless using VLAs) for taut or tension leg moorings; ii) the required effort to gain capacity; and iii) the uncertainty associated with resultant capacity and anchor final location. The latter may make other fixed location options, such as DPAs or suction caissons, more favorable for large scale floating farms where anchor location is more controllable. Additional investigation into the use of DPAs and suction caissons for ECD anchoring is recommended. Additional efficiency would arise through anchoring multiple floating turbines to a single anchor point (Tong 1988). The complex multidirectional cyclic loading of these foundation elements requires further study. 4.2

Soil behavior

The typical design life of an offshore wind turbine is 20 to 30 years. A similar design life is expected for tidal, wave and current devices. During the design life, structures experience sustained environmental and extreme storm loading that tends to be cyclic in nature. Both static and cyclic soil response must be assessed, noting that analysis of soil behavior for foundation response will differ significantly for fixed and floating structures and for foundation types. The magnitude and anisotropy of cyclic loading will be important for quantifying soil response. For example, tidal turbine seabed mounted arrays will experience two tide cycles per day with alternating current direction, while ocean current seabed mounted arrays will mainly experience one way cyclic loading. Gravity anchors offer a simple solution for floating structures. If motion can be handled by the structure, anchors can be installed with minimal geotechnical investigation. For structures with high lateral loads, more detailed analysis and design are required, which may lead to alternative gravity anchor options such as box or grillage and berm anchors. Design of gravity base shallow foundations systems can require some of the most detailed information on soil response. Cyclic tests of high quality soil samples are required to assess generation of pore pressures, reduction in strength, and accumulation of displacements. As it is essentially impossible to collect a nominally undisturbed sand sample, cyclic tests are often

596

Table 2.

Utilization of various foundation concepts in O&G and offshore renewable industries.

Foundation Option

O&G

Fixed GBS shallow foundations Driven piles Bored piles Suction caissons

Y Y Y Y

Floating Gravity anchors Piles Suction caissons Drag anchors Vertically loaded anchors Dynamic penetration anchors

Y Y Y Y Y Y

Deep water offshore wind (>30 m)

Shallow water offshore wind (<30 m)

Y Y Y Y

Y

Tidal energy

Ocean current

Y Y

Y

Y

Y

Y

Y Y Y

Y

performed on samples reconstituted using relative density estimates from piezocone penetration testing (CPTU), which neglects in situ aging or structure. Knowledge of time dependent changes in total and effective radial stresses is of paramount importance for axial loading of displacement foundations, which include driven piles, suction caissons, and dynamically penetrating anchors. The complex nature of displacement foundation installation typically results in designs based on correlation to CPTU data. Correlations to CPTU data will predominantly be influenced by soil type and foundation geometry, although soil structure and time between installation and loading are also important considerations. Cyclic loading is usually accounted for empirically. Prediction of drag anchor response has a low level of certainty, as evident by keying and proof testing requirements for drag installation that are current practice. The potential of encountering problems with anchoring systems is high when soil-dependent anchor capacity factors are relied upon. A more rational analysis will include analysis of pore pressure and rate effects, such that anchor resistance is based on undrained strength for rapid (undrained) loading, and friction and dilation angles for slow (drained) loading. Undrained strength and friction and dilation angles can be estimated from CPTU penetration resistance, although laboratory testing of good quality samples is required to verify correlations to CPTU data, particularly for critical layers. For fixed offshore ECDs, deformation analysis may be more critical than ultimate capacity. This is particularly true for offshore wind turbines with strict tolerances on dynamic response and foundation rotation. Shear modulus is a difficult design parameter to quantify due to its nonlinear response from small strains. In situ seismic piezocone (SCPTU) measurements of small strain behavior provide the most cost effective assessment of shear modulus, but these values need to be reduced by factors of 2 to 4 to reach © 2011 by Taylor & Francis Group, LLC

Wave energy

597

operational values. Laboratory testing of small strain modulus and modulus reduction with increasing shear strain is recommended for foundation designs where deformations control. 4.3

Site investigation

Foundation design analyses for many offshore ECDs, particularly those of high capacity, are as technically challenging as those for multi-well offshore O&G platforms. Soil samples and in situ test data are typically needed at each foundation location, although this depends on uniformity and familiarity with the local geology. A typical offshore O&G site investigation includes mobilization costs in excess of $500,000 and daily operations costs of $250,000 to $500,000 (Randolph et al. 2005). A 100 m deep boring with continuous downhole CPTU tests may take 2 to 3 days. Borings with continuous sampling may take a similar duration, and should be performed adjacent to at least one CPTU location. For a four legged jacket, investigation costs may be on the order of $5 million, while a floating structure with 8 anchor points may be $10 million. This is a small fraction of a $2 billion O&G platform, but when compared to either a $5 million monopile or $35 million jacket structure supporting an offshore wind turbine, site investigation becomes significant to total project costs. ECD arrays will benefit from cost efficiency of multiple foundation installations once equipment is mobilized, particularly closely spaced arrays (e.g., wave & tidal) in homogeneous sediments with similar anchoring systems. ECD arrays in laterally variable sediments (common in nearshore, previously glaciated regions) or that require large spacing (i.e., wind) will require that more locations are surveyed to determined geotechnical properties. The greater the heterogeneity, the greater numbers of investigations will be required. To overcome these issues, the following site

investigation approach is outlined for use at offshore ECD facilities (e.g. SUT-OSIG 2005): • Develop understanding of local geology through a

detailed desk top study. • Perform a geophysical investigation using mul-

tichannel receivers to assess local variability in geological and geotechnical conditions. • Perform a targeted site investigation consisting of CPTUs and SCPTUs to characterize typical properties and variability within each major soil layer, relying mostly on existing correlations for interpretation. Increased use of cyclic in situ testing should be explored to further streamline the subsequent drilling and sampling program. • Perform drilling and targeted high quality sampling at a select fraction of locations, including typical and atypical layers important for design. Samples should be tested to reduce uncertainty in strength and stiffness design correlations, as well as assess cyclic response and modulus reduction. • Increase use of seabed frames for sampling and in situ testing in applicable soil conditions to eliminate the need for expensive drilling vessels. 5

CONCLUSIONS

A number of different fixed and floating substructure options are in development for offshore energy conversion devices. As developments increase in variety, size and loading complexity, foundation resistance will become a more critical component of design, and alternative foundation solutions will be needed for cost effective development. Future designs will involve complex anisotropic cyclic loading, and require a thorough site investigation (SI). Cost effective SI will likely revolve around detailed near surface geophysics complemented by CPTU and SCPTU. Targeted sampling to address cyclic response and modulus reduction may be replaced partially once cyclic in situ testing is further developed. REFERENCES American Petroleum Institute (API) 2000. Recommended Practice for Planning, Designing, and Constructing Fixed Offshore Platforms – Working Stress Design, API RP2A, 21st Ed., American Petroleum Institute, Washington, D.C. Bedard, R. 2008. Prioritized research, development, deployment, and demonstration (RDD&D) needs: Marine and other hydrokinetic renewable energy, Final Report, EPRI, Dec. Bedard, R. Previsic, M., Polagye, B., Hagerman, G., and Casavant, A.2006. North America tidal in-stream energy conversion technology feasibility study, Report EPRI TP-008-NA.

© 2011 by Taylor & Francis Group, LLC

Bedard, R., Hagerman, G., Previsic, M., Siddiqui, O., Thresher, R., and Ram, B. 2005. Final Summary Report, Project Definition Study, Offshore Wave Power Feasibility Demonstration Project, EPRI Global WP 009 US Rev 2, 22 Sept. Driscoll, F.R. Alsenas, G.M., Beaujean, P.P., Ravenna, S., Raveling, J., Busold, E., and Slezycki, C. 2009. A 20 KW Open Ocean Current Test Turbine, Report of the Center for Ocean Energy Technology, Florida Atlantic University. European Wind Energy Association (EWEA) 2009. Oceans of Opportunity: Harnessing Europe’s largest domestic energy source. Report of EWEA. Harris R.E, Johanning L, and Wolfram J. 2004. Mooring systems for wave energy converters: A review of design issues and choices. In: 3rd Int. Conf. on Marine Renewable Energy, Blyth, UK. Gerdes, G., Tiedemann, A. and Zeelenberg, S. 2006. Case Study: European Offshore Wind Farms: A Survey to analyse experiences and lessons learnt by developers of offshore wind farms, Final Report, Deutsche WindGuard GmbH. Houlsby, G.T., Ibsen, L.B., and Byrne, B.W. 2005. Suction caissons for wind turbines. In: Proc., Int. Symp. Frontiers Offshore Geomech, ISFOG, Perth, 19–21 Sept.: 75–93. Jonkman, J.M., and Buhl, M.L. Jr. 2007. Development and verification of a fully coupled simulator for offshore wind turbines. In: Proc. 45th AIAA Aerospace Sciences Meeting and Exhibit, Reno, NV, 8–11 Jan. 2007. Musial, W. 2008. Status of Wave and Tidal Power Technologies for the United States, NREL/TP-500-43240. Musial, W. and Ram, B. 2007. Status of offshore wind energy projects, policies and programs in the United States. In: 2007 European Offshore Wind Conf., Berlin, Germany, 4–6. Musial, W., Butterfield, S., and Ram, B. 2006. Energy from offshore wind, In: Proc. 2006 Offshore Technology Conf., OTC18355: 11. MMS 2006. Technology White Paper on Ocean Current Energy Potential on the U.S. Outer Continental Shelf, Report of the US Minerals Management Service (MMS) Renewable Energy and Alternate Use Program. Ming, H. and Aggidis, G.A. 2008. Developments, expectations of wave energy converters and mooring anchors in the UK, Journal of Ocean University of China, 7(1): 10–16. O’Doherty, T., Egarr, D.A., Mason-Jones, A., and O’Doherty, D.M. 2009. An assessment of axial loading on a fiveturbine array, Proc. ICE, Energy,162(EN2), 57–65. Randolph, M.F., M. Cassidy, S. Gourvenec and C. Erbrich 2005. Challenges of offshore geotechnical engineering. In: Proc. of the 16th Int. Conf. Soil Mech. and Geotech. Eng. pp. 123–176. Rotterdam: Millpress Robinson, M. & Musial, W. 2006. Offshore Wind Technology Review. [http://www.nrel.gov/docs/gen/fy07/ 40462.pdf] Accessed Feb. 23, 2010. SUT-OSIG (2005). Guidance Notes on Site Investigations for Offshore Renewable Energy Projects. OSIG-Rev. 2, March 2005. Prep. by Offsh. Site Invest. and Geotechnics Group. Tong, K.C. 1998. Technical and economic aspects of a floating offshore wind farm. J. of Wind Engineering& Industrial Aerodynamics, 74–76, 399–410.

598

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Engineering issues for fixed offshore wind turbines on Lake Michigan Mid Lake Plateau, USA P.J. Lang & J.A. Schneider University of Wisconsin-Madison, Madison, WI, USA

K. Smith & T. McNeilan Fugro Atlantic, Norfolk, VA, USA

ABSTRACT: Selection of substructure type and resulting foundation loads for offshore wind turbines are largely dependent on water depth. The water depth of Lake Michigan in the Great Lakes region of the United States (US) rapidly exceeds 50 m, or the maximum depth of previous experience with fixed offshore wind turbines in the North and Baltic Seas. This paper focuses on an area having a water depth less than 60 m which may be suitable for wind turbines supported by fixed jacket structures. Engineering aspects related to offshore wind turbine development at a site on Lake Michigan are presented, specifically loading conditions and subsurface geology, with preliminary calculations performed for foundation sizing. Horizontal loads from freshwater ice was assessed to be the critical loading case, and drilled shaft foundations with a diameter between 0.76 m (30

) and 1.37 m (54

) and lengths of 3 m to 10 m were found to be acceptable for the cases assessed.

1

INTRODUCTION

2

It is a goal of the United States (US) Department of Energy to supply at least 20% of the total US energy use from wind sources by the year 2030.To achieve this goal the potential of offshore wind in the US will need accelerated development. Offshore wind turbines are now being considered at various locations in coastal regions of the US. Initial development will likely begin in the shallow and transitional water depths off of New England and the Mid Atlantic regions of the eastern US, followed by the Great Lakes region in the US Midwest and Canada. Offshore wind in the Great Lakes region has a significant potential to service major metropolitan areas, such as Buffalo, Cleveland, Detroit, Rochester, Toronto, and the area between Milwaukee and Chicago. The additional benefits of excess manufacturing capacity in the area make it a promising region for offshore wind energy development. The deep water of Lake Michigan, as defined for offshore wind turbines, creates challenges for development of offshore wind energy in this area. The shallowest portion of Lake Michigan that is at a distance that eliminates unwanted visual impacts and has optimal wind energy available is located at an area known as the Mid-Lake Plateau. Turbine size, spacing, layout and design loads are discussed for this area, with emphasis on wind turbines supported by quadrapod jacket substructures. Specifically, foundation sizes for 3 MW and 7.5 MW turbines are assessed for the inferred local geologic conditions. © 2011 by Taylor & Francis Group, LLC

SUBSTRUCTURE OPTIONS

There are currently four different wind turbine substructure and foundation options with commercial scale turbines in operation: (1) gravity base systems (GBS); (2) monopiles; (3) jacket quadrapods; and (4) ballast stabilized floating Spars with catenary moorings. In shallow water, depths less than 30 m, gravity base and monopile foundation options are typically used. Approximately 75% of installed capacity to date is supported by monopiles. These substructure options most often support relatively small turbines with rated energy ranging from 0.5MW to 3.6 MW. In transitional water depths, between 30 m and 60 m, practice has tended towards use of the jacket quadrapod (Seidel 2007), although tripod substructures have also been investigated. In water depths greater than about 60 m, floating substructures have been assessed to be the most economically feasible option for offshore wind turbines (e.g., Musial et al. 2006). This paper will focus on water depths on the order of 60 m and a quadrapod jacket substructure option, as illustrated in Figure 1. 3

GREAT LAKES BATHYMETRY

The area of interest for the installation of offshore wind turbines within this study occurs at a minimum distance of 15 km from shore. This distance is selected for two reasons: (i) faster and more uniform wind speeds

599

occur for sites located off western shores (Milwaukee, Chicago, Detroit, Toronto); and (ii) wind turbines can no longer be seen from shore, which may reduce cause for public opposition. At this distance the water depth in all of the Great Lakes, except Erie, is typically greater than 30m depth, with most areas deeper than 60 m. The most cost efficient substructure options for development in these water depths have been assessed to be jacket quadrapods, tripods, and floating options (e.g., Musial et al. 2006). It is important to note that much of southern Lake Michigan (close to high population areas) is considered too shallow for floating

Figure 1. Illustration of turbine with jacket substructure on Lake Michigan Mid Lake Plateau.

Spar substructures yet too deep for fixed substructure options. This leaves development in a large majority of Lake Michigan an extrapolation from experience gained during development of existing offshore wind energy converters. The shallowest portion of Lake Michigan that is greater than 15 km from shore is known as the MidLake Plateau. Three areas of the Mid-Lake Plateau are shallower than 60 m in depth. The first two locations are 15 km and 40 km from shore, but only have sufficient area with water depth less than 60m to develop pilot scale projects of 3 to 10 turbines. This paper focuses on the largest of the three areas located approximately 60km east of the Wisconsin coast and 70 km ENE from Milwaukee, as shown in Figures 2 through 4.

Figure 3. Subsurface area than can be utilized that is in less than 60 meters of water for Lake Michigan Mid-Lake Plateau site.

Figure 2. Water depths and major cities in the US and Canadian Great Lakes region.

© 2011 by Taylor & Francis Group, LLC

600

A cross section running eastward from Milwaukee, as well as a second profile of water depth running east from a point 30 km north of Milwaukee (through the Mid Lake Plateau), are shown in Figure 4. Near Milwaukee the water depth increases at a rate of about 3.5 m/km, but further north, where wind conditions are improved, the rate is twice that value at about 7 m/km. Transitional or deep water substructures would be required relatively close to shore.

4

LAKE MICHIGAN GEOLOGY

Figure 4 illustrates that the subsurface conditions at the Lake Michigan’s Mid-Lake Plateau consists primarily of exposed dolomite bedrock, with up to 10m of till in some locations. The dolomite of the mid Lake Plateau has been inferred to be the more resistant dolomite of the Middle Devonian Traverse formation (Wold et al. 1981). No rock cores were available at the actual site location, and rock strength for preliminary analyses is based on samples tested for the Marquette Interchange transportation project in Milwaukee. Bedrock below this area of Milwaukee is inferred to be dolomite of the Middle Devonian Detroit River formation or Middle Silurian dolomite, both expected to be somewhat softer than the Traverse formation of the Mid Lake Plateau (Wold et al. 1981). A boring where rock properties were obtained is included at the west end of the cross section in Figure 4. A specimen was collected from a depth of 90 m, located within the upper 5 m below the rock-soil interface. The dolomite bedrock from the onshore boring had a wet density of 26.5 kN/m3 , an unconfined compression strength of 115 MPa, and a rock quality designation (RQD) of 90%. Site specific geophysical and geotechnical investigations are necessary to update this preliminary assessment of subsurface conditions.

5 TURBINE LAYOUT The portion of the Lake Michigan Mid Lake Plateau of interest is the shaded area in Figure 3 and includes areas on both sides of the Wisconsin-Michigan state boundary. For a water depth of less than 60m, an area of approximately 75 km2 is delineated, 57 km2 of that in Wisconsin waters and 18 km2 in Michigan waters. A turbine spacing of 3 rotor diameters perpendicular to the prevailing wind and 10 rotor diameters parallel to the prevailing wind has been selected to minimize wind turbulence and maximize energy output. The layout is organized to optimize production from a west-northwest prevailing wind, although further optimization would likely result from a site specific investigation into wind and subsurface conditions. Table 1 summarizes the number of turbines that each area will accommodate based on the relationship between rated output and blade diameter for offshore turbines. For the selected layout, use of 3 MW turbines results in the maximum installed capacity. Larger turbines significantly reduce the number of installations that are required, and a hypothetical 7.5 MW turbine would result in 10% less capacity for approximately one third the number of installations. Foundation sizes for 3 MW and 7.5 MW turbines supported by jacket substructures are analyzed as example cases in this paper.

6

LOADING

6.1 General conditions Loading cases for a 3 MW and 7.5 MW wind turbine have been analyzed for a quadrapod steel jacket substructure and foundation. Wind (W), hydrodynamic (H, wave and current), ice (I), and dead loads (D) have been taken into consideration. A water depth of 60 m and a turbine hub height of 100m above the water are assumed for these preliminary analyses. Using load and resistance factor based design (LRFD), the factored resistance needs to be greater than the factored load:

Table 1. Number of turbines that fit on Mid Lake Plateau as a function of blade diameter (D).

Figure 4. Cross section of Lake Michigan geology from Milwaukee east with overlay of Mid Lake Plateau bathymetry (geologic cross section after Wold et al., 1981).

© 2011 by Taylor & Francis Group, LLC

601

Wisconsin (57 km2 )

Michigan (18 km2 )

Total (75 km2 )

Rated Output (MW)

Blade D m

#

Cap (MW)

#

Cap (MW)

#

Cap (MW)

2 3 3.6 5 7.5

80 90 107 129 150

297 235 166 114 84

594 705 598 570 630

94 74 52 36 27

188 222 187 180 203

391 309 218 150 111

782 927 785 750 833

Where φ is resistance modification factor, Rn is the calculated (nominal) resistance. A resistance modification φ factor of 0.7 has been used for these studies. With respect to loading, γi is the load factor and Qi is the calculated load, where the subscript i indicates load type, such as wind, wave, current, or ice. DNV (2007, Table F1) recommends using a load factor of unity for the following probabilistically based environmental ULS load combinations: 1. 2. 3. 4/5.

50-yr Wind / 5-yr Wave / 5-yr Current 5-yr Wind / 50-yr Wave / 5-yr Current 5-yr Wind / 5-yr Wave / 50-yr Current 5/50-yr Wind / 5-yr Current / 50-yr Ice

The 50-yr event corresponds to 98% nonexceedence, while a 5-yr event corresponds to 80% nonexceedence. For cases 4/5, a 50-year wind is used only if ice occurs at the location every year.

Figure 5. Distribution of wave height on southern Lake Michigan (modified after Liu 1986).

6.2 Dead loads Dead loads result from the weight of the turbine and substructure. As water depth and turbine size increase, jacket size and dead load increase. Dead loads for the quadrapod jacket structure are estimated from a summary presented by Seidel (2007), and result in a 10 MN structure for a 3MW turbine and a 20 MN structure for a 7.5 MW turbine. Figure 6. Distribution of wind speeds over southern Lake Michigan (modified after Liu 1986).

6.3 Wind & wave loading Both wind and wave loading have been analyzed based predominantly from a buoy in Southern Lake Michigan (National Buoy Data Center Station 45007). Analysis is based on discussion by Liu (1986), and representative gamma (probabilistic) distributions for waves and wind are shown in Figures 5 and 6. Wind data were collected at a height of 5 m, and were scaled to values at 60 m, 80 m, and 100 m based on information provided by PSCW (2008). There is a relatively small wave environment in Lake Michigan, with a 50-year significant wave height of approximately 3 m having a period of 8 s, and a 5-year current of approximately 0.25 m/s. Hydrodynamic loads with a 50-yr wave and 5-yr current on the order of 0.2 MN to 0.3 MN are calculated to act on substructures supporting the 3 MW and 7.5 MW turbines, respectively. Wind loading cases are somewhat less clear. Figure 6 indicates a 5-year wind speed at a hub height of 100 m to be 14 m/s, increasing to 24 m/s for a 50year event. Since wind turbines adjust the pitch of the blades to keep a constant power output when the rated energy is reached, Figure 7 shows that horizontal force decreases with increasing wind speed after approximately 12 m/s. For a hypothetical 7.5 MW turbine, the maximum calculated force is 1050 kN, but the force corresponding to a 5-year event is 865 kN reducing to 335 kN for a 50-year event. The maximum force (1050 kN for the 7.5 MW turbine) will occur during the life of the turbine and may occur simultaneously with © 2011 by Taylor & Francis Group, LLC

602

Figure 7. Wind forces as a function of wind speed for hypothetical turbines of various rated outputs.

extreme hydrodynamic and ice loading conditions.The maximum wind force from Figure 7 is used for both 5-yr and 50-yr wind loads in these calculations. 6.4

Ice loading

Ice loading will increases with contact area (element diameter multiplied by ice thickness), ice compressive or flexural strength, and is modified by a shape factor (α). Flexural strength is often 1/4 to 1/2 that of compressive strength, and therefore a structures with ice cones at the water line are analysed to induce a failure controlled by flexural strength. Figure 8 illustrates

7 7.1

Figure 8. Distribution of ice thickness for two Lake Michigan locations (data from NOAA 2005). Table 2.

Summary of foundation loads for cases considered.

Case W1 H1 I1 D1 L2 C2 T2 (MW) (MN) (MN) (MN) (MN) (MN) (MN) (MN) 3 7.5

0.4 1.0

0.2 0.3

1.0 1.5

10 20

2.0 3.6

6 12

0.5 1.0

1 W = max Wind Load; 50-yr H = Hydrodynamics Load; 50-yr I = Ice Load; D = Dead Load (self weight of structure) 2 L = Critical lateral load; C = Compression load on leeward legs of structure; T = tension load on windward legs of structure.

gamma distributions for ice thickness on Lake Michigan for two locations, the Mid Lake Plateau assessed in this paper, and a shallower water coastal area near Muskegon, Michigan. Ice thickness tends to decrease as water depth increases, and design conditions at the Mid Lake Plateau are 0.3 m for the 50-year case. Shallow waters areas (i.e., Muskegon) may have design ice thickness exceeding 0.8m. For simultaneous impact on two legs, a 3 MW turbine supported by a jacket with 0.85 m diameter legs will result in a horizontal load of 1 MN. The load would increase to 1.5 MN for a 7.5 MW turbine supported by a jacket with 1.3 m diameter legs. These horizontal loads are significantly higher than hydrodynamic loads mentioned in the previous section, and will control design. It is important to note that uplift loading from seiches is possible (Wortley 1984), but analysis procedures require further study. 6.5 Load summary Two turbines sizes have been analyzed in this paper to determine the loading conditions for foundations supporting a quadrapod jacket offshore wind turbine substructure on the Lake Michigan Mid-Lake Plateau. The critical design case consists of a maximum wind load, 5-year current, and 50-year ice loading scenario. Table 2 summarizes the critical loading conditions. At the foundation level, a lateral load (L) will result, and the foundation elements will resist moment from that lateral load through compression (C) in the leeward leg and tension (T) in windward leg of the structure. © 2011 by Taylor & Francis Group, LLC

FOUNDATION DESIGN General conditions

Foundations for the quadrapod jacket substructure analyzed in this paper are designed to have four concrete drilled shafts fixed in the bedrock of the Mid-Lake Plateau. In areas where overlying till exists, the till is assumed to offer negligible resistance as compared to the bedrock. Three diameters for drilled shafts are compared for this assessment, 0.76 m (30

), 1.07 m (42

), and 1.37 m (54

). A concrete annulus of 100 mm is assumed to surround a steel pile insert. The lateral shear loads (L), summarized in Table 2, will be resisted using conventional pipe pile inserts with a ratio of pile diameter to wall thickness (D/t) of 20 to 40. The following discussion focuses on axial resistance of the foundation due to the moment induced by horizontal loading. The piles are assumed to generate all of their resistance through shaft friction. End bearing is neglected due to the potential for a soft base response that may occur from drilling debris remaining beneath each pile. Three failure mechanisms are discussed in the following (i) axial resistance in compression; (ii) axial resistance in tension, and (iii) an uplift tension cone failure. 7.2 Axial resistance Axial compressive loading is assumed to be resisted solely by friction (τf ) on the side of the drilled shafts as end bearing of the drilled shaft is neglected. Shaft friction on drilled shafts in rock tends to increases with the square root of the unconfined compressive strength (UCS). For rock with UCS values greater than 4 MPa, such as fresh Lake Michigan dolomite, the shaft friction will be limited by failure at the rock-concrete interface.This limiting strength is approximately equal to 750 kPa, which was used for shaft friction values when assessing axial resistance. Since end bearing is neglected, tension resistance is calculated to be the same as compressive resistance. For the 3 MW turbine, the critical foundation load is 6 MN occurring for the compression case. For the 7.5 MW turbine foundation the critical load is twice as high at 12 MN. Friction on the sides of the drilled shafts increases linearly with pile length and diameter due to the constant limiting shaft friction (τf ) of 750 kPa used for analysis (Rn /L = π·D·τf ). Required pile lengths (using a strength reduction φ factor of 0.7) ranged from 3 m to 10 m, as summarized in Table 3. The minimum slenderness ratio (L/D) is calculated to be nearly 2 for the 1.37 m diameter shaft and 3 MW turbine, and was calculated to be as high as 13 for the 0.76 m diameter shaft and 7.5 MW turbine. 7.3 Tension cone failure An alternative failure mechanism may occur for near surface rock socketed drilled shafts if the rock itself

603

Table 3.

Summary of foundation pile design lengths.

3 MW Turbine

installation of drilled shafts. A drive-drill option is often used in this case where a steel pile is driven to the top of the bedrock followed by construction of the drilled shaft into rock. The same failure mechanisms analyzed for the rock socketed drilled shafts discussed above need to be considered, as well as a more detailed analysis of lateral loading of the driven pile.

7.5 MW Turbine

Comp. Tension Comp. Tension γ Qi = 6 MN γ Qi = 1.8 MN γ Qi = 12 MN γ Qi = 3.6 MN D (m) Pile length (m) 0.76 5.0 1.07 3.5 1.37 3.0

<2 <2 <2

9.5 7.0 5.5

3.0 2.0 <2

8

For offshore wind turbines, water depths of Lake Michigan are considered deep, creating challenges for development. The relatively small wave environment and high strength of freshwater ice indicate that horizontal ice loading will be the most critical case for design. Foundations for jacket structures assessed resist moment induced by horizontal loading through compression in the leeward leg and tension in the windward leg. Compression is assessed to be the critical design case for foundation sizing. Geophysical and geological maps indicate that the Mid-Lake Plateau consists of shallow bedrock, which makes drilled shafts (drilled and grouted piles) a likely choice for foundation construction. Foundation resistance results solely from side friction on the drilled shafts, and is limited by the bond strength of the concrete and rock interface in the high strength rock that is expected. Depending on subsurface conditions, required foundation lengths may be an order of magnitude greater than those calculated for strong rock, highlighting the importance of high quality site investigation early in a project.

yields prior to failure at the rock-concrete interface. A cone shaped failure surface will form and include the drilled shaft and a surrounding volume of rock. Axial resistance to a cone failure is controlled by friction on the side of the drilled shaft below the cone and the uplift resistance of the rock cone itself. The shaft friction is calculated in the same manner as described for axial loading while the cone failure uplift capacity is calculate as a function of the tensile strength of the rock, the apex angle of the cone shaped failure mass, and the depth of the apex below the ground surface (e.g., Wyllie 1992). Tensile strength of the rock is estimated from UCS. The foundation for the 3 MW turbine has a maximum uplift load of 1.8 MN and the foundation for the 7.5 MW turbine foundation has a maximum uplift load of 3.6 MN. The nature of the strong unfractured rock expected near the seabed at this site resulted in a minor effect of drilled shaft diameter on the embedment length required to resist failure by a tension cone mechanisms. In all analyzed cases the required drilled shaft length is calculated to be less than 2 m. 7.4

Foundation pile design implications

Table 3 compares minimum pile lengths for compressive and tension loading cases as a function of pile diameter and turbine size. Pile lengths are controlled by the compression case. It is important to note that a larger horizontal load, a lighter structure, a larger diameter pile, weaker or fractured rock could shift the control of the design to the case of an uplift failure. Alternatively, much longer driven piles would be required to resist these design loads in different geological conditions.A driven pile with a 1.22 m diameter in continuous stiff clay or dense sand would need to be approximately 23 to 45 m long. A driven pile of the same diameter in continuous soft clay or loose sand would need to be on the order of 110 to 230 m in length. These vast differences in pile length highlight the importance of geologic conditions on foundation design and construction. Areas of Lake Michigan’s Mid-Lake Plateau have till overlying the dolomite bedrock. This till is assumed to be too thin to add significant levels of support for the foundation, but may cause difficulties during

© 2011 by Taylor & Francis Group, LLC

SUMMARY & CONCLUSIONS

REFERENCES DNV. 2007. Design of Offshore Wind Turbines, Offshore Standard DNV-OS-J101: 140. Liu, C.P. 1986. Estimating long term wave statistics from long term wind statistics, Proc. 20th Coastal Eng. Conf, Taiwan, 512–521. Musial, W., Butterfield, S. & Ram, B. 2006. Energy from offshore wind, Proc. 2006 Offshore Technology Conference, OTC18355. Houston OTC: 11. NOAA (2005) An Electronic Atlas of Great Lakes Ice Cover: 1973–2002. http://www.glerl.noaa.gov/data/ice/atlas/. PSCW 2008. Harnessing Wisconsin’s Energy Resources: An initial investigation into Great Lakes Wind Development, Docket 5-EI-144, Wisconsin Public Service Commission. Seidel, M. 2007. Jacket substructures for the REPower 5M wind turbine, Proc. Eur. Offshore Wind 2007, Berlin. Timco, G.W., and O’Brien, S. 1994. Flexural strength equation for sea ice, Cold Region Sci. and Tech., 22: 258–298. Wold, R.J., Paull, R.A., Wolosin, C.A., and Friedel, R.J. 1981. Geology of Central Lake Michigan, AAPG Bulletin, 65(9): 1621–1632. Wortley, C.A. 1984. Great Lakes Small-Craft Harbor and Structure Design for Ice Conditions: An Engineering Manual. WIS-SG-84-426, University of Wisconsin Sea Grant. Wyllie C. D. 1992. Foundations On Rock. E&F Spon, London, UK.

604

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Centrifuge model tests on piled footings in clay for offshore wind turbines B.M. Lehane University of Western Australia

W. Powrie University of Southampton

J.P. Doherty University of Western Australia

ABSTRACT: The paper presents the results from a series of centrifuge tests which examine the benefits of employing a footing together with a ‘standard’ monopile as a foundation solution for an offshore wind turbine. The experiments were carried out in firm to stiff kaolin clay and involved monotonic application of lateral loads at an equivalent prototype height of 30m above the foundations. Tests were conducted on piled footings, monopiles and an un-piled footing. The experimental results and observations are compared with those obtained from a parallel series of 3D Finite Element analyses.

1

INTRODUCTION

It is widely acknowledged that the world needs to move rapidly away from a dependence on fossil carbon fuels. Wind power represents a major potential source of renewable energy, but obstacles to exploiting its full potential remain. For example, as the preferred turbine size increases (noting that 7 MW turbines with a rotor diameter of 126 m are already in operation), there are ever increasing demands on foundation stiffness and capacity, particularly offshore where wave loading acts in addition to wind loading. The most popular offshore wind turbine foundation solution is a single pile (or monopile) with a maximum (currently achieved) diameter of about 5 m. The lateral and rotational stiffness provided by such a pile is not likely to be sufficient to support the new generation of wind turbines and therefore a range of different foundation solutions are being examined. One type of offshore wind turbine that has been proposed but not extensively investigated comprises a monopile combined with a footing base (Dixon (2005); Stone et al., (2007)), illustrated schematically in Figure 1. The intended principle of operation is similar to that of an embedded retaining wall with a stabilising platform (Powrie & Chandler 1998; Powrie & Daly 2007). The platform/footing potentially contributes to the performance of the foundation system in four ways: (i) As the pile tends to rotate under the action of the horizontal and moment loading, the development of bearing pressure on the underside of the footing imparts an additional restoring moment to the foundation system. © 2011 by Taylor & Francis Group, LLC

605

Figure 1. Hybrid monopile/footing foundation (Dixon 2005).

(ii) The bearing pressure on the underside of the footing also increases the vertical stresses in the soil in front of the pile, potentially increasing the lateral stresses in front of the pile and hence further resisting rotation.

Figure 2. Profile of T-bar undrained shear strength measured in centrifuge sample.

(iii) Shear stresses on the underside of the footing can provide resistance to the horizontal loads. (iv) If the footing is embedded to some extent, horizontal loads may also be resisted by the soil in front of the platform. Initial tests on piled footing models, carried out in sand at 1 g, were reported by Stone et al. (2007), with apparently promising results. This paper reports, discusses and analyses the results of a series of tests on piled footing models carried out in a clay soil in a geotechnical centrifuge. The effects of the footing on foundation system stiffness, pile bending moments and ultimate system capacity are investigated. 2

MODEL PREPARATION AND TESTING PROCEDURE

The model foundations were tested in a speswhite kaolin clay specimen, which was prepared from a slurry at a water content of 120% (about twice the liquid limit) by one-dimensional consolidation in a laboratory press to a vertical effective stress of about 225 kPa. The clay block was then placed in the University of Western Australia (UWA) beam centrifuge and allowed to come into equilibrium (over a period of 3 days) at a centrifugal acceleration of 200 g. The T-bar shear strengths, assessed using a T-bar factor of 10.5 (Randolph 2004), recorded at various stages of the test programme are plotted on Figure 2 and indicate good repeatability, particularly over the uppermost 100 mm (=20 m prototype), where clay strength has a controlling influence on foundation performance. © 2011 by Taylor & Francis Group, LLC

Figure 3a. Pile with solid footing attached.

These su-Tbar strengths correspond to a normally
consolidated undrained strength ratio (suT-bar /σv0 )nc of 0.16 assuming a variation in strength with overconsolidation ratio raised to the power of 0.8, i.e.

) = 0.16.OCR0.8 . (suT-bar /σv0 )nc ratios of (suT-bar /σv0 0.17 ± 0.02 are typical of those indicated in centrifuge tests on kaolin at UWA; see Lehane et al. (2009). All foundation tests were conducted at a centrifugal acceleration of 200 g, on one of two 1:200 scale model piles. The first model pile was a 1 mm thick, 19 mm diameter (Dp ), closed-ended aluminium pipe with strain gauge bridges wired to respond to bending installed at depths below the underside of the footing of 0.5, 1.5 and 3.25 times the pile diameter, Dp . The second pile was made from a solid aluminium rod, also of diameter Dp = 19 mm, and was used in tests carried out to investigate the ultimate capacity of the foundation at failure in the ground. Footings were 15 mm high, circular aluminium discs attached rigidly to the pile at the mudline (Figure 3a). Two of the footings were solid with diameters (Dfooting ) of 60 mm and 75 mm, while a third footing (with Dfooting = 75 mm) incorporated a 12 mm skirt around its outer periphery (Figure 3b); the latter was used to investigate the potential influence of suctions underneath the platform on the tension side of the pile. In all, six different foundation geometries were tested, as summarised in Table 1. The centrifuge was halted prior to the installation of each foundation. Foundations were pushed using an actuator into the clay sample at a rate of 2 mm/s to a

606

Figure 3b. Three footing employed: from left: skirted (Dfooting = 75 mm), solid (Dfooting = 60 mm) and solid (Dfooting = 75 mm).

Table 1.

Test 1 2 3 4 5 6

Details of centrifuge tests performed.

Foundation geometry Pipe pile with 60 mm diameter footing Pipe pile only Pipe pile with 75 mm diameter footing Solid pile with 75 mm diameter footing Pipe pile with skirted 75 mm diameter footing 75 mm diameter skirted footing only

Prototype dimensions (m)

Max H applied (N, model ≡ MN, prototype)

Dpile = 3.8 m Dfooting = 12 m

220N ≡ 8.8MN

Dpile = 3.8 m Dpile = 3.8 m Dfooting = 15 m

230N ≡ 9.2MN 220N ≡ 8.8MN

Dpile = 3.8 m Dfooting = 15 m

295N ≡ 1.8MN

Dpile = 3.8 m Dfooting = 15 m

220N ≡ 8.8MN

Dpile = 3.8 m Dfooting = 15 m

21N ≡ 0.83MN

Figure 4. Lateral load – displacement response observed in centrifuge (prototype units).

3 3.1

point at which the underside of the footing was embedded by a distance of 4 ± 1 mm below the clay surface with the pile tip at an embedment of 211 mm (=11Dp ). The centrifuge acceleration was then increased back to 200g and the model left for a period of at least 30 minutes to allow the clay to reconsolidate. In all tests, lateral loads were applied at a distance of 150 mm (7.9Dp ) above the underside of the base of the footing (when present) using the arrangement shown in Figure 3a incorporating an above-ground tower section with the same properties as the pile under test. The height at which the lateral load was applied (equivalent to 30 m at prototype scale) gives a moment (M) to lateral load (H) ratio within the range experienced by offshore wind turbines (Byrne & Houlsby 2003). Lateral loads were applied in a series of load (or reload)/unload cycles of increasing magnitude at a deflection rate of 6 mm/min at the actuator location. Structural analyses show that this rate corresponds to a rate at the mudline of at least 1.5 mm/min (equating to 8% of the pile diameter per minute), which experience with overconsolidated kaolin at UWA indicates is fast enough to ensure essentially undrained conditions. © 2011 by Taylor & Francis Group, LLC

RESULTS Load-displacement response

The lateral load (H) – lateral displacement (y) relationships (with the unload-reload loops omitted) observed in all six centrifuge tests are plotted using prototype units in Figure 4. The lateral displacement is measured at the actuator location and is normalised by the pile diameter (Dp ). It is apparent that:

607

(i) The ultimate capacities of the solid pile and unpiled footing are attained at lateral displacements (y) in excess of 1Dp and at 0.1Dp , respectively. Based on the flexural rigidity of the tower, the movement at the soil surface for the solid pile when y = 0.1Dp is estimated to be about 0.03 Dp . (ii) The capacity of a pile foundation (with or without a footing) is at least 10 times greater than that of a footing alone. (iii) Although the initial stiffness of the un-piled footing is comparable to that of a pile, the response of the pipe pile with no footing is virtually indistinguishable from that of pipe piles with footings. Analyses indicate that the higher stiffness of the footing-solid pile combination can be attributed to the higher flexural rigidity of the solid pile compared with that of the flexible pile.

(iv) The ultimate lateral capacity (Hult ) of the solid pile (i.e. the lateral geotechnical capacity) is approximately 12MN. The experiments also showed little difference between the skirted and solid footings. This is likely to be because the vertical stress on both footings (up to 80 kPa at footing/skirt tip level) was such that the both footings remained in full contact with the soil under the maximum applied lateral loads. 3.2 Operational soil strength The response of the footing with no pile (Test 6) was dictated by the relatively high moment (M) and vertical load (V). The buoyant weight of the footing and tower elements above the footing level (noting that the water level was maintained at the level of the top of the footing during testing) would, if taken entirely by the footing, apply a net vertical stress of approximately 80 kPa. The ratio of the moment at failure to the product of the footing area and diameter (M/ADfooting ) was 9.4 kPa; and the applied shear stress on the footing base (H/A) at failure was 4.7 kPa. The capacities of circular footings on a uniform clay subjected to undrained moment and vertical loading are presented by Taiebat & Carter (2002) as an envelope of combinations of M/ADsu and V/Asu . Their chart indicates that, for this geometry and vertical load, the average operational shear strength controlling failure in Test 6 was 19 kPa. Figure 2 shows that this strength is broadly consistent with suT-bar at a depth of one quarter the footing diameter. In reality, much of the vertical load is likely to have been carried by pile skin friction, reducing the net stress beneath the footing perhaps to ∼40 kPa. The chart given by Taiebat & Carter (2002) indicates that the footing moment capacity is not much affected by a reduction in the vertical stresses from 80 kPa to 40 kPa for the operational su value of 19 kPa. An ultimate lateral geotechnical capacity of the pile may be estimated from the difference between the ∼12MN ultimate lateral load of the solid pile and footing in Test 4 and the ultimate lateral load of 0.83MN resisted by the un-piled footing in Test 6 as about 11 MN. The Oasys ALP program (Oasys 2006) was used to backfigure operational undrained soil shear strengths for this ultimate lateral load of 11MN.Adopting the API recommendations for ultimate lateral soil resistance (API 2000), it was found that, in contrast to the trend shown by the un-piled footing (Test 6), backcalculated operational strengths were 60% higher than the T-bar strengths shown on Figure 2. These higher strengths are consistent with a normally consolidated
)nc of 0.25, which is undrained strength ratio (su /σv0 typical of that shown by kaolin samples tested under triaxial compression conditions (see Lehane et al. 2009). 3.3 Pile bending moments The (prototype) bending moments in the pipe pile, which were inferred from the pre-calibrated strain gauge outputs, are plotted in Figure 5 for Test 2 (pile © 2011 by Taylor & Francis Group, LLC

Figure 5. Measured (prototype) bending moments in centrifuge Tests 2 and 3.

only) and Test 3 (pile with 15m diameter footing). The moments derived for the pile only case are in close agreement with those determined using the OasysALP laterally loaded pile analyses, with the clay undrained shear strength set to 1.6suT-bar (for the reason mentioned above). The moments measured in the piled footing were virtually identical to those measured in the pile only case. Maximum moments were generally at z/Dp = 3.5, although these were only marginally (∼10%) higher than moments recorded at z/Dp = 0.75. This indicates that the increase in moment with depth due to the higher lever arm of the applied lateral force is approximately balanced by the soil passive resistance over the upper 3.5 diameters of the pile depth.

4

FINITE ELEMENT RESULTS

Three-dimensional finite element analyses were carried out using Plaxis 3D Foundations (V2.2, Brinkgreve & Swolifs 2009) to calculate the response in Tests 2, 3 and 6 (see Table 1) and to assess the relative stiffness and capacity of the piled footing, the un-piled footing and the pile only. The mesh used for the piled footing analysis is shown in Figure 6. The prototype pile diameter of 3.8 m was modelled using elastic continuum elements. The Young’s modulus was taken to give the pile a flexural rigidity equivalent to a 3.8 m diameter aluminium circular hollow section with

608

Figure 6. Plaxis 3D mesh used for the FE analyses.

Figure 7. Plaxis 3D predictions for the response of three foundation types (Dp = 3.8 m and Dfooting = 15 m).

a wall thickness of 200 mm. The soil was modelled as an elastic perfectly plastic Tresca material with a tension cut-off of zero. The soil undrained shear strength (su ) was taken as the best linear fit to the T-bar data shown in Figure 2, su (kPa) = 7.5 + 1.75z, where z is the depth in metres. For the purposes of this study, a constant Young’s modulus of 8 MPa was specified for all the clay and Poisson’s ratio was taken as 0.49 to represent undrained conditions. As in the centrifuge tests, the pile extended 40 m below the base of the footing and the horizontal load was applied 30 m above the top of the footing. The case involving the pile only was modelled by removing the footing from the model and the footing only case was modelled by removing the embedded portion of the pile (and retaining the footing). The footing, which had the equivalent prototype diameter and thickness of 15 m and 3 m respectively, was modelled as a linear elastic material with the properties of aluminium. The footing was embedded 0.6 m below ground surface (i.e. its upper surface was 2.4 m above the ground surface). The lateral load-displacement responses calculated using Plaxis 3D for the three cases examined are shown in Figure 7. A comparison of Figure 4 with Figure 7 indicates close agreement between the centrifuge tests and the FE results. The FE-calculated lateral stiffnesses (load:deflection ratios) at the point of load application (i.e. 30 m above the ground surface) are summarised in Figure 8. This indicates that the combined stiffness

Figure 8. Lateral stiffness at point of load application for the three foundation types (Dp = 3.8 m and Dfooting = 15 m) calculated using Plaxis 3D.

© 2011 by Taylor & Francis Group, LLC

609

of the footing and pile (kpf ) is about 25% greater than that of the pile alone (kp ) at small displacements. However, the benefits of the footing reduce with continued displacement with the kpf /kp ratio approaching unity at a displacement of 20% of the pile diameter (which equates to 0.8 m at the point of load application or ∼0.25 m at the ground surface). Figure 8 also shows that the combined stiffness of the pile and footing may be taken approximately as the sum of the individual stiffness values of the pile and footing. It is apparent that the initial stiffness of the un-piled footing indicated by the FE analysis is less than that observed (see Figures 4 and 7), suggesting that the operational G value for the footing is higher than the Plaxis input value of 8 MPa. The high initial footing stiffness shown experimentally (Test 6) is, however, inconsistent with the similarities of the responses seen in Figure 4 between the piled footings and pile only (Tests 2, 3 and 5). Further experimentation is required to investigate this anomaly. It is also of interest to note that, as with the Oasys ALP analyses, the FE analysis under-estimates the capacity of the single pile when the suT-bar values are used as the input undrained shear strengths. 5

CONCLUSIONS

This paper has presented the results of a pilot centrifuge study and supplementary numerical analyses carried out to investigate the benefits of employing a footing together with a pile to provide resistance to large lateral moments and loads typical of those applied to the foundations of offshore wind turbines, in a clay soil. For clay with a near surface undrained shear strength su ∼ 20 kPa, the contribution to both stiffness and ultimate capacity of a footing with a diameter four times that of the pile was shown to be low. This result was confirmed by 3D Finite Element analysis, which further indicated that the stiffness of the combined pile/footing foundation might be taken approximately as the sum of the stiffnesses of the individual components in isolation. This might suggest that the approach could be more successful in deposits with a greater surface stiffness and strength.

© 2011 by Taylor & Francis Group, LLC

ACKNOWLEDGMENTS The authors gratefully acknowledge the assistance provided by both Don Herley, the chief beam centrifuge technician at UWA, and Shainal Patel, a final year student at UWA in 2009. REFERENCES American Petroleum Institute (API) 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms-Working Stress Design, RP2A, Edition: 21, Washington. Brinkgreve, R.B.L. and Swolifs, W.M. (2009). Plaxis 3D Foundation manual, Version 2.2., Delft University of Technology, The Netherlands. Byrne, B. & Houlsby, G.T. (2003). Foundations for Offshore Wind Turbines. Philosophical transactions – Royal Society. Mathematical, Physical and engineering sciences, Vol. 361, no. 1813, 2909–2930. Dixon, R.K. (2005) Marine Foundations. WO 2005/038146 (Patent application). Lehane B.M., O’Loughlin C.D., Gaudin C. and Randolph M.F. (2009). Rate effects on penetration resistance in kaolin. Géotechnique 59(1), 41–52. Oasys (2006). Program manual for Geotechnical suite, Oasys, Newcastle, UK. Powrie, W., and Chandler, R. J. (1998). The influence of a stabilizing platform on the performance of an embedded retaining wall: a finite element study. Géotechnique 48(3), 403–409. Powrie, W., and Daly, M. P. (2007). Centrifuge modelling of embedded retaining walls with stabilizing bases. Géotechnique 57(6), 485–497. Randolph, M.F. (2004). Characterisation of soft sediments for offshore applications. Proc. 2nd Int. Conf. on geotechnical and Geophysical Site Characterisation, Porto, 1, 209–232. Stone K., Newson, T. and Sandon, J. (2007). An investigation of the performance of a ‘hybrid’ monopile-footing foundation for offshore structures. Proc. 6th Int. Conf. of the Society of Underwater Technology, London, 391–396. Taiebat, H., and Carter, J. P. (2002). A Failure Surface for the Bearing Capacity of Circular Footings on Saturated Clays. Numerical Models in Geomechanics – NUMOG VIII, Rome, Italy, A.A. Balkema, Rotterdam, 457–462.

610

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Design of monopile foundations in sand for offshore windfarms M. Saue & T.E. Langford Norwegian Geotechnical Institute, Oslo, Norway

N. Mortensen nmGeo, Denmark

ABSTRACT: Wind farm structures are dynamically loaded, which must be accounted for when defining soil properties for design. Depending on the load case, the load history may imply a drained, partially drained or undrained soil response which will lead to different dimensions for a selected supporting structure. This paper presents some analyses for a large diameter monopile in dense sand evaluated with both drained and undrained parameters extracted using the methodology described in Andersen (2009). Both simple limit equilibrium analyses and more complex 3D FE models are used to evaluate the pile capacity and load-displacement response. Results from the different methods are compared and discussed with respect to general trends and recommendations for better practice.

1

INTRODUCTION

The dimensions of large diameter monopile foundations for offshore wind turbines are governed by shear forces and overturning moments generated by wind and waves. The design of such a monopile should properly account for the cyclic nature of the loading when defining the stress strain response of the soil. The cyclic shear strength of the soil and the failure mode depend strongly on the stress path and the combination of average and cyclic shear stresses. A research project was established at NGI to develop knowledge and improve practice in offshore windfarm foundation design, starting from an initial literature review into the current state of practice. The project focus was large diameter monopiles in coarsegrained soils. Ultimate capacity (ULS) and operational load-displacement behavior (SLS) were evaluated for both ‘undrained’ (total stress) and ‘drained’ (effective stress) conditions. Analysis methods included simple limit equilibrium analyses and 2D/3D FE models. The analyses used two ‘model’ sands from NGI’s database, namely ‘Baskarp’ and a ‘North Sea’ sand. Environmental load histories were based upon data from a recent UK windfarm project.

2

to evaluate the load-displacement response under SLS conditions. The beam-spring approach is especially popular because it allows rapid analysis using simple finite difference programs. In many cases, standard API p-y curves are used to model the soil behaviour, although in some cases user-defined curves are also used. Nonetheless, little data is available to calibrate these curves for large diameter piles subjected to cyclic loading with a period of 3 to 4 s. Recent research has investigated the cyclic behaviour of piles, but this has generally been limited to a ‘drained’ soil response. Achmus et al. (2008) used cyclic drained triaxial tests on sand to evaluate the cumulative displacements and change in stiffness of the soil-pile system. This approach is more suitable for a case with a large load period; however it is realistic for many windfarm monopile design cases, even in coarse sand where the load period may be just a few seconds and the pile diameter may be 5 m.

BACKGROUND

2.1 Evaluation of current state of practice The current industry practice for monopile design in sands uses a combination of beam-spring and Finite Element models to evaluate the necessary pile length under ULS conditions. The same approach is then used © 2011 by Taylor & Francis Group, LLC

611

2.2

Discussion of generic design input

Input for the current analyses was based on typical offshore windfarm projects. The pile diameter was given as 4.7 m, and the wall thickness was increased to avoid significant bending in the embedded section of the pile. The loads and load history were generously supplied from a project in UK waters with a soil profile of dense sand, as confirmed by CPT data. A generic reference profile of dense sand was therefore used for all analyses. Two ‘model’ sands were evaluated, namely North Sea sand and Baskarp sand. Both sands are classified as very dense sands. Some geotechnical parameters are listed in Table 1.

Table 1.

Geotechnical parameters for the two ‘model’sands.

Soil parameters Relative density, Dr %) Drained friction angle [◦ ] Grain size, D10 [mm] Grain size, D50 [mm] Maximum void ratio, emax Minimum void ratio, emin Permeability [m/s · 10−5 ] Angularity

Baskarp sand

‘North Sea’ sand

95–100 44.5 0.08 0.15 0.86 0.51 N/A Sub-angular to angular

80–95 39.0 0.08 0.15 1.07 0.58 1.0 to 1.5 Angular to sub-rounded

Figure 2. Normalized storm history grouped in load parcels for use in the pore pressure accumulation.

τcy /τa -ratios (0.5 to 1.5). A 600 second time history representing the peak period of the storm has been implemented. The duration of a storm would typically be significantly longer resulting in a larger number of cycles at different stress levels. Different storm durations should be considered in a real wind farm design.

3.2 Figure 1. M/Mmax versus time, idling, 50-year return period.

3

DEVELOPMENT OF UNDRAINED SAND PARAMETERS

In order to investigate the more likely ‘undrained’ soil response due to cyclic loading, the cyclic shear strength for ULS capacity has been evaluated based on methodology given in Andersen (2009). A methodology for evaluation of the cyclic stress-strain response of the soil was established for use in SLS loaddisplacement analyses. 3.1

Environmental loads and load history

Load histories were processed to derive the maximum design loads and group the load parcels. The so-called ‘idling’ case with 50-year environmental conditions in Figure 1 was used for the current study. Loads include both wind and wave loads at seabed. The characteristic horizontal load was given by Hmax = 4.2 MN. The corresponding overturning moment at seabed level was given by Mmax = 129 MNm. The cyclic shear strength of the soil should be evaluated as the sum of the average and cyclic shear stresses, τa and τcy . The average and cyclic load components derived from the load history indicate ratios between the cyclic and average stress components (τcy /τa ) in the range 2 to 10, suggesting cyclic loading is dominant in this case. The load cases reflecting normal operation showed lower maximum loads and smaller © 2011 by Taylor & Francis Group, LLC

Cyclic pore pressure accumulation

The drainage condition around a top loaded monopile is complex. In addition to passive and active zones, there is a large rotational zone and tendency for the soil to flow laterally around the pile. NGI has developed a constitutive model for pore pressure accumulation and drainage to be used within FE analyses. However, there is currently no simplified drainage model accounting for pore pressure dissipation for such conditions. The current analyses therefore used a practical approach, whereby an equivalent number of undrained cycles (Neq ) was used to define a normalised cyclic shear strength. In order to provide justification for the approach described above, it was decided to perform a ‘simplified’ check of the pore pressure response for the North Sea sand case, based on the normalised storm history grouped in load parcels shown in Figure 2. This check was based on a pore pressure accumulation procedure considering both pore pressure generation and dissipation in a characteristic point selected in the passive failure zone at 5 m depth. The results show that the soil at the characteristic point acts almost fully undrained during one load cycle of typically 5 seconds, and only 10% difference between pore pressure generation and dissipation is observed during 10 load cycles. It is therefore believed that the soil will behave undrained during the maximum load cycle whereas a drained response may be seen for the average load. Results from the pore pressure accumulation procedure are presented in Figure 3. In this case, the derived Neq of about 10 is found for the combination of load history and cyclic contours in this case. A similar process for the contours based on CAU triaxial tests revealed a value of Neq closer to 1.

612

Figure 5. Cyclic contours for North Sea sand, N=1, average shear stress applied drained. Figure 3. End points from the pore pressure accumulation plotted with cyclic contour lines for DSS tests on North Sea sand.

Figure 6. Relationships between shear stresses and shear strains for τcy = 5 · τa and N = 1.

sand, as presented in Figure 6. The following criteria were used:

Figure 4. Cyclic contours for Baskarp sand, N=1, average shear stress applied drained.


τcy /σvc for τa = 0 based on test data

τcy /σvc constant at γcy = 15% for τa /σvc ≤ 0.15
τa /σvc = 0.60 (assumed) at γa = 15% and τcy = 0

τa /σvc for τcy > 0 scaled up to maximum τcy /σvc

3.3 Derivation of sand parameters for the ‘undrained’ model

• • • •

Two ‘model’ sands were used for evaluation of cyclic undrained shear strength properties, namely Baskarp sand and North Sea sand, with typical properties given inTable 1. More cyclic test data is available for Baskarp sand (e.g. Andersen and Berre, 1999). However, the sand is very angular and dilatant in shear, mobilizing large shear stresses. The results may therefore overestimate the performance of many offshore sands which are more rounded. Figures 4 and 5 present cyclic contour plots for Neq = 1 for the Baskarp and North Sea sands, respectively. The North Sea sand shows a weaker response, and may be considered more typical of dense offshore sand, with a relative density of about 85%. However, there is limited data relating to the cyclic performance of the North Sea sand when subjected to a drained average shear stress. Some experience-based judgement was therefore required to derive cyclic contour plots for this

The normalised ‘cyclic’ shear strength is given by

the expression τf ,cy /σvc = (τcy,f + τa,f )/σvc, which is based on a ratio of average to cyclic shear stresses, in this case taken as τcy = 5 · τa . The resulting maximum normalised cyclic shear strength for Neq = 1 is 1.50 for
Baskarp sand and 0.74 for the North Sea sand. τf ,cy /σvc ratios for N = 10 are 0.94 and 0.55 for the Baskarp sand and the North Sea sand, respectively. These values have been used in calculating the maximum pile capacity (ULS). The cyclic test data for both the North Sea and Baskarp sands was extracted from tests run at relatively high consolidation stresses to represent the stress conditions below a GBS. Higher normalised strengths may be expected for the consolidation stresses typical for windfarm monopiles. The parameters for the Baskarp sand are limited to data for shear strains of 2 to 3%. A significantly higher cyclic strength may

© 2011 by Taylor & Francis Group, LLC

613

be expected for larger strains due to the sand’s highly dilatant behaviour. However, the designer should judge whether such dilation should be included. This choice will be dependent upon many factors, including the complexities of the failure mechanism and soil anisotropy. For shallow water, cavitation may govern the pore pressure response. Stress strain curves may be established from the cyclic contour plots for use in load displacement response analyses (SLS), assuming the same ratio of τcy = 5 · τa . Figure 6 shows these curves for both ‘model’ sands assuming Neq equal to 1. 4

CAPACITY CALCULATIONS

The minimum required monopile penetration has been calculated for both drained (effective stress) and undrained (total stress) conditions using the same maximum load. A load factor of γL = 1.35, and material factors of γm = 1.20 (drained) and 1.3 (undrained), were applied, according to DNV (2004). Note that the Table 2. Results from ULS analyses for drained parameters for a 4.7 m diameter monopile. Required pile penetration [m]
◦

φ[ ]

API p-y

Brinch Hansen

3D Plaxis

35 40

21.0 19.0

19.5 17.2

15.5 13.0

material factors have recently been decreased (1.25) for undrained analyses and (1.15) for drained analyses. The calculations included the Brinch Hansen (1961) method, API (2000) p-y curves within a finite difference beam-column code and Plaxis 3D Foundation. Undrained analyses were also performed using NGI’s in-house quasi-3D FE program Bifurc. 4.1

Drained analyses may often be used for monopile design within windfarm projects. Drained analyses were therefore included as a reference in the project, based on an effective unit weight of 10 kN/m3 and friction angles of 356◦ and 40◦ , representing dense to very dense sands according to API (2000). The Plaxis 3D analyses used an elastic-perfectly-plastic material with Mohr Coulomb failure criterion. Associated flow (ψ
= φ
) was assumed. Table 2 presents results from the different analyses. Figure 7 shows contours of total shear strain at failure for 15.5 m penetration (Plaxis 3D). The failure mode is a combination of passive, active and rotational failure. Passive failure zones have developed along the front of the pile and at the bottom rear side of the pile where the pile kicks back. A smaller active failure mode forms at the upper rear part of the pile. This complex failure mechanism is very sensitive to assumptions regarding soil dilation. The assumption of associated flow included in these analyses is considered non-conservative in this case. 4.2

Figure 7. Total shear strain at failure for a 15.5 m long pile, drained ULS analyses applying associated flow (loads applied to right and clockwise). Table 3.

Drained analyses

Undrained analyses

Undrained capacity analyses have been performed for the Baskarp and the North Sea sands, assuming a ‘clay’ type total-stress model. The equivalent static undrained shear strength profile is based on the cyclic DSS test results discussed above. The potentially significant anisotropic behaviour of the sands has therefore not been considered at this stage in the undrained analyses. The required pile penetrations calculated using the different methods are summarized in Table 3, together with the normalised strength,
, used in each case. τf ,cy /σvc The results show a significant difference in required pile penetration and monopile capacity depending on the calculation approach used. The more simplified (Brinch Hansen and API p-y curves) methods typically

Results from ULS analyses for undrained parameters for a 4.7 m diameter monopile. Required pile penetration [m]

Soil type

Neq


τf ,cy /σvc

API p-y

Brinch Hansen

Birfurc 2D1)

3D Plaxis

Baskarp sand

1 10 1 10

1.50 0.94 0.74 0.55

19.5 24.5 26.5 28.0

20.3 23.9 25.9 28.8

16.5 19.5 21.5 24.0

16.5 19.5 21.5 24.0

North Sea sand

1)

The FE program Bifurc 2D accounts for 3D effects by use of shape factors. Results are presented for shape factors of unity

© 2011 by Taylor & Francis Group, LLC

614

Figure 9. CAD and CAU triaxial tests on Baskarp sand compared to simulation of tests in Plaxis with HSS model.

Figure 8. Incremental displacements at failure in a selected cross section of the Plaxis 3D FE model for a 16 m long pile.

result in a lower undrained capacity compared to FE analyses (note that the drained capacity from the simplified methods is higher). Figure 8 presents the typical failure mechanism showing near-symmetrical active and passive zones in the upper half of the pile, with rotational failure in the lower part. One reason for the range in results is that the simplified methods are based on smaller diameter piles and do not directly account for the contribution from base shear, which may be significant for a large diameter monopile. Another reason is that the API method includes the ultimate resistance (pu ) increasing from around 3· su at seabed to a maximum of 9·su at greater depths. This increase with depth is highly dependent on the depth-diameter ratio. In the case of the Baskarp sand analyses for Neq = 1 with API p-y curves, the ultimate resistance increases from 3·su at the surface to around 6·su at 20 m depth. For a 4.7 m diameter pile, the penetration would need to be around 50 m long to reach an ultimate resistance of 9·su . These values of ultimate resistance are therefore low compared to FE analyses which typically include a maximum lateral resistance of 9 to 12·su for depths where lateral flow-around occurs. As discussed earlier, many monopile foundations are designed using drained analyses. This approach has developed from typical offshore practice for smaller diameter foundation piles; however the extrapolation to much larger pile diameters and higher frequency loading is not well understood. The only input parameters applied in the simplified effective stress calculations are the submerged unit weight and effective friction angle. The API p-y curves account for cyclic loading by multiplying the static soil resistance by a factor A = 0.9. Using A = 0.9 implies a reduction in the static resistance of 70% at seabed level, 55% at a relative depth (depth below seabed normalized with respect to pile diameter) of 1.25 and 0% at a relative depth of 2.63. Otherwise, the API-method appears to assume that the sand acts fully drained during cyclic loading. This may be suitable for slender jacket piles for which the calculation methods were originally developed. However, these assumptions are not realistic for large diameter monopiles, as © 2011 by Taylor & Francis Group, LLC

demonstrated by pore pressure accumulation analyses discussed previously. Even though the results presented here may appear to suggest a drained approach gives pile penetrations which are similar to or more conservative than an undrained approach, this may not be representative for other cases. Rather, the main purpose of the work presented here is to demonstrate the importance of including the cyclic behaviour of the sand within the design of a monopile foundation. 5

LOAD-DISPLACEMENT ANALYSES

The load-displacement response (SLS) of the monopile foundation has also been considered. In principle, a similar approach should be used as for the capacity analyses, accounting for the response of the foundation system to cyclic loading. Such an approach would use stress-strain relationships based on the results of cyclic testing, as shown earlier for the two ‘model’ sands. Analyses should account for non-linear stiffness, anisotropy and permanent deformations due to long-term loading. At this stage, the analyses presented herein are limited to drained analyses, including the beam-column approach with API (2000) p-y curves currently used as standard practice in many commercial projects. However, this is not considered appropriate for monopiles subjected to high frequency cyclic loading. 5.1

Drained analyses

The drained response has been evaluated with a finite difference beam column code using API (2000) p-y curves.Additionally, FE analyses have been performed applying the Hardening Soil model with Small Strain (HSS). A penetration of 16.0 m has been modeled, based upon the capacity analyses for Baskarp sand. Input parameters for the HSS model have been calibrated against drained and undrained static triaxial tests as presented in Figure 9. Figure 10 shows the pile displacement versus depth based on the API (2000) p-y curves and the Plaxis 3D analysis with HSS constitutive model. The range of

615

• The soil response will depend on the load history.

Depending on the load case considered, the soil may act undrained or drained, but it may not be realistic to use the same maximum design load for different types of soil response. • The drainage conditions in a sand/silt are crucial in design. If the pore water cannot drain away within half a typical load period, the soil must react undrained for the maximum load.

Figure 10. Drained pile displacement response (SLS).

pile response for the API p-y curves compares well with the results from the FEM analyses using Baskarp HSS parameters. It should be noted that Baskarp sand is highly angular and dilatant in shear, which gives a different response to many typical offshore sands. 5.2

Undrained analyses

Undrained analyses have also been performed in Plaxis 3D using an in-house multi-linear elastic constitutive model applying the stress strain behaviour derived from the contour diagrams such as presented in Figure 6. Total displacements are based on analyses for the average and cyclic load components. Results will be published after further internal validation. It should be noted that load-displacement criteria may be governing for design, since the limiting rotation of the nacelle may be limited to perhaps 0.25 degrees. Analyses shall include both the cyclic and permanent shear strains, as well volumetric strains which may result from long term loading. 6

CONCLUSIONS

This paper has presented a framework for evaluating the capacity of monopile foundations in dense to very dense sand assuming cyclic loading. Such an approach is required to improve the understanding of the empirical design rules adopted through the API p-y model. It may be difficult to draw clear conclusions from a rather limited number of analyses, but it has been indicated that: • The API p-y approach is an empirical design method

which may not realistically capture all relevant mechanisms for such large diameter piles. More advanced models exist which consider both the structure and soil response.

© 2011 by Taylor & Francis Group, LLC

A detailed design of a large wind farm may consider more refined analyses for a limited number of turbine locations. Based on these positions, a p-y type approach may then be calibrated to fit the results from the detailed analyses. Such a calibration should focus as a minimum on establishing the correct stress distribution along the pile and a realistic centre of rotation. Although the focus of this paper is on evaluation of capacity, the load displacement response may be governing for design. ACKNOWLEDGEMENTS The research project covering this work was partly financed by the Norwegian Research Council and partly by nmGeo, Denmark. Furthermore, the authors wish to acknowledge input and advice from colleagues at NGI and others within the industry, especially Knut H. Andersen and DONG Energy. REFERENCES Achmus, M., Abdel-Rahman, K. and Kuo,Y.-S. 2008. Design of Monopile Foundations for Offshore Wind Energy Plants. 11th Baltic Geotechnical conference -Geotechnics in maritime engineering, Gdansk, Poland, Vol. 1, pp. 463–470. Andersen, K.H. 2009. Bearing capacity under cyclic loading – offshore, along the coast, and on land. The 21st Bjerrum Lecture presented in Oslo, 23 November 2007. Canadian Geotechnical Journal, Volume 46, No. 5. Andersen, K.H., and Berre, T. 1999. Behaviour of a dense sand under monotonic and cyclic loading. In Proceedings of the 12th ECSMGE, Geotechnical Engineering for Transportation Infrastructure, Amsterdam, the Netherlands, 7–10 June 1999. A.A. Balkema, Rotterdam, the Netherlands. Vol. 2. pp. 667–676. API 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms-Working Stress Design. American Petroleum Institute, Washington, D.C., 21st ed., 2000. Brinch Hansen, J., 1961. The ultimate resistance of rigid piles against transversal forces. Bulletin No. 12, Danish Geotechnical Institute, Copenhagen, Denmark, 5–9. DNV 2004. DNV-OS-J101, Design of Offshore Wind Turbine Structures, 2004.

616

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Experimental evaluation of backfill in scour holes around offshore monopiles S.P.H. Sørensen, L.B. Ibsen & P. Frigaard Aalborg University, Denmark

ABSTRACT: When a monopile is installed in sandy or silty soil a scour hole will form around the pile due to erosion. A scour protection consisting of rock infill can be employed to protect against the forming of a scour hole. As scour protection is highly expensive the most economic solution might be to allow the forming of a scour hole and hereby design the monopile with a larger penetration depth. The depth of the scour hole will change over time as the scour depth will increase when currents are dominating and backfilling of the scour hole will take place when waves are dominating. Several researchers have investigated the time scale of erosion in small scale experiments and found a dependency on Shields parameter and the Keulegan-Carpenter number. Further knowledge is however needed in larger scale. Regarding the rate of backfill, further knowledge is needed for both small and large scale. Currently there is no knowledge concerning the relative density, and hereby also the strength and deformation properties of the backfilled soil material. The strength of the backfilled soil and the time scale of respective erosion and backfilling is of high importance when designing the steel structure of the foundation for fatigue. A backfill test has been performed in the Large Wave Channel (GWK) of the Coastal Research Centre (FZK) in Hannover. The relative density of the backfilled soil material has based on soil samples and CPT measurements been determined to be in the range of 60–80%. The normalized time scale of backfilling is found to be small in comparison with existing small scale experiments.

1

INTRODUCTION

For offshore wind turbines several types of foundation exist. The choice of foundation depends on the site and loading conditions. The monopile foundation concept, in which a pile made of welded steel is driven open-ended into the soil, is often employed. Typically, the pile diameter, D, is in the range of 4–6 m and the embedded pile length, L, around 20–25 m. Hereby the slenderness ratio, L/D, has a magnitude around 5. Around monopiles installed in silty or sandy soil erosion will take place. The waves and current can result in the forming of local scour around the monopiles. The depth of these scour holes can according to the design regulations, e.g. DNV (2004), be up to 1.3 times the pile diameter. Due to the low slenderness ratio typically employed for these monopile foundations the depth of the scour hole can be up to 25% of the embedded pile length. When designing monopiles situated in sandy or silty soil scour protection consisting of rock infill is therefore often used. Scour protection is highly expensive and the most economic solution might therefore be to design the monopiles without scour protection and hereby allow the forming of scour holes. A larger penetration depth is however needed for monopiles designed without scour protection to account for the loss of soil resistance caused by the scour hole. © 2011 by Taylor & Francis Group, LLC

Due to changing sea conditions the depth of the scour hole will change over time. Hereby, also the total stiffness of the monopile foundation will be time dependent. Today the variation of the total stiffness of the foundation is not taken into account when designing the steel pile for fatigue. Instead the depth of the scour hole is taken as a constant value corresponding to the maximum scour depth, resulting in a conservative design of the wall thickness of the monopile. In order to incorporate the variation of scour depth further research concerning the time scale of backfilling is needed in order to estimate the variation of the scour depth with time. Moreover, research is needed concerning the relative density, and hereby also the strength and deformation properties of the backfilled soil material. Typically sand in the marine environment is packed rather dense due to the wave action. Internal friction angles above 40◦ are therefore not unusual. Augustesen et al. (2009) has reported friction angles between 36.6◦ and 45.4◦ at Horns Rev, Denmark. In the present paper the rate of the backfilling and the relative density of the backfilled soil material are assessed on the basis of experiments at the Large Wave Channel (GWK) of the Coastal Research Centre (FZK) in Hannover, Germany.

617

2

SCOUR PHENOMENON

In river hydraulics the scour phenomenon has been studied intensively for bridge piers, cf. Breusers & Raudkiwi (1991) and Melville & Coleman (2000), as the forming of scour holes have proven to be an important cause of failure. In the coastal and offshore region, the scour phenomenon is complex as both waves and currents cause scour. Sumer & Fredsøe (2001) studied the equilibrium depth of scour holes, S∞ , for combined current and wave conditions by conducting small-scale tests in wave channels. They found that the equilibrium depth of the scour holes depends on the dimensionless parameter, Ucw , and the Keulegan-Carpenter number, KC. Ucw is given by Equation 1 where Uc is the undisturbed current velocity determined D/2 above the seabed and Um is the maximum undisturbed orbital velocity at the seabed.

θcr , sediment transport is initiated. According to DNV (2004) the critical value of Shields parameter is about 0.05–0.06. As the equilibrium depth of the scour hole depends on whether the sea conditions are current or wave dominated, the scour depth will change with time. Hereby also the total stiffness of the monopile will change with time. The timescale of scour has been investigated by several authors, e.g. Sumer et al. (1992), Fredsøe et al. (1992), Sumer et al. (1993) and Hartvig et al. (2010). According to Hartvig et al. (2010) the time variation of the scour depth can be approximated by:

Where S0 denotes the initial scour depth, t denotes the time and T is defined as the timescale of scour representing the time period during which substantial scour takes place. Equation 7 is valid for both the scour and backfilling process. The normalised time scale, T
, is defined as:

The Keulegan-Carpenter number is for regular waves calculated as:

where T is the wave period. According to Sumer & Fredsøe (2002) the equilibrium scour depth for a combined current and wave condition can be estimated as:

According to Sumer et al. (1992) and DNV (2004) the normalised time scale is for current dependent on the water depth compared to the pile diameter and Shields parameter, cf. Equation 9, and for waves dependent of Shields parameter and the KeuleganCarpenter number, cf. Equation 10.

where S∞, c is the equilibrium depth of the scour hole for current only. A and B are estimated from Equation 4 and 5, respectively. According to DNV (2004) S∞, c have a magnitude around 1.3.

Equation 3 is valid when Equation 6 is fulfilled. Further, Equation 3 is only valid for Keulegan-Carpenter numbers in the range of 6 < KC < 26.

θ is a dimensionless parameter termed Shields parameter, Uf is the maximum undisturbed bed-friction velocity, g is the acceleration due to gravity, s is the specific grain density defined as the ratio between the grain and water density and d is the specific grain size. The median grain size, d50 , can according to DNV (2004) be employed as the specific grain size. When Shields parameter is equal to a critical value, © 2011 by Taylor & Francis Group, LLC

618

Research regarding the time scale of backfill has only been conducted to a minor extent. Fredsøe et al. (1992) investigated the time scale of backfilling below pipelines. The conclusions were that the time scale decreases for increasing Shields parameter. Further the time scale is high when the change in wave climate results in a large difference in the Keulegan-Carpenter number. Hartvig et al. (2010) investigated the time scale of backfilling around piles in small scale and concluded that the time scale for the backfilling process is roughly 10 times longer than the erosion process assuming similar values of the Shields parameter. 3 TEST SETUP Experimental tests of the backfill rate and of the relative density of backfilled sand have been conducted at the Large Wave Channel (GWK) of the Coastal Research Centre (FZK) in Hannover, Germany. The length, width, and height of the wave channel are respectively, 324 m, 5 m, and 7 m. A piston-type wave

generator with a capacity of 900 kW made it possible to generate waves with periods between 1 and 15 s and heights up to 2.5 m. A cylindrical pile with an outer diameter of 0.55 m has been fixed to the bottom of the wave channel. Hereby the geometric scale between the model pile and the target field pile is in the range of 1:7-1:11. The pile was fixed in the centre of the wave channel. Near the pile well-sorted sand with a depth of 1 m was situated. The sand had a median grain size, d50 , of 0.15 mm. The relative density was not measured prior to the testing, but as no compaction of the sand had taken place, the sand was expected to be rather loose. The water level during the tests was 4 m above the bottom of the channel corresponding to a water depth of 3 m near the pile. In order to measure the wave parameters wave gauges were installed in several places along the length of the channel. Further, wave gauges were installed around the pile. The procedure for the tests is as follows:

Figure 1. Dimensions of the scour hole.

• At first when no water was in the channel, a scour

hole was manually prepared around the pile. • Water was led into the channel. The wave gauges

were calibrated by respectively raising and lowering the water level. • The waves were generated and the scour-depth was regularly measured in three positions. • Water was drained from the channel. • Soil samples and cone penetration tests (CPT) were conducted. 3.1

Preparation of scour hole

A scour hole around the pile was prepared manually. The excavated soil material was distributed evenly to the surrounding soil ensuring a plane soil surface outside the scour hole. The approximate dimensions of the scour hole are shown in Figure 1. A scour depth of approximately 0.7 m has been employed corresponding to 1.3D, where D is the pile diameter. Hereby the depth of the scour hole corresponds to current dominated conditions. The ratio between the width of the channel and the width of the scour hole is around 2. Minor boundary effects from the channel wall might therefore take place.

Figure 2. Forming of stream-channel during the water fill-in causing minor damages to the scour hole.

was approximately KC = 9 and Shields parameter approximately θ = 0.41. Hereby, the equilibrium depth of the scour hole was expected to be close to zero. No current was generated during the tests. The depth of the scour hole was measured in three places meanwhile the waves were generated, cf. Figure 3. The depth was measured manually by a large weight connected to a string. Waves were generated over a time period of 25 minutes.

3.2 Water led-in After the preparation of the scour hole, water was led into the channel. The water was led slowly into the channel until the scour hole was filled with water in order to minimize the damages of the scour hole. However, minor damages occurred to the scour hole while water was led into the channel causing deposition of sand into the scour hole, cf. Figure 2. 3.3 Running of waves After the calibration of the wave gauges the tests were initiated. Irregular wave series with significant wave heights, Hm0 , of 1.06 m, peak periods, Tp , of 5.9 s and peak enhancement factors of γ = 3.3 were generated. Hereby, the Keulegan-Carpenter number © 2011 by Taylor & Francis Group, LLC

3.4

Soil samples and CPT’s

Soil samples and CPT’s were taken both near the pile and away from the pile. The soil samples and CPT’s within the boundaries of the scour hole are shown in Figure 3. The soil from the soil samples has been tested in the Laboratory of Geotechnical Engineering at Aalborg University where the relative density, ID , of the samples has been determined. A laboratory CPT-cone has been used for the CPT’s. A sectional view of the CPT-cone is shown in Figure 4. The test setup for the CPT’s is shown in Figure 5. The CPT-cone was pressed into the sand material by a

619

Figure 5. Test setup for the CPT’s.

Figure 3. Position of soil samples, CPT’s and measurements of scour depth.

Figure 4. Sectional view of the laboratory CPT-cone. Measures in mm. Table 1. Variation of scour depth. The positions of the scour depth measurements, SD 1-3, are shown in Figure 3. Time [min]

SD 1 [m]

SD 2 [m]

SD 3 [m]

0 5 10 15 20 25

0.10 0.10 0.10 0.00 0.01 0.02

0.30 0.20 0.12 0.01 0.04 0.02

0.38 0.25 0.16 0.05 0.01 0.02

hydraulic piston mounted on a frame. Heavy concrete blocks were used as counterweight.

4

RESULTS

The scour hole before and after the running of waves is shown in Figure 1 and Figure 6, respectively. It can be observed that almost the entire scour hole has been filled with backfilled sand material. From Equation 3 employing KC = 9 the expected scour depth can be estimated to S∞ = 0.06 m. Hereby the observed equilibrium scour depth is in good agreement with Equation 3. © 2011 by Taylor & Francis Group, LLC

620

Figure 6. Scour hole after the generation of waves.

4.1 Time scale of backfill During the generation of waves the scour depth was measured regularly at three positions. The variation of the scour depth with time is shown in Table 1. After approximately 10 min the scour hole has approached the equilibrium scour depth and the time scale of backfill is therefore estimated to 10 min. The variation of scour depth at SD 3 is illustrated in Figure 7, where also the variation of scour depth calculated using Equation 7 is indicated. A timescale of 10 min has been employed. The normalised time scale can be determined to 0.0147 by use of Equation 8. Assuming that the scour hole had been established by currents the normalised time scale for erosion can be estimated to 0.02 by use of Equation 9 and θ = 0.41. Hereby the time scale for the backfilling process is in the same order as the expected time scale for the erosion process. This is in contrast to the results presented by Hartvig et al. (2010), where the normalised time scale of the backfilling process was found to be around a factor of 10 larger than the normalised time scale for the erosion process. However, it should be emphasized that Equation 9 has been determined based on small scale experiments and therefore might be out of range when employed for D = 0.55. For KC = 3 and θ = 0.2 Hartvig et al.

Table 2. Relative density determined from soil samples. Soil sample 1–5 is taken within the scour hole and the position of these soil samples are shown in Figure 3. Soil sample 6 and 7 are taken outside the boundary of the scour hole. Soil sample

Relative density, ID

1 2 3 4 5 6–4 m behind pile 7–5 m behind pile

70% 85% 90% 80% 76% 32% 40%

Figure 7. Variation of scour depth normalized with the initial scour depth at SD 3. The continuous curve is calculated using Equation 7 with a time scale of 10 min. The position of SD 3 is shown in Figure 3.

(2010) determined the normalised time scale to 4.0 for small scale experiments with D = 0.1 m. The normalised time scale is hereby approximately 250 times smaller for the experiments at GWK, although the Keuler-Carpenter number and Shields parameter is in the same order. Hereby, scale effects that are not included in Equation 8 might be introduced when changing the scale of the experiments. During the first 10 min of waves the scour depth in position SD 1 does not change. Hereby, it seems that the deposition of backfill for the first 10 min is concentrated at the base of the pile. After the first 10 min the deposition of backfill takes place at all three positions. This observation is in agreement with observations from small scale experiments reported by Hartvig et al. (2010). It should be emphasized that the method employed for the measuring of the scour depth is rather inaccurate. This inaccuracy can be observed in Table 1 as the scour depth at all positions alternately increases and decreases after 15 min of waves. 4.2

Relative density of backfilled sand

The relative density of the soil was determined based on the intact soil samples and is shown in Table 2. The relative density of the samples taken outside the boundary of the scour hole is in average 36% corresponding to rather loose sand. However, the sand was not compacted prior to the tests and a loose soil compaction was therefore expected. An average relative density of 80% has been obtained from the soil samples taken within the boundary of the scour hole corresponding to dense sand. The results from the CPT’s are used to predict the variation of relative density with depth. The relative density from the CPT’s is estimated as:

© 2011 by Taylor & Francis Group, LLC

Figure 8. Variation of relative density with depth determined by CPT. The position of the CPT’s is shown in Figure 3.

where c1 , c2 and c3 are constants, qc is the tip resistance
and σv0 is the effective vertical stress. Equation 11 has been proposed by Ibsen et al. (2009). The constants c1 and c3 are given as 0.75 and −0.42, respectively. The constant c2 is calibrated to the relative density obtained from the soil samples. In the calibration of c2 , a relative density of 80% at a depth of 100 mm is required for the CPT’s carried out within the boundary of the scour hole. The value of c2 is determined to 4.0. It should be emphasized that the interpretation of the CPT’s as given in Equation 11 is only valid for depths larger than 100 mm. At depths smaller than 100 mm a different failure mode appears in the soil. In Figure 8 the variation of relative density of the backfilled soil material with depth, z, is shown. The relative density of the backfilled soil is found to decrease from approximately 80% near the soil surface to approximately 60% at a depth of 400 mm, i.e. at the bottom of the original scour hole. This observation might be caused by the fact that the backfill rate decreases as the scour depth approaches the equilibrium scour depth. Hereby, the grains have more time to rotate into a dense configuration at the surface compared to the bottom of the original scour hole. However, it should be emphasized that the low number of conducted CPT’s causes large statistical

621

uncertainties regarding the variation of the relative density with depth. The relative density of the backfilled soil material determined based on the soil samples and CPT’s is in comparison with typical values for sand in the marine environment of the same magnitude or slightly smaller. 5

CONCLUSION

This paper presents an experimental evaluation of the backfill rate in a sea condition with no current. Further, the relative density of the deposited backfill material has been investigated. The tests have been conducted at the Large Wave Channel (GWK) of the Coastal Research Centre (FZK) in Hannover. The major conclusions that can be drawn from the experiments are: • Compared to the studies of Hartvig et al. (2010) the

• •

•

•

normalized time scale was found to be a factor of approximately 250 smaller although the KeuleganCarpenter number and Shields parameter were in the same order for the two studies. Hereby, scale effects that are not included in Equation 8 might be introduced when changing the scale of the experiments. During the backfill process, the sand material was at first deposited near the base of the pile. From soil samples and CPT-measurements the relative density of the backfilled soil deposit was found be approximately 80% near the surface. Near the bottom of the original scour hole, the relative density was determined to approximately 60%. The denser compaction near the soil surface might be caused by the decrease in backfill rate as the scour depth approaches the equilibrium state. As the backfilled soil deposit can be expected to be rather dense, the total stiffness of the foundation can be expected to increase by a large amount when the sea conditions changes from current dominated to wave dominated. If accounting for the variation of the total stiffness of the foundation in the fatigue limit state, large savings in the amount of steel used for the monopile might therefore be the result. Further research is needed regarding the time scale of backfill and the relative density, e.g. a broad spectrum of wave and current conditions needs to be examined. It needs to be examined how the relative density of the soil varies during the backfill process.

© 2011 by Taylor & Francis Group, LLC

ACKNOWLEDGEMENTS The authors would like to acknowledge the assistance and support provided by staff of the Large Wave Channel (GWK) of the Coastal Research Centre (FZK) in Hannover, Germany. The experiments in the Large Wave Channel (GWK) were supported by the European Community under the project “Load on entrance platforms for offshore wind turbines, Hydrolab III-GWK-02 (6th EU Framework Programme). REFERENCES Augustesen, A. H., Brødbæk, K. T., Møller, M., Sørensen, S.P.H., Ibsen, L. B., Pedersen, T. S. & Andersen, L., 2009. Numerical Modelling of Large-Diameter Steel Piles at Horns Rev, Proceedings of The Twelfth International Conference on Civil, Structural and Environmental Engineering Computing, September 1 to September 4, Funchal, Madeira, Portugal, 239. Breusers, H. N. C. & Raudkiwi, A. J., 1991. Scouring: Hydraulic structures design manual Series, No. 2. Routledge, Taylor & Francis Group, London, England. DNV, 2004. Design of Offshore Wind Turbine Structures – Offshore Standard, DNV-OS-J101. Det Norske Veritas, Norway. Fredsøe, J., Sumer, B. M. &Arnskov, M. M., 1992. Time Scale for Wave/Current Scour Below Pipelines, International Journal of Offshore and Polar Engineering, 2(1), 13–17 Hartvig, P. A., Thomsen, J. M., Frigaard P. & Andersen, T. L., 2010. Experimental study of the development of scour & backfilling. In print. Ibsen, L. B., Hanson, M., Hjort, T. H. & Thaarup, M., 2009. MC-Parameter Calibration for Baskarp Sand No. 15. DCE Technical Report No. 62, Department of Civil Engineering, Aalborg University, Denmark. Melville, B. W. & Coleman, S. E., 2000. Bridge Scour, Water Resources publications, Colorado, USA. Sumer, B. M., Christiansen, N. & Fredsøe, J., 1992. Time Scale of Scour around a Vertical Pile, Proceedings of the Second International Offshore and Polar Engineering Conference, San Francisco, USA, 308–315 Sumer, B. M., Christiansen, N. & Fredsøe, J., 1993. Influence of Cross Section on Wave Scour around Piles, J. Waterway, Port, Coastal and Ocean Eng., 119(5), 477–495. Sumer, B. M. & Fredsøe, J., 2001. Scour around Piles in Combined Waves and Current,ASCE Journal of Hydraulic Engineering, 127(5), 403–411. Sumer, B. M. & Fredsøe, J., 2002. The mechanics of scour in the marine environment. Singapore: World Scientific Publishing Co. Pte. Ltd. Reprinted 2005. ISBN 981-024930-6.

622

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

An investigation of the use of a bearing plate to enhance the lateral capacity of monopile foundations K.J.L. Stone University of Brighton, Brighton, UK

T.A. Newson, M. El Marassi & H. El Naggar University of Western Ontario, Canada

R.N. Taylor & R.J. Goodey City University London, London, UK

ABSTRACT: Current offshore foundation technology is being transferred successfully to the renewable energy sector, but there is clearly scope to develop foundation systems that are more tuned to the needs of the renewable power systems such as wind powered generators. One such approach would be for foundation systems that combine several foundation elements to create a ‘hybrid’ system. In this way it may be possible to develop a foundation system which is more efficient for the combination of vertical and lateral loads associated with wind powered generators. This paper reports a series of small-scale centrifuge model tests designed to investigate if the hybrid system offers a significant advantage in terms of lateral and axial load capacities to a conventional monopiled foundation. It is apparent that the lateral response of a single monopiled foundation can be enhanced by the presence of a bearing plate. Whilst the effect on the initial lateral stiffness was not observed to be significant, the lateral stiffness beyond this initial movement was significantly enhanced.

1

INTRODUCTION

Offshore foundation systems are constantly evolving to meet the needs of developments in the energy sector. These developments may be induced by the requirements of moving into ever deeper water for hydrocarbon recovery, or creating foundation systems for renewable energy sources such as offshore wind farms. These massive turbine structures induce complex loading patterns on foundations as the result of combined wind, wave and self weight loading effects, all of which must be accommodated within very small displacement envelopes to allow the turbines to operate. The loading requirements associated with wind turbines are characterised by relatively low vertical loads but high overturning moments. As such, conventional shallow foundations are not well suited to resist these loads, although in some instances gravity base units have been successfully employed. The preferred foundation solution is a monopile. These foundations are attractive because they can be employed in a variety of soil conditions that might exist over the large areas occupied by commercial wind farms. Their design also draws on much offshore geotechnical experience in the design of cyclically loaded piles (e.g. Poulos, 1988). This paper investigates the feasibility of enhancing the lateral capacity of a monopile through the use of an integral bearing plate. In essence, the foundation system can be considered as a monopiled foundation © 2011 by Taylor & Francis Group, LLC

623

Figure 1. Schematic representation of prototype monopiled footing foundation.

plate under combined horizontal, vertical and moment loading as is shown schematically in Figure 1. The proposed foundation system has a 2dimensional analogy in the case of a retaining wall with a stabilising base (Powrie and Daly, 2007), and in a 3-dimensional case that of a single capped pile. Poulos and Randolph (1983) developed methods for analysing the relative influence of the pile and pile cap

under axial loading, and some studies of the influence of the pile cap on the lateral performance of single piles has also been reported by Kim et al. (1979), Mokwa and Duncan (2001, 2003) and Maharaj (2003). The dimensions of pile caps are generally relatively small, but the role they play in determining restraint conditions at the pile head is significant. It has also been demonstrated that for individual piles, the presence of a relatively thick pile cap can provide a significant contribution to lateral resistance through the development of passive soil wedges. A similar resistance mechanism develops where a skirt is present (Bransby and Randolph, 1998). By studying the performance of the monopiled footing it may be possible to relate the response of the system to the cumulative effect of the constituent elements namely, the lateral response of the single pile and the bearing capacity of the footing. Much work has been undertaken on the former and reported, for example, by Matlock and Reese (1960), Broms (1964), Poulos (1971), Reese et al. (1974), Randolph (1981), Duncan et al. (1994) and more recently Zhang et al. (2005). Similarly the bearing capacity problem has been investigated for complex loading conditions associated with offshore foundations by, for example, Houlsby and Puzrin (1999) and Gourvenec and Randolph (2003). Recent studies reported by Stone et al. (2007), Stone et al. (2010) and El-Marassi et al. (2008) have indicated that the lateral capacities of a single pile can be considerably increased through the use of a foundation plate. Figure 2 shows a typical set of single gravity tests results (after Stone et al. 2007) and illustrate that the lateral capacity of a 10 mm diameter pile is increased in excess of 50% through the use of a 50 mm diameter foundation plate. This paper builds on these single gravity model studies and extends the physical modeling to a series of centrifuge model tests. As in the single gravity model studies, the model tests are intended to investigate the influence of the bearing plate on the lateral load response of the monopile. The centrifuge model was instrumented with strain gauges to provide some qualitative information concerning the contribution to lateral resistance from the soil reaction on the pile. The lateral resistance contributed by the bearing plate can then be deduced by considering the overall lateral capacity of the combined pile and bearing plate.

The critical state friction angle, determined from direct shear testing, was 32 degrees. The model foundation system was fabricated from a 19 mm diameter thin-walled, open-ended steel tube (t = 0.5 mm) and instrumented with foil strain gauges at four locations along its length (refer to Figure 3a). The bearing plate was 100 mm in diameter and formed from a 5 mm thick aluminum plate with a clamping arrangement allowing the location of the plate to be varied in relation to the pile, i.e. the length of pile protruding below the plate can be adjusted (refer to Figure 3b). Prior to centrifuge testing the model foundation system was calibrated for bending in the pile. This involved clamping the bearing plate vertically to a rigid support. The instrumented pile was then loaded as a cantilever beam by the application of discrete weights, and the output from the strain gauges was recorded. Since the strain gauges are located on opposite external faces of the pile their signal can be added to eliminate the effect of axial loading.

2

2.2

2.1

Figure 2. Lateral load versus displacement plots for single gravity experiments with single pile (PB2 – PB6) and piled footing models (HB2-HB6), after Stone et al. 2007.

EXPERIMENTAL PROCEDURES Materials and model preparation

Medium dense sand models were prepared by dry pluviation into 420 mm diameter circular containers. The sand used was a rounded to sub-rounded fine grained, uniformly graded, quartz sand with an average particle size of 0.25 mm (Fraction D from David Ball Ltd.). The maximum and minimum void ratios were determined to be 1.06 and 0.61 respectively, which corresponds to dry unit weights of 12.6 and 16.1 kN/m3 . © 2011 by Taylor & Francis Group, LLC

Centrifuge test procedure

All the centrifuge model tests were conducted on the Acutronic 661 balanced beam centrifuge at the Centre for Geotechnical Modeling, City University, London. Air pluviation created samples with a unit weight of 13.7 kN/m3 , void ratio of 0.89 and a relative density of 36%. On completion of pouring the sand bed, the container was mounted onto the centrifuge platform. Installation of the foundation was then undertaken.The installation method consisted of pushing the pile by hand to about 80% of its desired penetration depth

624

Table 1.

Summary of centrifuge model tests.

Test ID

Test type

Pile Vert load∗ embedment @50 g (mm) (N)

FV PV PL 1 HL1 HL2 HS 1

Footing only vertical push Vertically loaded pile only Laterally loaded pile only Monopiled footing Monopiled footing Monopiled footing

n/a 200 180 180 180 50

∗

n/a n/a 600 600 1100 600

self weight of foundation neglected

Figure 3. (a) Centrifuge model foundation system (b) Photograph of centrifuge model embedded in sand. Figure 4. Vertical load versus vertical displacement for 50 mm diameter plate (FV) and single pile (PV).

and then final driving of the pile by light tapping with a hammer to the desired depth of installation. For the tests involving the combined pile and bearing plate, care was taken to ensure that the bearing plate was in firm contact with the soil surface on completion of installation. For tests when only the single pile response was required, the bearing plate was fixed to the pile shaft some 20–25 mm clear of the soil surface. Lateral loading of the model foundation was provided via a steel wire looped around the pile and connected to a load actuator. Vertical loading was provided through the use of dead weights placed on the bearing plate.A linear variable differential transformer (LVDT) was used to record the lateral displacement at the pile head. A summary of the centrifuge model test programme is presented in Table 1 below. All the centrifuge tests were conducted at 50 gravities.

3

CENTRIFUGE TEST RESULTS

3.1 Vertical loading To obtain basic information on the vertical capacity of the pile and the bearing capacity of the plate, vertical loading tests were conducted. Due to the limitations © 2011 by Taylor & Francis Group, LLC

of the actuation system, to estimate the ultimate bearing capacity of the foundation plate a smaller 50 mm diameter plate was used. Figure 4 shows the results of the two vertical loading tests (refer to Table 1). The ultimate capacity of the bearing plate can be estimated to be between 880 and 1250 N (line FV), depending on the methodology used. These equate to ultimate bearing stresses of 448–637 kPa and back analysis with the bearing capacity equation (Terzaghi, 1943) gives values of the bearing capacity factor Nγ of 44 to 62. These correspond to mobilised friction angles of 36 to 38 degrees for rough foundations (Davis & Booker, 1971; Hansen, 1961). Allowing for increasing capacity with settlements (Ovesen, 1975) and stress level related dilation effects (Kimura et al., 1985), this seems reasonable for this loose material state. Given the same foundation behaviour, the ultimate capacity of the larger 100 mm diameter bearing plate would therefore be in the region of 4480 to 6365 N. From the same plot, it is apparent that the ultimate capacity of the pile is approximately 850 N (line PV). Back analysis using the pile equation (Meyerhof, 1976; Berezantsev et al., 1961) and assuming only skin friction (i.e. no plugging of the open pile), gives a

625

Figure 5. Lateral load versus lateral displacement response for monopiled footing (HL 1) and single pile (PL 1) with a vertical load of 12 N. Figure 6. Pile bending moments versus applied lateral load (black lines for single pile, grey for pile with bearing plate).

mobilised friction angle of 32–33 degrees. This agrees with the observations of Craig and Sabah (1994) who found that mobilised pile friction angles fell between the peak and critical state friction angles.

3.2

Combined loading

The results of the combined vertical and lateral loading tests are best represented through plots of lateral load versus lateral displacement. Figure 5 shows a plot of the lateral load versus lateral displacement for the monopiled footing (HL 1) and single pile (PL 1) with a vertical load of 600 N at 50 g. The horizontal loads were applied at 30 mm above the ground level, giving eccentricities e/L for these foundations of 0.17. It is apparent from this plot that the initial lateral stiffness of the monopiled footing and pile are similar for the first 1–1.5 mm of lateral displacement. However the monopiled footing continues to exhibit a stiffer response than the single pile as the lateral displacement increases. If the lateral capacity of the soil adjacent to the piles is fully mobilised, then solutions such as Zhang et. al’s (2005) or Brom’s (1964) solution for ‘short’ piles in cohesionless soils can be applied. Assuming a friction angle of 32 degrees (with the appropriate L/D ratio) for the two pile systems, suggests that the hybrid system has a behaviour between a fully restrained pile and one with e/L = 0, whereas the pile alone behaves in a manner consistent with a pile foundation horizontally loaded with the appropriate eccentricity. This also suggests a reduction in rotation of the hybrid pile system as it begins to fail, since a fully © 2011 by Taylor & Francis Group, LLC

626

restrained pile would be expected to fail in horizontal translation, rather than rotation. Also presented in Figure 5 is the response for a monopiled footing with a short 50 mm pile embedment depth (HS1). Unfortunately no data are available for the lateral response of the single short pile, although use of Brom’s approach again suggests a behaviour somewhere between a fully restrained pile and one loaded horizontally at the surface (e/L = 0.6). It is also apparent from the shorter pile that the embedment depth has a significant influence on the lateral response of the monopiled footing and highlights the significant influence that the geometry of the respective foundation elements have on the response of the foundation system.

3.3

Bending moments

Figure 6 presents the measured bending moments observed for test HL1 (pile with bearing plate) and test PL1 (pile only) against the applied lateral load. From these data it is apparent that the pile bending moments increase more rapidly in the pile only case, and that the flattening out of the moments is consistent with a rotational failure mechanism of the pile as the lateral resistance of the soil is exceeded. A less rapid increase in pile bending moments is observed with the bearing plate present. It is also apparent that the presence of the bearing plate leads to the generation of higher moments in the pile as the plate rotates into the soil surface. The rigid connection between the pile and the bearing plate results in the transfer of the moments,

developed by the rotating plate, into the pile, such that even as the ultimate lateral resistance of the foundation is being reached, bending in the pile is still increasing.

4

DISCUSSION AND CONCLUSIONS

The presence of the foundation plate rigidly attached to a monopile provides a degree of moment restraint to the pile head through the soil reaction acting on the underside of the bearing plate. In order for this soil reaction to be generated it is necessary that the plate rotates into the soil surface. Consequently it is likely that the very initial response of the monopiled footing remains governed by the lateral stiffness of the pile, but as the foundation system rotates the lateral stiffness of the monopiled footing is influenced by the moment restraint provided by the bearing plate. The efficiency of the bearing plate to stiffen the response of the monopiled footing will be greatly influenced by the initial contact between the bearing plate and the soil, and any vertical pre-stress acting under the plate. In the tests conducted here it is evident that the axial capacity of the pile was significantly greater that the applied vertical load acting on the foundation. Consequently there would not be any initial vertical pre-stress acting on the underside of the bearing plate. Also any gap between the soil and the plate would eliminate any influence of the bearing plate on the initial lateral response. Nevertheless, the small number of tests conducted did provide some valuable information from which the following conclusions can be drawn: • It is apparent that the lateral response of a single

monopiled foundation can be enhanced by the presence of a bearing plate resulting in a greater ultimate lateral capacity. • Whilst the effect on the initial lateral stiffness was not observed to be significant in the tests reported here, the lateral stiffness beyond this initial movement was significantly enhanced through the presence of the bearing plate. • Further research is required to explore the influence of the geometry of the respective foundation elements, and the response of monopiled footing foundations in different soils. REFERENCES Berezantsev et al. (1961). Load bearing capacity and deformation of piled foundations. Proc. 5th Int Conf Soil Mechanics and Foundation Engineering, Paris, 2, 11–12. Bransby, M. F. and Randolph, M. F. (1998). Combined loading of skirted foundations. Géotechnique, 48(5), 637–655. Brinch Hansen, J. (1961). The ultimate resistance of rigid piles against transversal forces. The Danish Geotechnical Institute Bulletin 12, 5–9. Broms, B.B. (1964). Lateral resistance of piles in cohesionless soils. ASCE Journal of the Soil Mechanics and Foundation Division Proceedings (JSMFD), 90(SM3), 123–156.

© 2011 by Taylor & Francis Group, LLC

Craig, W.H. and Sabagh, S.K. (1994). Stress-level effects in model tests on piles. Canadian geotechnical journal. 31(1), 28–41. Davis, E.H. and Booker, J.R. (1971). The bearing capacity of strip footings from the standpoint of plasticity theory. Proc. 1st Australia-New Zealand Conf. on Geomech., Melbourne (Australia), 276–282. Duncan, J. M., Evans, L. T. and Ooi, P. S. (1994). Lateral load analysis of single piles and drilled shafts. ASCE Journal of Geotechnical Engineering, 120(6), 1018–1033. El-Marassi, M., Newson, T., El-Naggar, H. and Stone, K. (2008). Numerical modelling of the performance of a hybrid monopiled-footing foundation. GeoEdmonton2008: Proceedings of the Canadian Geotechnical Conference. (Paper # No. 420, 1–8). Gourvenec, S. and Randolph, M. (2003). Effect of strength non-homogeneity on the shape of failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique, 53(6), 575–586. Houlsby, G. T. and Puzrin, A. M. (1999). The bearing capacity of a strip footing on clay under combined loading. Proc. R. Soc. London Ser. A. 455, 893–916. Kim, J. B., Singh, L. P. and Brungraber, R. J. (1979). Pile cap soil interaction from full scale lateral load tests. ASCE Journal of Geotechnical Engineering, 105(5), 643–653. Kimura, T., Kusakabe, O. and Saitoh, K. (1985). Geotechnical ModelTests of Bearing Capacity Problems in a Centrifuge. Géotechnique, 35(1), 33–45. Maharaj, D.K. (2003). Load-Deflection Response of Laterally Loaded Single Pile by Nonlinear Finite Element Analysis. EJEG. Matlock, H. and Reese, L. C. (1960). Generalized solutions for laterally loaded piles. ASCE Journal of Soil Mechanics and Foundations Division, 86(SM5), 63–91. Meyerhof, G.G. (1976). Bearing capacity and settlement of pile foundations. ASCE, Journal of Geotechnical Engineering. 102(GT3), 197–228. Mokwa, R.L. (1999). Investigation of the Resistance of Pile Caps to Lateral Loading. Ph.D Thesis, Virginia Polytechnic Institute, Blacksburg, Virginia. Mokwa, R.L. and Duncan, J.M. (2001). Experimental evaluation of lateral-load resistance of pile caps, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 127(2),: 185–192. Mokwa, R.L. and Duncan, J.M. (2003). Rotational restraint of pile caps during lateral loading, Journal of Geotechnical and Geoenvironmental Engineering,ASCE, 129(9),: 829– 837. Ovesen, N.K. (1975). Centrifugal testing applied to bearing capacity problems of footings on sand. Geotechnique, (25)2, 394–401. Poulos, H.G. and Randolph, M.F. (1983.) Pile group analysis: a study of two methods, Jour. of Geot. Eng., ASCE, 109(3), 355–372. Poulos, H. G. (1971). Behaviour of laterally loaded piles: Part I-single piles. ASCE Journal of the Soil Mechanics and Foundations Division, 97(SM5), 711–731. Poulos, H.G. (1988). Marine Geotechnics. Unwin Hyman, London. Powrie, W. and Daly, M.P. (2007). Centrifuge modeling of embedded retaining wall with stabilizing bases. Geotechnique, 57(6), 485–497. Randolph, M. F. (1981). The response of flexible piles to lateral loading. Géotechnique, 31(2), 247–259. Reese, L. C., Cox, W. R. and Koop, F. D. (1974). Analysis of laterally loaded piles in sand. Offshore Technology Conference, Vol. II, Paper No. 2080, Houston, Texas, 473–484.

627

Stone K.J.L., and Newson, T.A. and El Marassi, M. (2010). An investigation of a monopiled-footing foundation. Int. Conf. Phys. Modelling in Geotech., ICPMG2010 (Zurich), Balkema, Rotterdam. Stone K.J.L., Newson, T.A. and Sandon, J. (2007). An investigation of the performance of a ‘hybrid’ monopole-footing foundation for offshore structures. Proc. 6th Int. Offshore Site Investigation Conf., 11–13th Sept 2007, London, UK.

© 2011 by Taylor & Francis Group, LLC

Terzaghi, K. (1943). Theoretical Soil Mechanics. John Wiley, New York. Zhang, L., Silva, F. and R Grismala, R. (2005). Ultimate Lateral Resistance to Piles in Cohesionless Soils. Journal of Geotechnical and Geoenvironmental Engineering, Vol. 131(1), 78–83.

628

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Optimizing site investigations and pile design for wind farms using geostatistical methods: A case study B. Stuyts, V. Vissers, D.N. Cathie & C. Jaeck Cathie Associates SA/NV, Belgium

S. Dörfeldt Offshore Wind Technologie GmbH, Germany

ABSTRACT: Offshore wind farms are currently being rapidly developed in the North Sea as alternative energy resources. In contrast to offshore oil and gas developments, wind farms consist of a large number of relatively closely spaced foundations. In order to comply with national and European standards, certifying authorities often require a borehole at each wind turbine location. This results in large site investigations and takes no account of geotechnical knowledge gained from nearby boreholes. Moreover, the overall achieved reliability for foundation design may be affected by soil variability. This can also be captured probabilistically using a geostatistical model of the site data. A case study of a North Sea windfarm is presented and the feasibility of using geostatistics to capture the structural variability of the site is demonstrated. Overall project decision making is incorporated into a simple foundation cost-reliability model which could become a component of a probabilistic project value model.

1

INTRODUCTION

The increasing demand for renewable energy resources is driving the rapid development of offshore wind farms in Western Europe. These wind farms typically consist of more than 40 turbines which are supported with piled tripods, jackets or gravity base foundations. Due to the large number of relatively closely spaced foundations, a very extensive site investigation program is often required by certifying bodies or national authorities. Such programs often require one or more boreholes at each turbine location. Drilling boreholes at every turbine location represents a significant cost before turbine construction and the project schedule often depends strongly on the availability of appropriate geotechnical investigation vessels. In this paper, a case study of a windfarm site in the North Sea consisting of 80 turbines is presented. The wind turbines are supported by piled foundations and the axial capacity in compression is the limiting factor for foundation design. One borehole with combined CPT and sampling was drilled at every turbine location to obtain the necessary geotechnical parameters for the ICP pile design method (Jardine et al. 2005). The data gathered during the geotechnical campaign was used to assess the feasibility of using a geostatistical model during the planning of the site investigation to optimize the number of required boreholes. In this work, the variability of pile capacity across the windfarm site is characterized both with conventional statistical analysis and geostatistical methods. The variability models are then used to estimate pile © 2011 by Taylor & Francis Group, LLC

capacity and associated variance at uninvestigated turbine locations through the technique of conditional simulation (Chilès & Delfiner 1999). First order reliability methods (Ditlevsen & Madsen 2007) can then be used to assess the probability of foundation failure at both investigated and uninvestigated locations. A simple cost-benefit analysis using the net present value (NPV) of the foundation construction project is used to determine whether the site investigation campaign can be optimized to minimize the overall project cost. 2

STATISTICAL AND GEOSTATISTICAL VARIABILITY ACROSS THE SITE

Geological information on the windfarm site under consideration shows that the soils are mainly glacial of origin, are relatively uniform, and consist of dense to very dense sands. Paleochannels which are infilled with various similar sediments cut across the site. A geophysical campaign also revealed these as reflectors and the channel features were mapped (Fig. 1). The ensuing geotechnical campaign showed that these paleochannels were not necessarily correlated with the presence of clayey or silty layers, but an increasing variability in pile capacity was observed inside and close to the mapped channel features. Due to the different levels of variability in- and outside the paleochannels, two sets of data were considered (Fig. 1) and statistical and geostatistical data analysis was carried out for both data sets. The results of the conventional statistical analysis are shown in

629

Figure 2. Normal distribution fit to axial pile capacity at 35 m.

Figure 1. Site map showing paleochannels cutting across the site. Table 1.

Statistical analysis results.

Axial pile capacity [MN] Pile length [m] Mean Minimum Maximum Standard deviation Coefficient of variation

Locations outside channels (59 locations)

Channel locations (24 locations) 25 m

30 m

35 m

25 m

30 m

35 m

44.0 16.3 82.7 18.8

56.3 17.5 96.4 22.6

64.2 21.1 107 23.5

38.8 12.9 65.9 11.1

52.6 20.1 77.0 12.1

68.1 37.1 91.9 11.7

0.43

0.40

0.37

0.29

0.23

Figure 3. Variogram for pile capacity at 30 m for boreholes outside channels (Number of data couples for each separation distance are given as labels).

0.17

Table 1. Three different pile lengths were considered 25 m, 30 m and 35 m. The high coefficient of variation of the boreholes inside the channel features shows that boreholes at every location would be warranted in this part of the site. Therefore, these locations were not considered in the remainder of the geostatistical analysis. For the remaining 59 turbine locations, normal distributions could be fitted to the pile capacities as shown in Figure 2. Structural variability of a geotechnical variable can be captured using the experimental variogram (Equation 1). For each separation distance, the variance of the couples of data points with this separation is computed:

 is the value data points for this spacing and γ ◦ (h) of the experimental variogram for the spacing under consideration. This procedure is repeated for each separation distance and an experimental variogram as shown in Figure 3 is typically obtained. For small distances, the variance is small or zero and as the separation between data points increases, so does the variance. Eventually, at a certain separation distance called the range, ha , the variance levels off and becomes equal to the population variance, which is called the sill. Experimental variograms for the boreholes outside the channels showed this structure whereas the variograms for the boreholes inside the channels were much more erratic.

where F(x) is the value of the geotechnical variable at point x , h is the spacing, N is the number of © 2011 by Taylor & Francis Group, LLC

630

3

GEOSTATISTICAL VARIABILITY MODELS AND GEOSTATISTICAL SIMULATION

Based on a given experimental variogram, a fitted variogram model can be obtained as shown in Figure 3. This variogram model is required in the further geostatistical analyses. For the axial pile capacities for the North Sea windfarm site, a Gaussian model (Equation 2) was fitted to the experimental variograms of the boreholes outside the channels.

The boreholes inside and close to the channels were not considered further in the geostatistical analysis.

The sill of the fitted Gaussian model was chosen equal to the population variance and the range was approximately 1000 m for both 30 m and 35 m long piles. This indicates that the pile capacity is quite variable across the site and that the correlation with boreholes further than 1000 m away will be limited. Geostatistical methods have been used in extensively in mining (Krige 1951), hydrology (Chilès & Delhomme 1975), and petroleum engineering (Chilès & Delfiner 1999) to estimate the value of geotechnical parameters at locations where measurements are unavailable. Kriging is a geostatistical technique which estimates the geotechnical variable as a weighted average of the values at surrounding locations. The weights are calculated by minimizing the variance on the estimate through the distance to available data points and the variogram model of the data. The main advantage of kriging is that the minimized variance on the estimate is obtained together with the estimate itself. More recently, the technique of conditional simulation (Alfaro 1979) has been used with the advantage that the simulated datapoints obey both the statistical distribution of the data and the variogram. The technique builds on kriging to condition the data to obey the spatial structure. Conditional simulation is usually preferred over kriging because it reflects patterns of local variability more satisfactorily (de Smith et al. 2006). Conditional simulations of the pile capacity at 30 m and 35 m have been carried out using the geostatistical software WinGSlib (Statios 2007). Four different site investigation scenarios were tested in which the number of drilled boreholes was increased progressively (15, 30 and 45 boreholes) until the site was eventually covered (57 boreholes) with one borehole per location. An example of the simulated pile capacity at 30 m for 30 available boreholes is shown in Figure 4. As more and more boreholes are drilled, the estimation variance at unsampled locations reduces. However, the simulated contour maps show that the estimation variance increases quickly with increasing distance from the sampled locations. 4

RELIABILITY CALCULATION

Figure 4. Simulated expected value (a) and standard deviation (b) of 30 m long piles for 30 available boreholes. (Investigated locations marked in white, uninvestigated locations in black).

reliability index of a piled foundation in axial compression can be obtained using the first order second moment method (FOSM) (Ditlevsen & Madsen 2007) if the mean and standard deviation of the load (S) and the resistance (R) are known (Equation 3).

At locations where geotechnical data is not available, the standard deviation of the estimate, σe , also needs to be taken into account. In the case study, the standard deviation from the estimation process is added to the standard deviation accounted for in the engineering codes of practice. The reliability index can then be calculated using the variables from Equation 4.

Partial safetly factors quoted in most engineering codes of practice for piled structures are calibrated to obtain specific reliability levels. For wind turbines, the target reliability index β is typically 4.1 corresponding to an annual probability of failure pF = 2.0e-5. (Vrouwenvelder 2003). The extreme design load and the soil resistance each have their specific probability distribution, depending on the wind, waves, currents and soil conditions. The © 2011 by Taylor & Francis Group, LLC

631

Figure 5 shows that the distribution of the soil resistance becomes wider as the standard deviation on the estimate increases. This results in more overlap between the distributions of load and resistance and thus a lower reliability index and a higher probability of failure.

Figure 5. Influence of standard deviation on estimate of reliability.

Figure 7. Contribution of different cost drivers to the overall project cost. • Pile installation costs, depending on pile length,

CI (L) The site investigation cost will have a significant impact on the other two cost components through the pile length. If geotechnical data is available, the pile length can often be reduced whereas a longer pile would have to be used at uninvestigated locations to account for the additional uncertainty. The contribution of each of the cost components to the total project cost is shown in Figure 7. Figure 7 shows that pile fabrication is the major cost driver for the wind turbine foundations. As more and more locations are investigated, the required pile length can be reduced at most locations. This results in cost savings on the fabrication and installation of the piles. Geostatistical simulations have the benefit that the risk of pile failure can be quantified in monetary terms. Even though the cost of a wind turbine foundation, CF , is extremely difficult to assess, the risk to investors can be calculated by multiplying CF by the probability of failure. The cost of failure contains not only the cost of removal of the failed foundation and replacement of the turbine sub-structure, but also other intangible costs such as the damage to the image of the windfarm developer or other parties involved, and litigation costs. For this case study, a failure cost of 10 times the foundation installation cost has been assumed but this could easily be higher. In investment analyses, all costs are related to the present through the net present value concept. The net present cost (NPC) of a windfarm foundation project is written in Equation (6) where r is the interest rate and T is the total windfarm lifespan.

Figure 6. Map of reliability index for 35 m long piles with 45 investigated locations available.

The results of the condition simulations were used to calculate the reliability index and corresponding probability of failure at each turbine location. Assuming load and resistance are statistically independent and the failure equation can be written as R = S, the reliability index using form is given in Equation 5.

Figure 6 shows the map of the reliability index for 35 m piles with 45 investigated locations available. The results clearly show that the target reliability index of β = 4.1 is reached at the investigated locations only. In between the investigated locations, the variability is too high to reach the target at any of the uninvestigated locations. 5

COST-RELIABILITY MODEL

Investors need to perform a risk-benefit analysis before taking investment decisions. For the wind farm case study, the main cost drivers for the pile foundations are: • Site investigation costs, CSI • Pile fabrication costs, depending on pile length,

CPF (L) © 2011 by Taylor & Francis Group, LLC

The NPC can be calculated for each of the four site investigation scenarios given in Section 3. Because of the calibration range of the ICP method, 30 m was taken as the minimum pile length. At locations where site investigation was not performed, a 35 m pile was

632

•

•

Figure 8. NPC of windfarm foundations vs number of investigated locations.

assumed to be needed in order to demonstrate the NPC model. However, the geostatistical simulations showed that the required reliability index was not achieved for any uninvestigated location, so a significant failure risk would still exist for this length pile. In order to demonstrate the NPC model, the site investigation cost was assessed as a cost per borehole and the pile fabrication and installation as a cost per meter pile length. Typical values for these cost drivers known at the time of writing were used to assess the NPC for the wind farm case study. The NPC model is a simplified cost model and includes only the cost components shown in Equation 6. Other costs components such as mobilization costs, pile transportation costs, etc. are not included but could easily be added to this formula if their values are known. Figure 8 shows the decrease of the net present cost as more and more locations are investigated. The results clearly show how site investigation can dramatically decrease the failure risk and thereby reduce the net present cost of the foundation project by more than 40%. Drilling a geotechnical borehole at every turbine location is clearly warranted, and necessary. 6

CONCLUSIONS

Site investigation can represent a major cost during the development phase of offshore wind energy projects. In this paper, the possibility of using geostatistical methods for characterizing the spatial variability of axial pile capacity across a windfarm site has been investigated. The results of geostatistical simulations have been combined with first order reliability methods to get an estimate of the net present cost of a windfarm foundation project. The following conclusions can be drawn from the analyses: • Spatial variability across the windfarm site is quite

large with limited correlation existing between neighboring boreholes for typical spacings of about

© 2011 by Taylor & Francis Group, LLC

•

•

•

•

1 km at the site studied. The heterogeneity of the glacial deposits results in a rapid increase in variance with distance from a borehole; Geostatistical simulations can be used to obtain an estimate which obeys both the spatial structure and the distribution of the site data. At uninvestigated turbine locations, the variance on the estimate is obtained together with the estimate itself; The reliability index required by the engineering codes of practice for 35 m long piles cannot be obtained for any of the wind turbines when a borehole is not drilled at each location. Longer piles would be required to achieve the required reliability index, resulting in heavily overdesigned foundations; The risk of failure at uninvestigated locations is too high to lead to cost savings by reducing the size of the site investigation. The net present cost is minimum when a borehole is drilled at each turbine location; The combination of geostatistical and probabilistic methods provides the possibility to quantitatively assess the foundation failure costs and incorporate these into a project value model; For new developments, the methodology in this work could be used to assess whether every foundation location needs to be investigated and to compute the NPC for different site investigation strategies; A foundation cost-risk model such as this could be incorporated into a full financial risk model for the overall development.

REFERENCES Alfaro, M. (1979). Etude de la robustesse des simulations de fonctions aléatoires. Doctoral thesis, E.N.S. des Mines de Paris. Chilès, J.P. & Delfiner P. (1999), Geostatistics: Modeling Spatial Uncertainty, Wiley series in probability and statistics, Book, ISBN 0 471 08315 3, pp. 695. Delfiner, P. & Delhomme, J.P. (1975). Optimum interpolation by kriging. Display and Analysis of Spatial Data, J.C. Davis and M.J. McCullagh, eds. Wiley, London, 96–114. de Smith, Goodchild, Longley (2006). Geospatial analysis, a comprehensive guide. 3rd edition. Ditlevsen, O. & Madsen, H.O. (2007). Structural Reliability Methods. Technical University of Denmark. JuneSeptember 2007. Jardine, R., Chow, F., Overy, R. & Standing, J. (2005), ICP Design Methods for Driven Piles in Sands and Clays, Thomas Telford, Book, ISBN 0 7277 3272 2, pp. 105. Krige, D.G. (1951). A statistical approach to some basic mine valuation problems on the Witwatersrand. Journal of the Chemical Metallurgical and Mining Society of South Africa, December 119–139. Statios (2007). WinGsLib. Geostatistical software libraries. Vrouwenvelder, T. (2003). Reliability Based Code Calibration – The Use of the JSCC Probabilistic Model Code. Joint Committee of Structural Safety, Workshop on Code Calibration, March 21/22, Zurich.

633

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Towards the FE prediction of permanent deformations of offshore wind power plant foundations using a high-cycle accumulation model T. Wichtmann, A. Niemunis & Th. Triantafyllidis Institute of Soil Mechanics and Rock Mechanics, Karlsruhe Institute of Technology, Germany

ABSTRACT: The paper discusses a possible application of the authors’ high-cycle accumulation (HCA) model for the prediction of long-term deformations of offshore wind power plant (OWPP) foundations. The calibration of the HCA model parameters for a typical North Sea fine sand is presented. These parameters have been used for exemplary finite element calculations of a monopile foundation, with variation of soil density, average load and cyclic load amplitude.

1

2

INTRODUCTION

Numerous offshore wind parks will be installed in the North Sea and in the Baltic Sea during the next years. The foundations of offshore wind power plants (OWPPs) are subjected to a multiaxial high-cyclic loading due to wind and waves, which may cause an accumulation of permanent deformations. The serviceability of the OWPPs may get lost due to an excessive tilting of the tower. No established methods for a prediction of the long-term deformations exist so far. Experience from conventional offshore foundations (e.g. oil rigs) cannot be easily adapted since the ratio of the horizontal cyclic load and the own weight of the structure is significantly larger for the new OWPPs. While existing offshore wind parks lay in relatively shallow water, the water depths will be up to 40 m at the locations of the new offshore wind parks. Therefore, the bending moments at the seabed level will be much larger for the new OWPPs. Furthermore, the new OWPPs will have larger dimensions than the existing ones due to increased power generation requirements. For example, monopiles with diameters between 5 and 8 m will be installed. Practical experience with such large pile diameters in combination with high ratios of cyclic horizontal load and self weight of the structure is missing. Existing methods for the prediction of long-term deformations were developed and proven for much smaller pile diameters and for lower cyclic loads. The applicability of these methods to the new OWPPs is questionable (Section 2). The present paper discusses a possible application of the authors’ high-cycle accumulation (HCA) model (Niemunis et al., 2005) for finite element (FE) predictions of the permanent deformations of OWPP foundations. © 2011 by Taylor & Francis Group, LLC

LITERATURE REVIEW

The guidelines for the design of OWPP foundations published by the certifier Germanischer Lloyd (2005) demand an investigation of both, the short-term and the long-term soil-structure interaction under cyclic loading. However, the methods and the extent of such investigations are not further specified. It is stated that neither a theory nor established investigation methods exist with respect to the long-term behaviour. It is recommended to utilize experience from similar projects in the past, which are missing in the case of the new OWPPs. As a possible approach the adherence of given maximum deflections under a static equivalent load is mentioned but judged as possibly non-conservative. Both, the recommendations of the American Petroleum Institute (1993) and the Offshore Standard DNV-OS-J101 (Det Norske Veritas, 2004) utilize p-ycurves in order to predict lateral deformations of piles subjected to horizontal loads. In the case of cyclic loading p is reduced. Based on tests of Reese et al. (1974), the API standard proposes a constant reduction factor 0.9. Since the load amplitude, the number of cycles and the soil conditions are not considered, such reduction seems to oversimplify the problem according to the authors’ opinion. The DNV standard recommends to use “suitable” p-reduction factors, which are not further specified. A calculation of the cumulative deformations in the soil in a “suitable manner” is demanded by DNV, but no respective method has been specified or recommended. Some more sophisticated modifications of p-ycurves considering the influence of a cyclic loading have been proposed in the literature (Welch and Reese, 1972; Swinianski and Sawicki, 1991; Long and Vanneste, 1994; Little and Briaud, 1988). For example,

635

Little and Briaud (1988) proposed to increase y with N according to y(N ) = y(N = 1) N a , where the parameter a can be obtained from cyclic pressuremeter tests. However, an extrapolation of the pressuremeter data obtained for small numbers of cycles to large N -values seems doubtful. It should be kept in mind that all these methods and equations recommended in the standard codes or proposed in the literature were developed for smaller pile diameters, lower ratios of cyclic horizontal load and self weight of the structure, and smaller number of cycles. The applicability to the large pile diameters of the new OWPPs in combination with the large cyclic horizontal loading has not been confirmed yet. This lack of knowledge served as a motivation for ongoing research on the long-term behaviour of the new OWPP foundations. Different approaches are followed. Achmus et al. (2008) calculated the lateral displacements with an elastoplastic constitutive model, reducing the constrained elastic modulus M in dependence of the number of load cycles and the load amplitude. Lesny and Hinz (2006) described a method in which the permanent deformations are predicted using a strain wedge model. The stress-strain behaviour of the soil is derived from a multi-stage cyclic triaxial test. Based on model tests and FE calculations, Dührkop (2010) proposed a p-reduction factor which increases linearly from zero at the seabed level to 0.9 for z/L > 0:5 with L being the pile length. Since in-situ measurements are not available yet for the new OWPP foundations, all these methods were calibrated or proven based on small-scale model tests or measurements at smaller pile diameters. Therefore, at present it is unclear if these procedures are able to predict realistic permanent deformations. The authors of the present paper intend to predict the permanent deformations of OWPP foundations by means of finite element calculations using their highcycle accumulation model. 3

HIGH CYCLE ACCUMULATION MODEL

The main constitutive equation of the HCA model reads

with the Jaumann stress rate σ˙ of the effective Cauchy stress σ, the strain rate ε˙ , the prescribed strain accumulation rate ε˙ acc , the plastic strain rate ε˙ pl (for stress paths touching the yield surface only) and the pressuredependent elastic stiffness E. In the highcyclic context “rate” means the derivative with respect to the number of cycles N . The accumulation rate is calculated as the product of the scalar intensity of accumulation ε˙ acc and of the direction of accumulation m (a unit tensor):

A multiplicative approach is used for ε˙ acc . Each function considers separately the influence of a different parameter (Table 1, fampl : strain amplitude, fe : void ratio, fp : average mean pressure pav (that means the average value of mean pressure p during a cycle), fY : average stress ratio ηav = qav /pav , f˙N : cyclic preloading, fπ : polarization changes). The spatial field of the strain amplitude can be obtained from a calculation of a few cycles using a conventional constitutive model. The authors use hypoplasticity with intergranular strain (vonWolffersdorff, 1996; Niemunis and Herle, 1997) for that purpose. 4

The HCA model parameters have been determined for a uniform fine sand (d50 = 0.14 mm, Cu = d60 = d10 = 1:5). This sand is also currently used in small-scale model tests on OWPP foundations performed at our institute. It is intended to verify the accuracy of the HCA model prediction by recalculations of these model tests. Stress-controlled drained cyclic triaxial tests with 105 load cycles applied at a frequency of 0.2 Hz have been performed in order to determine the parameters Campl , Ce , Cp , CY , CN 1 , CN 2 and CN 3 (Table 1). Four different amplitudes, seven different initial relative densities ID0 , four different average mean pressures pav and four different average stress ratios ηav = qav /pav have been tested in separate tests. No multistage tests have been performed. Figure 1 shows a typical plot of the vertical strain ε1 (t) measured during the first 24 cycles and during five cycles recorded at N = 50, 100, 200, …, 105 . The upper row of diagrams in Figure 2 shows the increase of the residual (= permanent, plastic) strain εacc with increasing number of cycles N measured in the four test series. Evidently, the rate of strain accumulation increases with increasing amplitude (Figure 2a), decreasing density (Figure 2c) and increasing average stress ratio (Figure 2g). Similar residual strains were obtained for different average mean pressures (Figure 2e) because the tests were performed with the same amplitude-pressure ratio ζ = qampl /pav = 0.3. Table 1. HCA model functions and parameters for the tested fine sand (emin = 0.677, emax = 1.054).

The flow rule of the modified Cam clay (MCC) model has been experimentally found to approximate m well. © 2011 by Taylor & Francis Group, LLC

CALIBRATION OF THE HCA MODEL FOR A FINE SAND

636

Function fampl = (εampl /10−4 )Campl 2 max fe = (C1e +− ee) (C1e+−eemax )2 av fp = exp[−Cp (p /100 − 1)] fY = exp(CY Y¯ av ) f˙N = f˙NA + f˙NB   A ˙f A = CN 1 CN 2 exp − g N CN 1fampl ˙f B = CN 1 CN 3 N

Mat. const.

Value

Campl Ce Cp CY CN 1

1.31 0.58 0.22 1.85 2.82·10−4

CN 2

0.37

CN 3

2.64·10−5

The HCA model parameter Campl was determined from a curve-fitting of the function fampl (Table 1) to the data shown in Figure 2b. In that figure the residual strain εacc after different numbers of cycles is plotted versus a mean value of the strain amplitude, calculated as ε¯ ampl = 1/N εampl (N )dN . This averaging is necessary since the tests have been performed stresscontrolled and thus the strain amplitude decreases slightly with N (especially during the first 100 cycles). On the ordinate the residual strain has been divided by the void ratio function f¯e of the HCA model in order to purify the data from the influence of slightly different initial densities and different compaction rates. f¯e has been  calculated with a mean value of void ratio e¯ = 1/N e(N )dN . The parameter Campl given in Table 1 is the average of the values determined for different numbers of cycles. A curve-fitting of the function fe to the data in Figure 2d delivered the parameter Ce given in Table 1. In Figure 2d the residual strain has been divided by

Figure 1. Vertical strain ε1 (t) measured during the initial phase of a drained cyclic test and after different numbers of cycles.

the amplitude function fampl in order to purify the data from the influence of slightly different strain amplitudes. The data are plotted versus a mean value of void ratio. The parameters Cp and CY (Table 1) were determined from a curve-fitting of the functions fp and fY to the data in Figures 2f and 2h. In those diagrams the residual strain has been divided by the amplitude and void ratio function and plotted versus pav or Y¯ av , respectively, where Y¯ av is a normalized stress ratio which is zero for isotropic stresses and 1 on the critical state line. The curves εacc (N ) from Figure 2 have been divided by the functions f¯ampl , f¯e , fp and fY of the HCA model (Figure 3) in order to determine the parameters CN 1 , CN 2 and CN 3 . A curve-fitting of the function fN = CN 1 · [ln(1 + CN 2 N ) + CN 3 N ] to the data in Figure 3 (solid curve) delivered the CNi -values specified in Table 1. The critical friction angle ϕc = 33:1◦ necessary for the cyclic flow rule m has been determined from the inclination of a pluviated cone of dry sand. Isotropic elasticity is assumed for E in Eq. (1). Therefore, two elastic constants (e.g. bulk modulus K and Poisson’s ratio ν) have to be determined. The bulk modulus K = u˙ /˙εacc v can be obtained as the ratio of the rate of pore water pressure accumulation u˙ in a stress-controlled undrained cyclic test and the rate of volumetric strain accumulation ε˙ acc v measured in a drained test. Both samples should have similar initial densities and the tests should be performed with identical consolidation stresses and cyclic loads. Six such test pairs have been performed on the fine sand so far. All specimens were prepared medium dense and consolidated isotropically. Different initial effective mean pressures in the range 50 kPa ≤ p0 ≤ 300 kPa and different amplitude-pressure ratios in the range 0.2 ≤ ζ = qampl /p0 ≤ 0.3 were tested (not all combinations have been tested so far). Based on the test results (documented in detail byWichtmann et al. (2010)) the pressure-dependent bulk modulus can be described by

Figure 2. Results of drained cyclic tests with different a,b) amplitudes, c,d) initial relative densities ID0 , e,f) average mean pressures pav and g,h) average stress ratios ηav .

© 2011 by Taylor & Francis Group, LLC

637

Figure 3. Determination of parameters CNi from a curve-fitting of the function fN to the curves εacc (N )/(f¯ampl f¯e f¯p f¯Y ). Table 2.

Hypoplastic parameters for the fine sand.

hs [MPa]

n [–]

α [–]

β [–]

ei0 [–]

ec0 [–]

ed0 [–]

862.6

0.32

0.21

1.5

1.212

1.054

0.677

Figure 4. FE model of a OWPP monopile foundation.

pav = 200 kPa, ηav = 0.75 and qampl = 60 kPa is reproduced well. Table 3. R [–] −4

10

Parameters of intergranular strain for the fine sand.

5

mT [–]

mR [–]

βR [–]

χ [–]

2.3

4.6

0.2

2.8

with patm = 100 kPa and with constants A = 467 and n = 0.46. No significant influence of the amplitude on K could be detected. Poisson’s ratio ν can be obtained from the shape of the average effective stress path in an undrained test with anisotropic consolidation stresses and strain cycles. A comprehensive test series on the fine sand is documented elsewhere (Wichtmann et al., 2010). A Poisson’s ratio of ν ≈ 0.32 has been found appropriate for calculations with the HCA model. The parameters of the hypoplastic constitutive model with intergranular strain are needed in the FE calculations for the determination of the spatial field of the strain amplitude. The parameters of the conventional hypoplastic model (Table 2) were determined according to Herle (1997). Drained monotonic triaxial tests on dense samples and oedometric compression tests on loose and dense samples were performed for that purpose. The parameters of intergranular strain (Table 3) were chosen such way that the strain amplitude measured in a drained cyclic triaxial test with © 2011 by Taylor & Francis Group, LLC

FE CALCULATIONS

The FE calculations have been performed in order to prove the HCA model prediction qualitatively. The geometry of a real OWPP project (Figure 4) supplied by “Germanischer LloydWind Energie GmbH” has been used. The inner and outer diameter of the monopile are di = 5.00 m and da = 5.09 m, respectively. The depth of embedding is 32.65 m. An erosion of the upper 3 m of the soil is likely to occur, thus only the lower 29.65 m were considered for the embedding. The commercial program Abaqus 6.7 was used in combination with the user-subroutine Umat, in which both the hypoplastic model with intergranular strain and the HCA model are implemented. The 3D-FE discretisation is shown in Figure 4. Since only an unidirectional cyclic loading has been considered so far, the symmetry of the system could be utilized and only one half of the problem’s geometry was modeled. Three-dimensional elements with reduced integration scheme (C3D8R) were used for the discretisation. Ten layers of elements were chosen along the depth of embedding and two elements were used for the thickness of the pipe.The soil was modeled within the radius of 20 m around the pile shaft and up to a depth of 20 m below the pile tip. The pile was modeled up to 1 m above the seabed. The Young’s modulus of the steel pipe was E = 2.1 · 108 kPa except the upper 1 m where it was chosen as E = 1010 kPa in order to distribute the concentrated loads applied to the head of the pile. A Mohr-Coulomb contact with a friction coefficient µ = 0.3 was used at the inner and outer side of the

638

Figure 5. Results of FE calculations: Horizontal displacement of the monopile as a function of depth z below seabed, calculated for different amplitudes and average values of the bending moment.

steel pipe. The bending momentM was applied as a pair of vertical forces acting on two nodes laying on the symmetry axis (Figure 4). The shear force Q was also equally distributed to these two nodes. The vertical load V due to the own weight of the OWPP tower was distributed equally to all nodes at the top of the pile. A uniform distribution of the initial void ratio was assumed. The initial earth pressure coefficient at rest was set to K0 = 0.5. The loading was applied in the following steps: 1. Application of the self weight of the soil with the geostatic initial stress (without generating deformations). 2. Application of the vertical force V = 9247 kN representing the own weight of the tower. 3. Application of the average values of the bending moment (M av ) and the shear force (Q av ). 4. Calculation of the first cycle, using the hypoplastic model with intergranular strain. The loading was applied sinusoidal with the amplitudes M ampl and Qampl . 5. Calculation of the second cycle, using the hypoplastic model with intergranular strain. During the cycle, the strain path was recorded in each integration point. The spatial field of the strain amplitude was determined from that strain path and is input for the following calculation with the HCA model. 6. Calculation of permanent deformations due to N = 106 cycles using the HCA model. The bending moment and the shear force were kept constant at their average values M av and Q av while the permanent deformations due to the cyclic loading with M ampl and Qampl were predicted by the HCA model. Calculations with different amplitudes M av and average values M ampl of the bending moment were performed. The chosen values cover the design range for real OWPP projects. A constant ratio Q/M = 0.027 1/m of shear force and bending moment was set into approach. Figure 5a presents the lateral pile displacements after N = 106 cycles applied with different amplitudes in the range 10 MNm ≤ M ampl ≤ 25 MNm. Results from calculations with different average values 20 MNm ≤ M av ≤ 50 MNm of the bending moment are shown in © 2011 by Taylor & Francis Group, LLC

Figure 5b, while Figure 5c provides the lateral displacement for different initial relative densities in the range 0.5 ≤ ID0 ≤ 0.9. The predicted increase of the permanent lateral displacements with increasing amplitude M ampl , increasing average value M av and decreasing initial density could be expected from the cyclic triaxial test results (Figure 2b) and is in accordance with model test data (Hettler, 1981; Long and Vanneste, 1994; Lin and Liao, 1999). A quantitative verification of the HCA model prediction based on model tests performed on the fine sand and (if available) in situ data is planned for the future. For a shallow foundation the HCA model prediction has been already proven by recalculations of a centrifuge model test (Niemunis et al., 2005). 6

SUMMARY, CONCLUSIONS AND OUTLOOK

The paper discusses a possible application of the authors’ high-cycle accumulation (HCA) model for the prediction of permanent deformations of OWPP foundations. The calibration of the HCA model for a typical North Sea fine sand is presented. The parameters have been used for finite element calculations of a real OWPP project founded on a monopile. The predicted increase of the permanent lateral displacements with increasing amplitude and average value of the applied bending moment and with decreasing soil density is qualitatively plausible and in accordance with model test results in the literature. A quantitative verification of the HCA model prediction is planned for the near future. Several series of cyclic triaxial tests will be performed on the fine sand in order to further develop the HCA model, in particular with regard to the application to OWPP foundations. Since the OWPP foundations are subjected to a very large number of load cycles, long-term tests with N ≈ 108 cycles are planned in order to evaluate the function fN of the HCA model for large N -values. The effect of changes of the polarization of the cycles (factor fπ of the HCA model) will be studied in cyclic triaxial tests with a simultaneous oscillation of the axial and lateral stresses. In

639

the case of OWPPs such changes are caused by the variation of the direction of wind and wave loading. ACKNOWLEDGEMENTS This work has been done in the framework of the project “Geotechnical robustness and self-healing of foundations of offshore wind power plants” funded by the German Federal Ministry for the Environment, Nature Conservation and Nuclear Savety (BMU) (grant No. 0327618). The authors are grateful to BMU for the financial support. Furthermore, the authors wish to thank H. Borowski who performed the cyclic triaxial tests. REFERENCES Achmus, M.,Abdel-Rahman, K. & Kuo,Y.-S. 2008. Design of monopile foundations for offshore wind energy converters. In Z. Mlynarek, Z. Zikora, and E. Dembicki, editors, Goetechnics in Maritime Engineering, Proc. of 11th Baltic Sea Geotechnical Conference, 1:463–470. American Petroleum Institute (API). 1993. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design. API RP 2A – WSD, Vol. 20, Dallas. Dührkop, J. 2010. Zum Einfluss von Aufweitungen und zyklischen Lasten auf das Verformungsverhalten lateral beanspruchter Pfähle in Sand. Dissertation, Veröffentlichungen des Institutes für Geotechnik und Baubetrieb der Technischen Universität Hamburg-Harburg, Heft Nr. 20. Germanischer Lloyd. 2005. Rules and Guidelines, IV Industrial Services, 2 Guideline for the Certification of Offshore Wind Turbines, 6 Structures. Herle, I. 1997. Hypoplastizität und Granulometrie einfacher Korngerüste. Promotion, Institut für Bodenmechanik und Felsmechanik der Universität Fridericiana in Karlsruhe, Heft Nr. 142. Hettler. A. 1981. Verschiebungen starrer und elastischer Gründungskörper in Sand bei monotoner und zyklischer Belastung. Institut für Boden- und Felsmechanik der Universität Karlsruhe, Heft Nr. 90.

© 2011 by Taylor & Francis Group, LLC

Lesny K. & Hinz, P. 2006. A concept for a safe and economic design of foundations for offshore wind energy converters. In New Approach to Harbour, Coastal Risk Management and Education, Proc. of LITTORAL 2006, Gdansk, Poland, 90–98. Lin, S.-S. & Liao, J.-C. 1999. Permanent Strains of Piles in Sand due to Cyclic Lateral Loads. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 125(9):798– 802. Little, R.L. & Briaud, J.-L. 1988. A pressuremeter method for single piles subjected to cyclic lateral loads in sand. Technical Report GL-88-14, US Army Corps of Engineers. Long, J.H. & Vanneste, G. 1994. Effects of cyclic lateral loads on piles in sand. Journal of Geotechnical Engineering, ASCE, 120(1):225–244. Niemunis, A. & Herle, I. 1997. Hypoplastic model for cohesionless soils with elastic strain range. Mechanics of Cohesive-Frictional Materials, 2:279–299. Niemunis, A., Wichtmann, T. & Triantafyllidis, T. 2005. A high-cycle accumulation model for sand. Computers and Geotechnics, 32(4):245–263. Reese, L.C., Cox, W.R. & Koop, F.D. 1974. Analysis of laterally loaded piles in sand. In Proceedings of the 6th Annual Offshore Technology Conference, Houston, Texas, OTC 2080, pages 473–458. Swinianski, J. & Sawicki, A. 1991. A model of soil pile interaction owing to cyclic loading. Canadian Geotechnical Journal, 28(1):11–19. Det Norske Veritas. 2004. Offshore Standard DNV-OSJ101: Design of Offshore Wind Turbine Structures. von Wolffersdorff, P.-A. 1996. A hypoplastic relation for granular materials with a predefined limit state surface. Mechanics of Cohesive-Frictional Materials, 1:251–271. Welch R.C. & Reese, L.C. 1972. Laterally loaded behavior of drilled shafts. Technical Report 3-5-65-89, Center for Highway Research, University of Texas, Austin. Wichtmann, T., Rojas, B., Niemunis, A. & Triantafyllidis, T. 2010. Stress- and strain-controlled undrained cyclic triaxial tests on a fine sand for a high-cycle accumulation model. In Proc. of the Fifth International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil Dynamics, San Diego, USA.

640

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Cyclic accumulation effects at foundations for offshore wind turbines H. Wienbroer, H. Zachert, G. Huber, P. Kudella & Th. Triantafyllidis Karlsruhe Institute of Technology (KIT), Institute of Soil Mechanics and Rock Mechanics, Karlsruhe, Germany

ABSTRACT: Offshore wind turbines will reach their serviceability limit at a certain permanent inclination of the structure, which can be caused by a strong storm event. If the permanent deformation after such an event is sufficiently large, the allowed range for the operation of the turbine may be exceeded. Self-healing means that this permanent inclination can at least partly be reverted by a rather regular exposure to wind and waves with smaller amplitude. The ability of the system to show this behavior depends on the design of the foundation and the type and state of subsoil. This paper focuses on shallow foundations on dense to medium dense sand. The setup for an experimental investigation of this phenomenon in small scale model tests is presented in detail. It can be shown that the tests confirm the self-healing ability of foundations exposed to cyclic loading. The influence of the loading sequence is also presented. 1

INTRODUCTION

Due to the growing interest in renewable energy sources the offshore wind energy production grows as one of the leading fields of investment in Europe. The German Federal Maritime and Hydrographic Agency (Bundesamt für Seeschifffahrt und Hydrographie, BSH) approved 25 offshore wind projects with a total number of 1769 wind turbines. Most of them are located in the North Sea and will be built in the next 10 to 20 years. The existing experiences in the installation and operation of offshore structures were gained in the field of oil and gas exploration; see for example Bjerrum (1973), Randolph et al. (2005) or Andersen (2009). As the loading conditions for wind turbines show high moments due to the wind exposure, the BSH issued standards with recommendations for the design of the structure and the foundation (BSH 2007). This standard demands compliance with the standards DIN 1045 (2005) and Eurocode 7 (2005) allowing several international standards to be taken as a basis. Different types of foundations like shallow foundations with and without skirts or deep foundations with a single pile or several piles (tripod, jacket) are discussed.

cyclic loading sequence). For a shallow foundation these verifications are normally done by static bearing capacity and eccentricity calculations (e.g. Randolph et al. 2005, Andersen 2009). The serviceability of such a structure can be defined via settlement and inclination (differential settlement). The settlements are caused by static (submerged weight of the structure) and cyclic load portions from wind and waves. Settlements due to cyclic loading can be estimated by multiplying the settlements from static calculations with a certain factor, including the number of cycles as a logarithmic function (e.g. Mallwitz & Holzlöhner 1996). Differential settlements can be caused by existing inhomogeneities in the density of the subsoil or by density changes due to static or cyclic shearing. Gajan & Kutter (2008) for example investigated the rocking behaviour of shallow foundations under earthquake like loading. Although there are some differences in the loading sequences, the observed dependencies on the contact area between foundation and subsoil should be comparable. The effect of self-healing discussed in the next section can be seen as a special case of differential settlement accumulation due to a special combination of loading sequences.

1.1 Limit states and accumulated deformations Following the BSH standard two so-called Limit States have to be considered regarding the foundation of an offshore wind turbine: Ultimate Limit State – ULS and Serviceability Limit State – SLS. The loads which have to be taken by the foundation in the ULS are mainly caused by water waves during a storm event. This can be a huge single wave – a so-called freak wave (Clauss 2008) – or several big waves. In this case the stability against collapse or overturning of the structure has to be proven (for a © 2011 by Taylor & Francis Group, LLC

1.2

Self-healing effect

We are starting with the idea that wind and waves of a strong storm may cause a permanent inclination of a turbine structure. This inclined state must not exceed the ULS – stability problems are not part of the discussion here. Only the Serviceability Limit State shall be exceeded after the storm. The SLS may here be defined by the maximum inclination of the structure allowed for the operation of the turbine. Self-healing now means that there are certain possibilities to at least

641

Figure 2. Qualitative density changes under a shallow foundation for typical loading.

Figure 1. Principal mechanism of self-healing.

partly revert this inclination due to regular exposure to wind and waves showing smaller load amplitudes (cf. Figure 1). The ability to show this behavior depends on the design of the structure and subsoil type and density. This contribution focuses on shallow foundations on dense sand.

sequence is essential to induce cyclic compaction. With monotonic loading no self-healing will be observed. After a sufficient number of cycles natural and man-made (installation process) inhomogeneities can get equalized. This is generally accompanied by a permanent inclination of the structure which can not be reverted. In addition the soil must have a sufficient high permeability to ensure that pore pressure can at least partially dissipate within a single load period. Self-healing also does not occur in very loose soil as there is no further loosening on the windward side but strong compaction on both sides – the foundation simply just settles. Another requirement is that the location of the structures centre of gravity is deep enough to avoid collapse in the inclined position. The following experimental investigations concentrate on a foundation with very deep location of the centre of gravity standing on very dense and dry sand. Therefore the expected self-healing effect should be very pronounced concerning these issues. The usage of dry samples can be justified by assuming fully drained conditions at any time.

2 1.3

MODEL TESTS

Shallow foundations

For a shallow foundation the actuation for self-healing is caused by different compaction accumulation rates at different positions under the foundation. The forces from wind and waves are transferred through the structure into the foundation interacting with the soil. The storm loading with Fmax (0 to 1 in Figure 2) causes a compaction of the soil on the leeward side and an unloading associated with increasing void ratio on the uplift or windward side. This results in a smaller compaction on the windward side after unloading (2 in Figure 2) and ends up with a residual leeward rotation of the structure. If the strong loading is followed by small cycles with Fampl , cyclic compaction occurs and the soil on the windward side densifies faster than on the leeward side (3 in Figure 2). As a result, the structure rotates partially back into a more horizontal position. Several requirements have to be fulfilled to enable this self-healing process. First a cyclic loading © 2011 by Taylor & Francis Group, LLC

2.1

Loading history

The real loading sequences for storm and regular weather conditions are very complex. First the direction of loading changes with time and, in addition, the wind and wave directions are not necessarily the same. For the storm event, the direction of loading changes only within a rather limited sector, but for the regular loading almost in all directions. The loading is composed from different parts, as there are wind, waves and current. For the storm event, one can assume that the loading of the structure is dominated by a rather low number of strong waves. The regular loading sequence can be considered as combination of a monotonic average force from current and wind and cyclic force amplitude from wind and waves. Both vary with time in rather long periods, and can therefore be called alternating or quasi-static (f < 5 Hz). There are also dynamic loads induced into the structure by the rotation of the turbine. But this is regarded to be

642

Figure 3. Idealized loading history (force vs. time).

of minor importance for the soil-structure-interaction compared to the wind and wave action. For a first experimental insight we used a simplified loading sequence consisting of a single large amplitude Fmax followed by an average force Fav , superimposed by a certain number N of smaller amplitudes Fampl . The loading was carried out in only one direction with a period of T = 6s (f = 0.17 Hz). The used sequence in the model tests can be idealized as shown in Figure 3.

Figure 4. Test setup with model foundation and loading frame.

2.2 Test setup The whole setup consists of a steel container (diameter Dcont. = 0.94 m, height hcont. = 1.5 m) filled with sand, the model foundation placed on it and the measurement and loading frames mounted on top (cf. Figure 4). Representing the soil conditions in the North Sea we used a natural fine grained quartz sand with d50 = 0.14 mm (uniformity coefficient Cu = 1.5, critical friction angle φc = 33◦ ). The sand is placed dry and in 15 vibro-compacted layers, which results in a mean relative density of about 97% (emax = 1.05, emin = 0.67). The loading frame carries the pneumatic cylinders which apply the load sequence. The cylinders have cardan joints and are weight compensated by springs in order to avoid additional forces on the foundation. The dimensioning of the foundation model was done for a circular plate and comprises: the diameter of the plate Dplate , the mass of the foundation m, the height of the point of load application h and the maximum horizontal load Fmax (cf. Figure 5). The diameter of the plate was chosen to be Dplate = 0.26 m, which corresponds to a ratio of diameters Dcont. /Dplate ≈ 3.5. For a first estimation of the horizontal force F we took the ULS design loads (horizontal force and momentum at the base plate) for a prototype foundation provided by our project partner Ed. Züblin AG and scaled them down with a factor of 1:125. This resulted in a horizontal force of Fmax ≈ 5N acting on the foundation shaft at a height of about 0.41m. This is the only dimension calculated with a scaling law as the design of the prototype foundation is completely different from our model foundation. © 2011 by Taylor & Francis Group, LLC

Figure 5. Principal scheme of the application points of the depicted measuring quantities.

The mass of the model foundation was estimated by limiting the eccentricity. With the eccentricity

and the already fixed values of Dplate , F and h this results in the mass m. An earlier experimental investigation was done with an octagonal plate (coextensive) dimensioned for Fmax ≈ 5N, see Wienbroer et al. (2010). The mass of this model was estimated as m ≈ 5 kg by limiting the eccentricity to r1 = 0.167 · Dplate according to Borowicka (1943). Later on, in order to enable higher relations Fmax /Fampl and for symmetry reasons, we changed the design to a circular plate and

643

doubled the weight of the foundation. The overturning stability was calculated according to the German standard DIN 1054 (2005) (e < r2 = 0.295 · Dplate ) as well as the safety against base failure. In addition, DIN 1054 (2005) requires e < r1 = 0.125 · Dplate under static loads as a serviceability limit state criteria to avoid a gap. So we obtained a model foundation with m ≈ 10 kg, achieving the r1 -limit due to DIN 1054 (2005) with Fstat = Fav ≈ 9N and the r2 -limit with Fmax ≈ 19N. Even for Fmax , the safety against base failure is still in the range of 13, which means that the DIN-compliant design is conservative and rather governed by serviceability considerations than by base failure. The plate was always equipped with a 0.55 m high rod for the application of the horizontal force.

the vertical transducers u1 , u2 and u3 . The displacements ux and uy can be calculated from the horizontal transducers u4 , u5 and u6 (Figure 5). The transducers work in a range of 0 to 30 mm with a resolution of about 0.1 µm. As the used displacement transducers work with pneumatically actuated feeler pins exerting forces on the plate (≈1.5N per transducer), this configuration has the advantage of a zero resulting horizontal force and moment. The disadvantage of the 120◦ -configuration is that the rotation θz can not be derived. A data acquisition system which works synchronous to the loading device was developed. It is triggered by the PLC. With this device only two points per cycle at the upper and lower reversal points are logged, which is quite appropriate for larger numbers of cycles.

2.3 Loading device As already discussed, the wind turbine structure can be regarded as a force controlled system. As this should also hold true for model tests, the loading device was chosen to be of pneumatic nature. With the estimated values of Fmax , the pneumatic system was designed for a force range between 0.05 and 50N. Considering the pneumatic cylinders to have a force-pressure relation of kcyl = 6N/bar, we are working in a pressure range of about pcyl = 0.01 to 8bar. The connection between cylinder and foundation system is a thin tension wire – therefore two cylinders are mounted in one direction operating against each other. Two of these pairs were mounted normal to each other to allow bi-dimensional loading sequences. For the realization of the simplified loading sequence in Figure 3, five independent air pressures per cylinder can be actuated via electromagnetic valves. The pressures can be adjusted with 5 pressure regulating valves, measured with a pressure transducer (10bar range, 10 mbar resolution). The system is controlled by a Programmable Logic Controller (PLC) and synchronized with a data acquisition unit. As the pressure build-up in the pressure chambers of the cylinders takes some time (≈1.5 s) the resulting real loading curve is smoother than the idealized one in Figure 3.

2.4 Instrumentation The horizontal load F was calculated from the pressures measured at the loading device, taking into account the friction loss of the pneumatic cylinders (≈1.5% of pcyl ). A direct force measurement via a load cell mounted on the pylon was done in earlier tests. The differences of the two redundant systems stayed inside the resolutions of the transducers (0.2N for the load cell). The motion of the foundation plate in 3D-space can by described by the displacements ux , uy and uz and the rotations θx , θy and θz (see Figure 5). In order to derive these motion quantities, 6 displacement transducers were placed in a 120◦ -configuration. The values of uz , θx and θy can be calculated from the measurements of © 2011 by Taylor & Francis Group, LLC

3 3.1

EXPERIMENTAL RESULTS Reference example

The presented tests were all performed onedimensional with the applied load in x-direction. In order to compare different loading schemes, one loading regime was set as a reference (cf. Test I in Table 1). This reference test was initially exposed to N = 500 symmetric cycles (Fav = 0.0N) with the loading amplitude Fampl = 3.0N (peak-peak). These pre-cycles should homogenize the subsoil and smooth possible inhomogeneities due to the installation of the foundation. They were followed by a maximum load Fmax = 20.0N accompanied by 10000 symmetric cycles with Fampl = 3.0N. This sequence of Fmax and Fampl was repeated 3 times. The resulting displacements and rotations are shown in Figure 6. Fmax in x-direction leads to a strong rotation around the y-axis (θmax ). θmax is reduced to θ0 (remaining rotation after Fmax ) during unloading. Due to the following cycles with Fampl this permanent rotation is significantly reduced. Beside this self-healing effect Figure 6 shows a decrease of θmax due to the second and the third maximum load Fmax . The cyclic loading leads to a compaction of the soil and a stiffer response of the system on the applied loads. This is supported by Figure 7, where the backward rotation or self-healing of the three sequences is compared. The self-healing θˆ during a single Fampl loading period was calculated with the following equation:

using the values of θ according to Figure 6. During the second and the third sequence it occurs slower than in the first step but reaches almost 100%. That means, only the first Fmax leads to a relevant remaining inclination. As this holds true for all performed tests (see Table 1) the backward rotations in Figure 8 are only shown for the third sequence.

644

Figure 6. Experimental result for symmetric loading cycles (u, θ vs. n) – reference test.

Table 1.

List of performed tests.

Test

Fmax N

Fampl N

Fav N

I II III IV V

20.0 20.0 20.0 20.0 20.0

3.0 4.0 2.0 3.0 3.0

0.0 0.0 0.0 1.5 3.0

Figure 7. Backward rotation for the reference test.

3.2 Amplitude variations In order to investigate the effects of different loading schemes on the self-healing, first experiments with different cyclic loads Fampl and static average forces Fav were performed. Figure 8 depicts the backward rotation θˆ of the 5 different experiments listed in Table 1. Test III shows a slower backward rotation for a smaller amplitude Fampl but Test II does not show a faster self-healing for a larger Fampl , as one would © 2011 by Taylor & Francis Group, LLC

645

Figure 8. Comparison of the backward rotations.

expect. This might be caused by experimental uncertainties, like small differences in density or homogeneity, which produce a certain scatter in the measurement results. There could also be something like a threshold in amplitude Fampl causing only small changes in the self-healing behavior beyond the threshold. Tests IV and V indicate that an average force Fav causes a clear deceleration of the backward rotation, which also would be expected. 4

CONCLUSIONS

The obtained test results evidently show that the developed experimental setup is able to reproduce the proposed self-healing effect. For the measurement equipment we concern a quite good resolution and sufficient arrangement concerning the motion quantities, although we did not realize a full 6 degree of freedom displacement measurement as it was done for example by Byrne & Houlsby (2005). The tests performed up to now show clear hints on the influences and limits of the amplitude size and the average force, but more tests are needed to support these first results. The supposed threshold in amplitude size has to be clarified, as this would have a strong influence on the needed time for a significant amount of backward rotation. The same holds true for the mean average force in comparison to the maximum load. The experimental setup will be improved by a different sand placement method, allowing medium dense to dense saturated samples. In a next step the influence of the orientation of loading on the deformation behavior of the foundation system will be studied. Furthermore arbitrary loading sequences will be applied leading to individual deformation paths. The observed self-healing effect is only one of many possible deformation path resulting from a special loading sequence. The obtained results will be used for some numerical studies of the system in model and prototype scale in order to provide a tool for a safer and more economic design of such foundation systems, enabling the designers to calculate serviceability states of the offshore wind turbine foundation. First numerical success in the simulation of self-healing could already be gained with back calculations of large scale in-situ model tests on caissons (Sturm et al. 2008). ACKNOWLEDGEMENT

Conversation and Nuclear Safety (BMU) under the research grant no. 0327618 – Geotechnical Robustness and Self-Recovery in Setting the Foundation of Offshore Wind Turbines (Geotechnische Robustheit und Selbstheilung bei der Gründung von OffshoreWindenergieanlagen). REFERENCES Andersen, K. 2009. Bearing capacity under cyclic loading offshore, along the coast, and on land. The 21st Bjerrum Lecture presented in Oslo, 23 November 2007. Canadian Geotechnical Journal, Vol. 46(5), 513–535. Bjerrum, L. 1973. Geotechnical problems involved in foundations of structures in the North Sea. Géotechnique, Vol. 23(3), 319–358. Borowicka, H. 1943. Über ausmittig belastete, starre Platten auf elastisch-isotropem Untergrund. Archive of Applied Mechanics, Vol. 14(1), 1–8. BSH – Bundesamt für Seeschifffahrt und Hydrographie 2007. Standard, Konstruktive Ausführung von Offshore – Windenergieanlagen. BSH-Nr. 7005. Byrne, B.W. & Houlsby, G.T. 2005. Investigating 6 degreeof-freedom loading on shallow foundations. In Gourvenec & Cassidy (eds.). Int. Symp. on Frontiers in Offshore Geotechnics, Perth, Australia, Taylor & Francis, 477–482. Clauss, G. 2008. The taming of the shrew – Tailoring freak waves for seakeeping tests. Journal of Ship Research, Vol. 52(3), 194–226. DIN 1054:2005-01. Baugrund – Sicherheitsnachweise im Erd- und Grundbau. Normenausschuss Bauwesen im Deutschen Institut für Normung e.V. DIN EN 1997-1:2005-10. Eurocode 7: Entwurf, Berechnung und Bemessung in der Geotechnik, Teil 1: Allgemeine Regeln. Deutsche Fassung EN 1997-1, Normenausschuss Bauwesen im Deutschen Institut für Normung e.V. Gajan, S. & Kutter, B.L. 2008. Capacity, Settlement and Energy Dissipation of Shallow Footings Subjected to Rocking. Journal of Geotechnical and Geoenvironmental Engineering, Vol. 134(8), 1129–1141. Mallwitz. K. & Holzlöhner, U. 1996. Verfahren zur Ermittlung der Setzung von Fundamenten infolge zyklischer Beanspruchung. Bautechnik, Vol. 73(3), 175–186. Sturm, H., Solf, O. & Kudella, P. 2008. Self-healing effects of shallow foundations for offshore wind turbine structures. In Z. Mlynarek, Z. Sikora & E. Dembicki (eds.). 11th Baltic Sea Conference on Geotechnics in Maritime Engineering, Gdansk, Poland, Vol. 1, 301–308. Randolph, M., Cassidy, M. & Gourvenec, S. 2005. Challanges of offshore geotechnical engineering. 16th Int. Symp. Soil Mech. and Geotech. Enging. (ISSMGE), Osaka, Japan, Balkema Vol. 1, 123–176. Wienbroer, H., Huber, G. & Triantafyllidis, Th. 2010. Cyclic deformations of foundations for offshore wind turbine structures. 7th International Conference on Physical Modelling in Geotechnics, Zurich, Switzerland, paper 10409.

The authors are grateful for the financial support of the Federal Ministry for the Environment, Nature

© 2011 by Taylor & Francis Group, LLC

646

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Study on soil-structure interaction of suction caisson by large-scale model tests B. Zhu, D.Q. Kong, L.G. Kong, R.P. Chen & Y.M. Chen MOE Key Laboratory of Soft Soil and Geo-environmental Engineering, Zhejiang University, Hangzhou, China Department of Civil Engineering, Zhejiang University, Hangzhou, China

ABSTRACT: A programme of large-scale model testes for the suction caisson subjected to overturning moment loads in silt were carried out in a large tank (15 × 5 × 6 m), to study the soil-structure interaction of the suction caisson used as the monopod foundation of offshore wind turbines. The caisson used was 1 m in diameter and 0.5 m in skirt length, with distributed soil pressure cells embedded in the lid and skirt. Interaction of the lid and soil plug inside the caisson as well as the distribution of lateral earth pressures on the caisson skirt was obtained. Displacement of the soil surface as well as the caisson was measured by LVDTs. Based on these test results, an assumption of a common position of instantaneous rotation center and dominating resistance forces on the caisson was presented, which can be used for further analytical calculating method of ultimate moment capacity of the caisson.

1

INTRODUCTION

As a strategic choice to reduce carbon dioxide emissions as well as to solve global energy shortages, developing wind energy is one of the most promising and fastest growing industries of renewable energy sources. A obligatory requirement that the total installed capacity of wind power in China should reach at least 30 million kilowatts by 2020 was presented by Renewable Energy Law of The People’s Republic of China in 2006. After that, China’s wind power industry has made rapid development in recent years, with a total installed capacity of 25.1 million kilowatts by the end of 2009, about 16% of the total capacity in the world. Due to the advantages of fast and stable wind speed, high energy yield, less wind turbulence as well as less constraint of land expropriation and noise, offshore wind power development has become an international trend. China’s first large wind turbine based on an offshore oil platform in Bohai Sea was put into operation in 2007 officially and the first offshore wind farm, Shanghai Donghai Bridge Offshore Wind Farm, has started construction in 2008. The first three wind turbines in this farm have been in operation in 2009 and the remaining will be finished construction by the end of 2010. This wind farm will have 34 3.0 MW wind turbines produced by Sinovel Wind Co., Ltd of China. The foundation used here is a pile group type with a concrete cap. In each foundation, there are 8 steel pipe piles each with a length of 85 m, diameter of 1.7 m and thickness of 3 cm. Besides, the present Dongtai Beach Wind Farm in Jiangsu province, the Cixi Beach Wind © 2011 by Taylor & Francis Group, LLC

Farm in Zhejiang province and a few other beach wind farms will be ex tended to offshore. Several other new offshore wind farms are also in plan as well. The foundation is used to carry all loads of superstructures, and it is one of the most important components in the design procedure. Compared with those built farms onshore, the offshore wind turbine foundation has an additional 30 percent in cost. The soil of present and potential sites for offshore wind farms in China are mainly soft clay and fine sand/sandy silt. In the case of soft clay seabed, multi-type foundations, such as multi-pile group or multi-caisson, are comparatively suitable; while for fine sand/sandy silt seabed, single footing foundations, for example, mono-pile, gravity base foundation (GBF) and the mono-caisson are more economical. The major load on the offshore wind turbine is overturning moment due to horizontal forces caused by wind and wave, and thus bearing capacity against overturning is the main issue to be considered in the design of single footing foundation. The three kinds of mentioned mono-type foundation show approximate bearing capacity, while the mono-caisson needs less quantity of steel but more work of welding, however, it is more economical in total (LeBlanc 2005). More importantly, large special vessels necessary for the construction of large-diameter piles and gravity base foundations are rarely available in China, which makes mono-caisson an important alternative in the construction of wind farms in fine sand/sandy silt seabed. The first suction caisson foundation was installed onshore for a V90 wind turbine in Frederikshavn,

647

Denmark, in 2002. Three years later, a large suction caisson, with a diameter of 16 m and a skirt length of 15 m, was constructed to support a 6 MW Enercon offshore wind turbine. However, the suction caisson was failed with skirt buckled due to a horizontal vessel impact during suction installation. Recently, Dong Energy has successfully installed a suction caisson foundation for a meteorology mast at Horns Rev 2 Offshore Wind Farm, Denmark, in February 2009, and this company seems to apply this novel foundation to offshore wind farms widely (LeBlanc, 2004). A lot of work has been carried out for the study on behaviors of the caisson, mainly concentrated on the matter of suction penetration and bearing capacity of pullout and overturning moment, in the form of monocaisson and multi-caisson (Byrne and Houlsby, 2003; Chen and Randolph, 2007). The available installation and design experiences of the monopod suction caisson foundation for offshore wind turbines is still of limitation, mainly due to its complicated interaction with the soil. The existing data obtained from neither laboratory model tests nor field trails is sufficient, compared with those of piles. Furthermore, the soil-structure interaction under overturning moment has not been investigated adequately yet. Thus, despite its promising prospect of engineering application, the suction caisson foundation developed rather slowly so far. Compared with the studies in clay, the loaddisplacement relationship and bearing capacity of the suction caisson in cohesionless soils are more difficult to investigate. Available calculation approaches of the ultimate moment capacity of caisson foundation installed in sand were based on test fitting test results of smallscale laboratory model tests, mainly provided by the research group at the University of Oxford (Byrne and Houlsby, 2003; Villalobos 2006). These empirical approaches may give satisfied prediction of ultimate moment capacity of the suction caisson for specified soils used in tests, but they are non-sensitive for properties of soils and geometric parameters of the caisson. It is an alternative way to calculate the ultimate moment capacity of the caisson on the basis of the study of soilstructure interaction. Furthermore, the soil type in the sites for most of present beach wind farms and potential offshore wind farms in China is silt, which calls for studies on overturning bearing capacity of foundations for the wind turbine in silt. A programme of large-scale model tests of suction caisson was carried out in a large soil tank (length of 15 m, width of 5 m and depth of 6 m) in Zhejiang University to learn the deformation mechanism as well as soil-structure interaction of the caisson foundation. The soil used in tests was silt, which is typically in most potential construction sites for offshore wind farms in China. Based on the test results, an assumption of a common position of instantaneous rotation center and dominating resistance forces on the caisson was presented.

© 2011 by Taylor & Francis Group, LLC

Figure 1. Large soil tank (15 × 5 × 6 m) in Zhejiang University.

2 TEST PROGRAMME Based on the non-dimensional analysis method for 1 g model tests of caisson foundations, the laboratory data from the small-scale moment loading tests are in well agreement with the field ones (Kelly et al. 2006). Thus, the laboratory large-scale model tests of the caisson will give good reliable results for in-situ full-size caissons. 2.1

Soil tank and soil preparation

The present large-scale model tests for suction caisson subjected to the overturning load were carried out in a large soil tank as illustrated in Fig. 1. This soil tank has an inside dimensions of 15 × 5 m in plan view and depth of 6 m. The side walls of the tank consist of 5 mm thick steel plate which is braced to provide lateral rigidity to the sidewalls. Using a separating steel plate, the tank can be divided into two parts with lengths of 5 m and 10 m respectively. All model tests of the caisson were carried out in the small part with the plan size of 5 × 5 m. A water level control system was designed to regulate the rise and drawdown of water in the tank. The water level control system consists of a movable water storage box along a guide, a main supply tube and a network of branch tubes with water exit points. The water storage box is equipped with a constant-head control device and can be elevated or lowered to any level through the use of a manual hoist. A network of branch tubes is designed to distribute water throughout the bottom of the box. The spacing of the water exit points along each branch tube is 200 mm. The network of branch tubes is embedded in a layer of coarse sand with a thickness of about 200 mm to provide uniformity of the water pressure distribution. The soil used in this model test is sandy silt obtained from an excavation near the Qiantang River in Hangzhou which is similar to the soils of two wind

648

Figure 2. Particle distribution curve of soils.

farms sites, Cixi of Zhejiang province and Dongtai of Jiangsu province, both of which are built in beach area and will be extended to the offshore. The particle size distribution was obtained by a hydrometer test and the results are shown in Fig. 2. The soil consists of 12% sand, 80% silt and less than 5% clay size particles. The specific gravity of the soil is 2.69. Its liquid index and plasticity index are 31.7 and 22.6, respectively. The maximum dry density from a standard compaction test is 1570 kg/m3 and the corresponding optimum water content is 18%. The coefficient of permeability, k of the sandy silt ranges from 4.2 × 10−6 m/s to 6.4 × 10−6 m/s. with an average value of 5.3 × 10−6 m/s. The shear strength of the silt was measured through conventional isotropic consolidated, undrained triaxial compression tests. The tests showed that the internal friction angle of the sandy silt is 30◦ . The soil was filled by dynamic compaction into the soil tank layer by layer, with a total depth of 520 cm. Soil in the bottom was divided into 21 layers, each had a thickness of about 20 cm after compaction while in the top 100 cm thickness each layer was 5 cm in thickness after compaction.After each layer was compacted, soil sample was taken from four different positions in the tank for the measurement of density by cutting ring method. The density of the soil sample ranges from 1630 kg/m3 to 1670 kg/m3 , indicating a desirable homogeneity. Water content was also obtained and the average value was 12%. After filling, the soil was saturated by elevating the water storage box along a guide, which would cause a uniformly upward penetration in the soil. The water storage box was elevated gradually with an approximate velocity of 0.3 m/day to prevent uneven seepage or piping. After the completion of soil saturation, the water level was kept 2 cm higher above the soil surface, to make sure that the soil was saturated during the tests. Saturated density of the soil was calculated as 1920 kg/m3 , with a relative density of 70%. © 2011 by Taylor & Francis Group, LLC

Figure 3. Arrangement of transducers (unit: m).

2.2 Test setup The used model caisson is with an internal diameter (D) of 100 cm and a clear skirt height (L) of 50 cm. Thicknesses of the lid and wall are 2 cm and 0.5 cm respectively. Including an exhaust valve and a pumping valve, there were 4 holes of 5 cm in diameter on the lid, which were sealed during suction installation and lateral loading but could be opened for pushing and “net” pullout the caisson without suction as well as CPT tests inside the caisson. A steel pipe connected to the lid of the caisson was used to apply horizontal loads and guide the suction installation of the caisson. It was 21.9 cm in outside diameter, 2 cm in thickness and 260 cm in height. The caisson and the pipe weight 170 kg and 236 kg respectively. As shown in Fig. 3, soil pressures between the caisson and soil were measured by BYB-3 soil pressure cells installed under the lid and on the walls both inside and outside, along the direction of lateral loading. Displacement of the caisson and the soil surface was measured by LVDTs during loading, while penetration depth was measured by two laser displacement transducers during suction penetration. A Fluke system was used for data collection. In addition, two vacuum meters were used to measure the suction under the lid of caisson during suction penetration. Arrangement of these transducers is illustrated in Figs. 3. A common caisson was suction installed and loaded at six different positions in the tank, in order to investigate the effects of the pumping rate during suction installation and the loading eccentricity M /(HD) in overturning loading. Before installation, the required

649

Figure 4. Relationship between moment and angular rotation.

suction was predicted according to the method mainly provided by Senders and Randolph (2009). In the overturning loading tests, weight load was laterally applied on the steel pipe using a series of pulleys. Loading heights to the lid are 1 m, 1.5 m and 2.6 m, respectively, which are same as their loading eccentricities. These values of loading eccentricity M/(HD) were chosen around the actual case in field, which was approximately 1.5 in most cases. Due to the limitation of the length of this paper, results of overturning loading test with loading eccentricity M /(HD) = 1 are mainly given herein. 3 3.1

RESULTS OF MONOTONIC OVERTURNING LOADING TESTS Moment loading

Instantaneous rotation centre of caisson

With the displacements measured by two vertical LVDTs applied on the edges of the caisson and one © 2011 by Taylor & Francis Group, LLC

horizontal LVDT applied on the connecting tube (see Fig. 3a), positions of instantaneous rotation centre of the caisson can be determined as given in Fig. 5. For different loading heights, the depth of Instantaneous rotation centre was about 0.8L below the soil surface with slight deviation, while the lateral distances away from the caisson center have some difference. It can also be observed that the Instantaneous rotation centre is closer to the skirt at the loading direction with decreasing of M /(HD). For an offshore wind turbine, the ratio of M /(HD) usually ranges between 1.5 and 2.6. From Fig. 5, the instantaneous rotation centre of the caisson with the aspect ratio L/D of 0.5 can be seemed as at 0.8 L depth and right below the centre of lid of the caisson with values of M /(HD) ranging from 1.5 to 2.6. 3.3

The overturning load was applied step by step, each load step was sustained for at least 30 minutes. The time is enough for the dissipation of 50 percent pore water pressure according to the calculated method by Byrne (2000). The vertical load applied in the overturning tests was the self weight of the caisson and the connecting tube (with a value of 4 kN). This value was chosen to be a little bigger than the one in field by the analysis of normalization (Kelly et al. 2006), to ensure a better study on the soil-structure interaction under the lid. If the angular rotation difference of the caisson foundation is smaller than 0.0002 in two continuous 10 minutes for three times, then the next load step was applied. The relationship between overturning moment and angular rotation of the caisson for three different loading eccentricities is presented in Fig. 4. 3.2

Figure 5. Positions of Instantaneous rotation centre.

Soil-structure interaction of caisson

For the two vertical LVDTs on the lid and the others on the ground surface (see Figs. 3), the ratio of their measured vertical displacements during lateral loading to diameter of the caisson is shown in Fig. 6. Displacement of the lid shows an upward trend as a whole. The ground surface at the loading direction is slightly upheaved with the maximum s/D of about 0.004, while in the other side the ground surface is subsided. The horizontal coordinate range the vertical displacement of the ground surface taking place can be approximately calculated by the Rankine’s earth pressure theory, and their values are 1.37D and 0.79D away from the central line of the caisson at and opposite the loading direction, respectively. The measured incremental lateral soil pressures inside the skirt are rather small that they are not presented in this paper due to the length limitation. Other incremental soil pressures obtained during overturning loading are given as follows. For different amplitude of the lateral loads with the ratio M /(DH ) of 1, the incremental soil pressures on the outer skirt of the caisson at the loading direction are shown in Fig. 7. It can be found that distribution

650

Figure 6. Vertical displacement of the ground surface and lid.

Figure 8. Distribution of incremental soil pressures outside skirt and opposite loading direction.

Figure 7. Distribution of incremental soil pressures outside skirt and at loading direction.

of soil pressures for a certain load is approximately binomial in relation to the depth. The position of the instantaneous rotation centre can be taken as the depth where the incremental soil pressures equal to zero. Distribution of incremental soil pressures on the outer skirt opposite the loading direction was also obtained. The results associated with the differential between the active soil pressures and the initial soil pressures are shown in Fig. 8. These incremental soil pressures are almost the same for different overturning moment loads. Moreover, only the shallow soil becomes active failure, and soil pressures at relatively large depth are between the initial one and the active one. Incremental soil pressures under the lid for different moment loads are presented in Fig. 9. Obvious reduction of soil pressures opposite the loading direction is observed as well as small increase at the loading direction. Ultimate reduction of the former ones is approximately the ratio of total weight of the caisson (together with the connecting tube) to the lid area. © 2011 by Taylor & Francis Group, LLC

Figure 9. Distribution of incremental soil pressures under lid.

Based on measured incremental soil pressures and positions of instantaneous rotation centre, a simplified force distribution for the failure mode of the caisson subjected to the overturning moment load is presented in Fig. 10. In the diagram, Ka is the active earth pressure coefficient, K0 is the coefficient of earth pressure at rest,V and H are the vertical and lateral load on the caisson, Kv is the vertical soil reaction coefficient, Kh is the lateral soil reaction coefficient, z0 is the depth of the instantaneous rotation centre of the caisson and θ is the angular rotation of the caisson, σzp and σzp are the soil pressures on the outer skirt at and opposite the loading direction, pu is the ultimate bearing capacity of the soil. Attention should be paid that soil pressures illustrated in the diagram are gained by adding incremental soil pressures to the initial soil pressures, which was taken as earth pressure at rest in the field. The suggested force distribution shown in Fig. 10 can be used for a new calculation method of ultimate moment capacity of the caisson in further work.

651

Natural Science Foundation of China (research grant: 50979097). All technicians especially Mr. S. Y. Wang in Laboratory of Soft Soil and Geo-environmental Engineering of Zhejiang University are also acknowledged. REFERENCES

Figure 10. Simplified force distribution of caisson.

4

CONCLUSIONS

A programme of large-scale model tests of suction caissons for the offshore wind turbine in silt was presented. Positions of instantaneous rotation centre of the caisson were obtained based on the displacement results. At the meanwhile, the measured soil pressures under the lid and on the outer skirt presented the soil-structure interaction of the caisson. Based on these results, a simplified force distribution for the failure mode of the caisson subjected to overturning moment was provided, which can be used for further analytical calculating method of ultimate moment capacity of the caisson. ACKNOWLEDGEMENTS The authors would like to acknowledge funds of National High-Tech R&D Program of China (863 Program) (research grant: 2007AA05Z427) and National

© 2011 by Taylor & Francis Group, LLC

Byrne, B. W. 2000. Investigations of suction caissons in dense sand. Ph.D thesis, University of Oxford, Oxford, UK. Byrne, B.W. & Houlsby, G. T. 2003. Foundations for offshore wind turbines. The Royal Society 361: 2909–2930. Villolobos. 2006. Model testing of offshore wind turbines Keble College, Michaelmas Term. Kelly, R. B., Houlsby, G.T. & Byrne, B.W. 2006. A comparison of field and laboratory tests of caisson foundations in sand and clay. Geotechnique 56, No.9: 617–626. Lehane, B. A., Schneider, J. A. & Xu, X. 2005. The UWA05 method for prediction of axial capacity of driven piles in sand. Frontiers in offshore geotechnics: ISFOG 2005, Taylor and Francis Group, London Houlsby, G. T., Kelly, R. B., Huxtable, J. & Byrne, B.W. 2006. Field trials of suction caissons in sand for off shore wind turbine foundations. Geotechnique 56, NO.1:3–10. Senders, M. & Randolph, M. F. 2009. CPT-Based method for the installation of suction caissons in sand. Journal of geotechnical and geoenviromental engineering 10.1061/ (ASCE) 1090-0241 135(1): 14–25. Chen, W & Randolph, M.F. 2007. Uplift capacity of suction caissons under sustained and cyclic loading in soft clay. Journal of geotechnical and geoenviromental engineering 10.1061/(ASCE) 1090-0241 133(11): 1352–1363. LeBlanc, C. 2004. Design of Offshore wind turbine support structures. Ph.D thesis, Technical University of Denmark, Denmark.

652

8 Jack-up units

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Simplified VH equations for foundation punch-through sand into clay J.-C. Ballard, P. Delvosal & P.H. Yonatan Fugro Engineers S.A. / N.V., Brussels, Belgium

A. Holeyman Université Catholique de Louvain, Belgium

S. Kay Fugro Engineers B.V., Leidschendam, The Netherlands

ABSTRACT: This paper gives simplified VH equations for surface strip foundation punch-through sand into uniform clay derived from 2D plane strain numerical simulations using Limitstate:GEO. The equations have been calibrated in the engineering ranges normally encountered in geotechnical engineering practice and have been inserted into proprietary computer program ISOBARE for everyday use. 1

INTRODUCTION

The assessment of bearing capacity of surface foundations on layered soils and subjected to combined VH loading is a common problem in the industry. For pure V loading, reliable analytical solutions are available. Of these, the “load spread” method (Young and Focht, 1981) is popular, in spite of the difficulty in selecting an appropriate “load spread factor”. For combined VH loading, reliable analytical solutions are scarce, especially for the most common offshore punch-through problem, sand over clay. Depending on the assumption made regarding horizontal load transfer, optimistic or pessimistic VH yield surfaces can result. 2 ANALYTICAL SOLUTIONS 2.1 Pure vertical loading Several analytical approaches are available to estimate the bearing capacity of sand overlying clay. Of these, the “load spread” method is frequently used. In this approach, it is assumed that the sand acts to spread the load and that the bearing capacity failure occurs within the clay. The load spread mechanism within the sand layer is modeled relatively simply by assuming that the vertical stresses associated with the footing load are confined to a zone defined by lines at angle β to the vertical, as shown on Figure 1. Load from the footing is assumed to be distributed uniformly over a width B
at the base of the sand layer, where B
= B + 2D tan β. For uniform clay, the bearing capacity V (or bearing pressure v) of the footing may be estimated using the following expression:

Figure 1. Load spread mechanism.

where su = clay undrained shear strength; Nc = standard bearing capacity factor for undrained loading; q = sand surcharge (= γD); and W = weight of sand confined in the load spread zone. It should be noted that the β angle implicitly accounts for any shear forces acting on the sides of the load spreading body. The problem with this method is that the chosen value of β can have an important influence on the calculated bearing capacity. A value of β between tan−1 (1/3) and tan−1 (1/5) is often adopted in practice (Kellezi, 2009) although it is generally accepted that the value of this parameter is influenced by the strength of the sand, the strength of the clay and the geometry of the problem. The use of these standard values may lead to results on the unsafe side in certain situations.

655 © 2011 by Taylor & Francis Group, LLC

2.2

Combined VH loading and yield surface

For combined VH loading, analytical solutions are scarce for the sand over clay case, which is commonly encountered offshore. Two extreme assumptions can be made. The first assumes that the bearing capacity of the clay layer is not influenced by the horizontal load, namely, the horizontal load is fully taken by the sand

uniform clay under pure V load and combined VH loading for a series of cases that cover the engineering ranges normally encountered in offshore geotechnical practice. Dimensionless groups were used to limit the number of analyses. The bearing pressure v may be shown to be given by the functional form (Michalowski et al, 1995):

where γ = sand unit weight; and φ
= sand friction angle. The VH curves were numerically derived for the following cases: – D/B = 0.25, 0.5, 0.75 and 1 – su /γD = 0.25, 0.5, 1, 2 and 4 – φ
= 25, 30, 35 and 40◦ Figure 2. Comparison between simplified assessments and finite element modeling for strip footing on sand over clay.

layer. The opposite one is to assume that the horizontal load is fully transferred to the clay layer, namely the sand layer does not take any horizontal load. Depending on the assumptions made, optimistic or pessimistic VH yield surfaces can result, as illustrated in the example presented in the next Section. 2.3

Reference problem

The engineering example that initiated this research is presented. The offshore spudcan foundation was circular with a 14 m diameter. Soil conditions consisted of a 8.5 m thick loose sand layer (γ
= 9.5 kN/m3 , φ
=29◦ ) overlying soft clay (su [kPa] = 18 + 1.6z [m], where z is the depth from top of clay layer). As a first approach, the foundation was treated as an infinite strip so that the analytical approaches discussed above could be directly compared with a 2D plane strain finite element simulation using Plaxis 2D v9 (Plaxis, 2002) and fully associated flow. The results of the comparison are shown on Figure 2. In this particular situation, a load spread angle β of tan−1 (1/3) is appropriate for pure vertical bearing capacity, which would be underestimated with β = tan−1 (1/5). With regard to the VH yield surface, it is either optimistic or extremely cautious depending on the assumption made with regard to horizontal load transfer (see Section 2.2). The conventional sand equations (Brinch-Hansen, 1970) are seen to be reasonable in the lower part of the VH yield surface (i.e. region where V < 200 kN/m), where sliding and general shear failure in sand dominate. 3

NUMERICAL MODELING

3.1 Introduction Numerical simulations were performed to determine the bearing capacity of strip footings on sand over © 2011 by Taylor & Francis Group, LLC

3.2

Software

Commercial programs Plaxis 2D v9 and Limitstate:GEO v2 (Limitstate, 2009) were used for these simulations. Limitstate:GEO (LSG) is very recent. It is designed to rapidly analyze the ultimate limit (or “collapse”) state for a wide variety of geotechnical problems. The soil failure model is rigid perfectly plastic (Tresca/Mohr-Colomb) with fully associated flow rule. The current version is limited to 2D plane strain analyses; extension to axisymmetry is under development and planned to be released shortly. LSG directly determines the ultimate limit state using the Discontinuity Layout Optimization (DLO) algorithm (Smith & Gilbert, 2007). DLO involves the use of rigorous mathematical optimization techniques to identify a critical layout of lines of discontinuity which form at failure. These lines of discontinuity are typically ‘slip-lines’ in planar geotechnical stability problems and define the boundaries between the moving rigid blocks of material that makeup the collapse mechanism. Associated with this mechanism is a collapse load factor, which will be an upper bound relative to the ‘exact’ load factor according to formal plasticity theory. Thus in essence the procedure replicates and automates traditional upper bound manual limit analysis. Solution accuracy can be improved by increasing the density of nodes covering the body under consideration, which in turn increases the number of discontinuities available for possible inclusion in the critical mechanism. 3.3

Comparison of results

The two programs were first evaluated by performing simulations in homogeneous isotropic soil conditions and comparing the results for vertical bearing capacity with classical analytical solutions. Highly accurate solutions have been obtained with both programs for bearing capacity in purely cohesive

656

Figure 3. Bearing capacity of strip footing on sand over clay, comparison of results. Note that x-axis is su /γD, not su /γB.

(Tresca) material. Bearing capacity in frictional media (with Mohr-Coulomb failure criteria) were found to be more challenging for both programs, although results within about 10%, compared to Martin’s exact solution (Martin, 2005), could be found, which is acceptable for this latter type of problem. Some of the cases published by Shiau et al (2003) were verified for the sand over clay problem. Shiau used finite element formulations of the limit analysis theorems to obtain rigorous plasticity solutions for the bearing capacity of a layer of sand on clay. As shown on Figure 3, both programs were able to reproduce reasonably well Shiau et al’s results.

non-associated flow rules, showing almost no difference in the results. However, these simulations were performed for small friction angles (φ
< 30◦ ). Loukidis et al (2009) recently published finite element simulations showing that the difference becomes more significant for higher friction angles (φ
> 30◦ ). They observed that the bearing capacity is 10–30% smaller (than when ψ = φ
) when realistic pairs of φ
-ψ values are assumed. The higher φ
, the higher the difference. The difference was found by the Authors of this paper to be less significant (5–15%) for the sand over clay case. This was expected as the failure mechanism in that case is only partly located in the sand layer. Errors of 5–15% tend to be masked by the fact that the bearing capacity factor Nγ in sand is highly sensitive to φ
value. Indeed, an error of 5–15% corresponds to assuming an angle of friction less than 1◦ smaller in an associated flow model. Therefore, the analyses were performed assuming associated flow.

3.4 Plaxis Vs. Limitstate:GEO As expected for surface foundations on sand, numerical problems were encountered with Plaxis. A phenomenon called “apparent strain softening” develops just next to the footing. As any softening phenomenon, the solution becomes very sensitive to mesh. Convergence was slow and “load-settlement” curves were irregular, the roughness increasing with increasing φ
. Our experience is that the problem is aggravated by mesh refinement while unrealistic failure modes emerge next to the footing. LSG was more robust and reliable. Results converged steadily towards analytical solutions by increasing the density of nodes. Solutions within 10% tolerance could generally be obtained within a few minutes compared to several hours with Plaxis. Therefore, LSG was found to be the most appropriate tool for our ultimate load problem. 3.5 Influence of non-associativity An advantage of the finite element method is that the influence of non-associativity in sand may be investigated while fully associated flow is assumed in LSG. It is generally believed that the assumption of fully associated flow rule in sand is reasonable for unconfined problems such as the bearing capacity of shallow foundations. Zienkiewicz et al (1975) published numerical simulations, assuming fully associated and © 2011 by Taylor & Francis Group, LLC

657

4

RESULTS AND DISCUSSION

4.1 Vertical bearing capacity LSG results for the vertical bearing capacity are given in Figure 4. In this paper, typical results for the D/B ratios of 0.5 and 1 are shown. Although no exact solutions are available, it is our opinion that, based on work of others and LSG cases where model geometry and nodal refinement were optimized, these results may in certain cases slightly overestimate the exact solution (up to about 10%). As expected, the normalized vertical bearing capacity increases with increasing su /γD, φ
and D/B. As the undrained shear strength of the clay increases, the curves flatten out for small friction angles and the failure mechanism becomes concentrated in the sand layer. Typical failure mechanisms observed are illustrated on Figure 5 for two specific cases. The first case is a dense sand layer overlying soft clay. The assumption of the load spread mechanism seems appropriate in this case. On the other hand, for the case of loose sand on stiffer clay, the assumption of a load spread mechanism with traditionally used load spread angles would lead to a severe overestimation of bearing capacity. The results from the parametric study are plotted in an alternative way in Figure 6 in order to illustrate the load spread mechanisms within the sand layer. In this plot, the parameter β is back-calculated using the standard expressions presented in Section 2. This load spread model is based on a highly simplified view of the mechanics of the system, hence particular β values do not have a precise physical interpretation. The load spread model becomes inappropriate when failure is confined to the sand layer, and so Figure 6 presents only results where failure occurred in sand and clay. Figure 6 also includes results for D/B ratios of 0.25 and 0.75. The results show that β is (1) remarkably insensitive to the value of D/B, (2) even more remarkably

Figure 5. Limitstate:GEO failure mechanisms for D/B = 0.5 (a) dense sand over soft clay: su /γD = 0.25, φ
= 40◦ and (b) loose sand over stiff clay: su /γD = 2, φ
= 30◦ .

Figure 4. Limitstate:GEO results, variation of v/γB with su /γD for (a) D/B = 0.5 and (b) D/B = 1.

varying linearly with the logarithm of su /γD and (3) varying approximately linearly with friction angle φ
for a given su /γD. The equivalent load spread angle β decreases with decreasing sand friction angle and increasing undrained shear strength ratio su /γD. For sand over normally consolidated clay (su /γD∼ 0.25), the values traditionally used for β (i.e. 11 and 18◦ ) are cautious. However, they may become unsafe when the clay undrained shear strength increases: equivalent β values can even become negative. Figure 6. Variation of equivalent β with su /γD for D/B = 0.25, 0.5, 0.75 and 1.

4.2 VH yield surfaces The computed VH yield surfaces are presented on Figure 7 for one D/B ratio (0.5) and one sand friction angle (35◦ ). The upper part of the yield surfaces are clearly curved so that the assumption of a flat top part (see Figure 2) is unconservative. A typical failure mechanism for inclined loading is shown on Figure 8. 5 5.1

SIMPLIFIED EQUATIONS Introduction

Simplified equations derived from the 2D plane strain numerical results presented above are proposed. These © 2011 by Taylor & Francis Group, LLC

equations can be used to compute VH yield surfaces of surface strip foundations on sand over uniform clay. 5.2

Pure vertical loading

When assessing the bearing capacity of shallow foundations on sand over clay, the results presented in section 4 showed that the load spread angle need not be an input variable. A relationship that computes automatically the appropriate load spread angle can easily be found: β is essentially independent of D/B and linearly varying with the logarithm of su /γD, and

658

Figure 9. Load transfer mechanism under inclined load.

As can be seen by comparing Fig. 9 with Fig. 1, the following assumptions have been made:

Figure 7. Limitstate:GEO results, VH yield surfaces for strip footing (B = 2 m, D/B = 0.5 and φ = ψ = 35◦ ).

– the LSB centerline, which has the same inclination as the VH load vector, defines the center point C of an equivalent footing on the underlying clay – the spreading geometry conforms to eq. (3), keeping the equivalent footing width B
resting on top of clay, i.e. B
= B + 2D tanβ – passive resistance mobilized along the LSB sides is schematized by a horizontal reaction Pm acting at height ηP D above the clay. The horizontal load HD transferred to the equivalent footing is obtained by horizontal equilibrium HD = H − Pm while the moment MD about point C applied to the lower equivalent footing is modeled by the following expression:

Figure 8. Limitstate:GEO’s example of failure mechanism under VH load (H = 0.3V, D/B = 0.5, su /γD = 0.25, φ
= 35◦ ).

with friction angle for a given su /γD. The following relationship fits the data reasonably well:

A priori assessments of ηH , ηP , and ηW can be sought by considering the curvature of the funicular thrust line throughout the LSB, a triangular distribution of passive pressures (ηP = 1/3), and the geometric determination of the centre of mass of the LSB. The mobilization of Pm follows an initial proportion α.π of the applied H according to:

where β and φ are both in degrees.

5.3 VH yield surface An attempt has been made to quantify the influence of the horizontal component of the foundation load as a further development of the load spread concept given in Fig. 1. Figure 9 highlights the horizontal action and reactions on the load spreading sand body (LSB) transferring part of the horizontal load onto its lower potential sliding surface. The need for partial transfer can be inferred from the examination of the intermediate position of the upper portion of the reference (Plaxis) failure envelope in Fig. 2 with respect to either total or zero transfer of horizontal load to the equivalent footing. © 2011 by Taylor & Francis Group, LLC

The initial rate of Pm mobilization is thus commensurate with the ultimate values of the potentially resisting terms, modulated by the parameter α. One intuitively expects passive resistance to be mobilized preferentially at low H since the base shear is acting at a further distance from the footing than the lateral boundaries of the LSB, especially for large D/B values. In addition, as H increases, displacement compatibility penalizes Pm mobilization that requires a larger displacement than base shear. An analogy is the better known problem of end bearing and friction resistance mobilization of a vertically loaded pile. Based on the above comparison, and allowing for the maximum Pm value to be lower than the reference ultimate

659

clay. It is fast and reliable and permitted about 800 analyses in a rigorous and consistent manner. Based on the results of these numerical simulations, simplified equations are proposed to compute the equivalent load spread angle β and the upper part of theVH yield surface for surface strip footing on sand over uniform clay. These equations have been calibrated in the engineering ranges normally encountered in geotechnical engineering practice. Given the encouraging results from this first step study assuming infinite strips, the method is being calibrated for other foundation shapes (circles and rectangles) and soil su profiles. REFERENCES

Figure 10. Comparison between analytical equation and Limitstate:GEO results for strip footing (B = 2 m, D/B = 0.5 and φ = ψ = 35◦ ).

value Pmax = 0.5απH ≤ Pult , Pm is suggested to be mobilized according to the following equation:

A typical comparison between the proposed simplified approach and LSG calculations is presented in Fig. 10 for the case D/B = 0.5 and φ
= 35◦ . The curves for the different values of su /γD = 0.25, 0.5, 1 and 2 have all been obtained using the same parameters: α = 1.1, ηP and ηW = 1/3, and ηH = 0.2. It can be observed that agreement is satisfactory with the upper part of the calculated failure envelope, which was the primary target of the developed approach, i.e. modeling the punch-through mechanism. Equations presented in this Section have been inserted into proprietary analytical “load spread” method software ISOBARE (Fugro, 2009) for everyday use. This program calculates shallow foundation capacity under general VHM loading on layered soils. Bearing capacity under pure V loading is calculated using the load spread angle deduced from Equation 3. The upper part of the VH yield surface is computed using Equations 4, 5 and 6 while the lower part, where sliding and shear failure in sand dominate, is computed using conventional sand equations (Brinch-Hansen, 1970). 6

Brinch Hansen, J. 1970. A revised and extended formula for bearing capacity. The Danish Geotechnical Institute Bulletin (28): 5–11. Fugro. 2009. ISOBARE, computer program for shallow foundation capacity on layered soils. Kellezi, L. & Kudsk, G. 2009. Spudcan penetration FE simulation of punch-through for sand over clay. In 12th Jack-Up Platform Conference, London, September 2009. Limitstate. 2009. Geotechnical software for stability analysis, Version 2. Sheffield: Limitstate Ltd. Loukidis, D. & Salgado, R. 2009. Bearing capacity of strip and circular footings in sand using finite elements. Computers and Geotechnics 36(5): 871–879. Martin, C.M. 2005. Exact bearing capacity calculations using the method of characteristics. In B. Giovanni & M. Barla (eds.), Proceedings of the Eleventh International Conference on Computer Methods and Advances in Geomechanics, Torino, Italy, 19–24 June 2005, Vol. 4: 441–450. Bologna: Pàtron Editore. Michalowski, R.L. & Shi, L. 1995. Bearing capacity of footings over two-layer foundation soils. Journal of Geotechnical Engineering 121(5): 421–428. Plaxis. 2002. Finite element code for soil and rock analyses, Version 9. Delft: Plaxis BV. Shiau, J.S., Lyamin, A.V. & Sloan, S.W. 2003. Bearing capacity of a sand layer on clay by finite element analysis. Canadian Geotechnical Journal 40(5): 900–915. Smith, C. & Gilbert, M. 2007. Application of discontinuity layout optimization to plane plasticity problems. Proceedings of the Royal Society of London; A 463(2086): 2461–2484. Young, A.G. & Focht (Jr.), J.A. 1981. Subsurface hazards affect mobile jack-up rig operations. Soundings (Summer). Zienkiewicz, O.C., Humpheson, C. & Lewis, R.W. 1975. Associated and non-associated visco-plasticity and plasticity in soil mechanics. Géotechnique 25(4): 671–689.

DISCUSSION AND CONCLUSIONS

Limitstate:GEO is an appropriate tool to derive VH yield surfaces for surface strip footing on sand over

© 2011 by Taylor & Francis Group, LLC

660

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Characterisation of undrained shear strength using statistical methods B. Bienen, M.J. Cassidy & M.F. Randolph Centre for Offshore Foundation Systems, University of Western Australia, Perth, Australia

K.L. Teh Centre for Offshore Research and Engineering, National University of Singapore, Singapore

ABSTRACT: Interpretation of shear strength measurements of a soil profile represents a crucial step before a bearing capacity analysis can be performed. In the case of mobile offshore drilling rigs, the installation process implies continuous bearing capacity failure. The final footing embedment is controlled by the depth at which sufficient soil resistance is reached. As this requires the expected bearing capacity with depth to be predicted accurately, uncertainty and conservatism in the soil shear strength profile need to be minimised. This paper uses a statistical approach to examine the undrained shear strengths derived from a variety of tools and to derive strength profiles for bearing capacity prediction. The main points are illustrated through examples of clay sites from the InSafeJIP data base.

1

INTRODUCTION

Jack-up drilling rigs are frequently re-located as they are typically employed for short periods of work-over of an existing well-head platform, for instance. For each installation process a prediction of the anticipated load-penetration response is required. The quality of the prediction is largely determined by the choice of the design shear strength profile for the bearing capacity analysis. Accurate prediction of the ‘spudcan’footing’s entire load-penetration behaviour, not only the final embedment depth, is crucial to the safety of the installation process. Comparison of the actual response with the predicted curve is a useful check of the understanding of the penetration behaviour, and significant deviation of the two curves should alert the personnel to potential problems (Houlsby 2010). The hazards to jack-up installations are numerous. In soil profiles where a strong stratum overlies a weaker layer, rapid uncontrolled leg penetration or ‘punch-through’ can pose a significant risk to the jackup installation. The potential consequences include lost time and revenue, structural damage but may be as severe as capsize and loss of the platform (Osborne et al. 2006). On soft soil sites, large penetrations (of the order of 30 m or two to three footing diameters) are common, such that the available leg length of the rig can become a limiting factor. Accurate prediction of the final penetration depth is therefore particularly important in this situation. As the jack-up industry expands from the Gulf of Mexico and the North Sea, with predominantly (single layer) clay or sand soil profiles, into South East Asia and Australia, where the soil profiles are

© 2011 by Taylor & Francis Group, LLC

661

multi-layered and often include silty deposits, these risks magnify (Osborne et al. 2006). Not only is the body of experience of site investigation, interpretation and jack-up installation significantly smaller in those “new” regions, also the difficulty of installation is much increased. Challenges in site investigation and measurement interpretation in these soil provinces include: • Identification of layer boundaries, • Large scatter of shear strength measurements

obtained through different methods, • Rate effects on the soil characterisation measure-

ment, and • Identification and understanding of rate effects

(drainage) when interpreting the shear strength measurements. As the interpretation of site investigation data and the derivation of a design shear strength profile for use in bearing capacity analysis have traditionally relied heavily on experience, there is potential risk associated with extrapolation of such experience from one soil type to another. Indeed, the subjectivity of the choice of a design shear strength profile was highlighted in an exercise within the InSafeJIP (Osborne et al. 2008, 2009a). The site investigation results of two jack-up locations were provided to a number of people. Their interpretation varied widely, as did the suggested design shear strength profiles. This showed the need for a more objective evaluation of each of the soil strength characterisation methods used as well as the process of deriving shear strength profiles for use in load-penetration predictions for jack-up rigs. Statistical methods provide a rational, consistent and objective approach to the derivation of shear

strength profiles that removes much subjectivity. An overview of related work is provided by Christian (2004). This paper examines the undrained shear strengths derived from a variety of strength measurement methods commonly conducted for jack-up site investigation. The discussion of example sites illustrates not only the direct benefits of application of the statistical approach to strength profile derivation but also shows the potential for longer-term refinement that will enhance the present understanding. 2 2.1

The definition of best fit that underlies Equations 2 and 3 is a mean difference of zero between the calculated and measured shear strength. Calculation of the standard deviation

STATISTICAL APPROACH

and coefficient of variation (CoV)

Deriving mean undrained shear strength profile for a site

The use of statistical methods to obtain shear strength profiles has been advocated by Lacasse et al. (2007), amongst others, and is detailed in DNV (2007). The treatment below (Equations (2) to (7)) follows that recommended in the latter publication. Within the InSafeJIP, the reasoning for employing statistics for the purpose of deriving undrained shear strength profiles was twofold:

provides an indication of the variation within a statistical sample (here: group of shear strength measurement points). These calculations were performed for each shear strength characterisation method in each layer. No data points were removed (possible “outliers”) for the statistical treatment as this would have introduced subjectivity. In addition, the overall mean shear strength profile was calculated, placing equal weighting on each characterisation method (rather than on each data point, which would have favored TV and PP measurements with numerous data points over UU results with only few data points, for instance). The statistical outcomes of a characterisation method relative to the average of all methods (calculated at the corresponding depth) used in that layer can be expressed as the bias:

• The primary aim was to evaluate the performance of

each the different soil shear strength measurement methods as well as their performance relative to each other. • The second aim was to ensure as much objectivity in the process as possible in order to arrive at design undrained shear strength profiles that could be used to benchmark the various methods of bearing capacity predictions against each other. For the statistical treatment of the shear strength measurement data, each soil layer with N data points is assumed to be continuous while individual layers are independent of other soil strata. Shear strength is often not constant but increases with depth. Here, a linear variation with depth was assumed. The profile can then be described by Equation 1:

with a0 being the strength intercept at the mudline and a1 the shear strength gradient with depth from the seabed z. The parameters a0 and a1 can be evaluated from Equations 2 and 3:

Thus, a bias of 0.1, for instance, indicates a shear strength profile based on one characterisation method that is on average 10% higher than the mean profile. Meaningful confidence bands to the design shear strength profile can be provided through the use of proportions of the standard deviation (Equation 6; Lacasse et al. 2007). For instance, framing the design shear strength profile with a profile of half a standard deviation above and below the mean attaches meaning to what is often referred to as “upper” and “lower bounds”. 2.2

Refinement within a data base

The statistical results of a data base can be used for further refinement in several ways: where the mean values of depth, z¯ , and undrained shear strength, s¯u , are calculated as:

© 2011 by Taylor & Francis Group, LLC

662

• Directly through deriving “conversion factors”

between shear strength characterisation methods, and • Through “closing the loop” with bearing capacity predictions and actual recorded load-penetration predictions.

The former are calculated taking into account the absence of some shear strength characterisation methods in some layers and can therefore not be directly derived from the bias values. This approach is similar to the practice of applying “weighting factors” which are based on experience of the performance of the soil characterisation method in local soils. Though this is understood to be routinely performed in jack-up assessments today (based on anecdotal evidence), this practice likely is qualitative and relies on individual experience and judgment. The use of statistics over as large a data base as possible provides a more rigorous approach. The data base can further be refined if the predicted and measured load-penetration curves of each rig move are compared as this allows conclusions of the performance of the prediction and its input to be drawn. Note, however, that this comparison assesses the combined performance of the shear strength profile interpretation and the bearing capacity prediction method. It is not possible, therefore, to attribute the successful/failed prediction to “correct” interpretation of the shear strength profile that was used in “accurate” bearing capacity prediction as cancellation of under-prediction of one and over-prediction of the other yields the same result. 3

Figure 1. Shear strength data and mean undrained shear strength profiles.

RESULTS AND DISCUSSION

characterisation method and (ii) different trends with depth, which is mirrored in the derived individual profiles. Note that although the “design” undrained shear strength profile is derived in an objective manner, some subjectivity remains attached to the process through the choice of layer boundaries. Further, the mean undrained shear strength profile as shown in Figure 1 was derived from equally weighted UU, MV and TV data, each of which exhibit different undrained shear strength measurement values and a different trend with depth (or increasing shear strength) in the same material. It can be argued that a different mean undrained shear strength profile would be obtained had (a) different (mix of) methods been employed during the site investigation (e.g. more or less different methods or PP instead of MV). The soil characterisation method(s) used will influence the derived shear strength profile as the soil’s strength depends on the mode of shearing. There is not one single or ‘true’ undrained shear strength. Further, natural soils are rate-dependent materials, implying that the shear strength varies with the shearing velocity. A multitude of methods exist to determine soil shear strength, differing in the type of shearing and rate with which the soil is brought to failure. Consequently, different shear strengths are obtained with each of these methods.

3.1 Soil characterisation methods discussed In the jack-up industry, the commonly used method for characterisation of undrained shear strength of clay are torvane (TV), miniature vane (MV), motor vane (MotV), undrained unconsolidated (UU) triaxial test, pocket penetrometer (PP) and piezocone (CPTU). Note that a distinction is made here between MV and MotV in which the former test is assumed to be manually operated while the latter is motor driven. In comparison, high quality laboratory tests such as anisotropically consolidated undrained compression/extension (CAUc/CAUe) triaxial and direct simple shear (DSS) tests are seldom carried out based on the InSafeJIP database. 3.2

Mean undrained shear strength profiles

In the context of the InSafeJIP (Osborne et al. 2008, 2009a) 14 clay sites in total were statistically analysed (Osborne et al. 2009b). Figure 1 shows the results of the statistical analysis of an example site. Based on the shear strength data, the borehole log and index test results the stratigraphy was assumed to consist of two layers, with the layer boundary at 13 m depth. The Figure shows the data points by method as well as the statistically derived undrained shear strength profile corresponding to each of the methods. It further contains the overall mean shear strength profile, derived given equal weight to each of the methods. The data points in Figure 1 show (i) different amounts of scatter depending on the shear strength © 2011 by Taylor & Francis Group, LLC

3.3

Information from CoV and bias

While in the statistical analysis each soil layer was assumed to be homogenous, the properties of a natural soil vary even within a layer. This natural variation is expected to be reflected in the shear strength

663

Table 1.

Comparison of CoV (InSafeJIP) with literature.

Method

CoV (InSafeJIP) Mean (Range∗ ) %

CoV Mean (Range) %

UU MV MotV TV PP CPTU

20 (3–37) 20 (0–48) 12 (3–18) 13 (2–31) 19 (4-44) 23 (2–41)

33 (11–49)1 , (10–30)2 N/A N/A N/A 20–403 27 (5–40)1 , 144

∗

Figure 2. CoV of different shear strength characterisation methods.

measurements taken. Further, the data will be influenced by measurement errors. These cannot be separated here from the natural variation. However, the latter is expected to be dominant. Figure 2 shows this variation within the InSafeJIP data base, expressed as the coefficient of variation (Equation 7). Each data point represents the statistical treatment of one shear strength characterisation method in one soil layer. Note that the 14 sites analysed within the scope of the InSafeJIP, though all classified as clay soils, span various regions worldwide and have been obtained from different operators, possibly adopting different testing procedures. The coefficients of variation shown in Figure 2 generally decrease with depth (or increasing shear strength). This trend as well as the absolute values of CoV are consistent with the literature (Table 1). The results shown in Figure 2 can further be used to identify “outliers” within the data base, points that do not conform with the general trend. Here labeled (a), further scrutiny of these points identified silt pockets and thin cemented layers as the likely reason for the increased variation. Somewhat counter-intuitively, therefore, the high values of CoV obtained for the CPTU demonstrate this method’s superior profiling capabilities. Further, though it is acknowledged that different Nkt factors apply to different soils, depending on anisotropy, rigidity index (Teh & Houlsby 1991), strain softening and rate effects, a single constant value of Nkt = 13.5 (from a study of soft clay sites worldwide, Low et al. 2010) was used in the InSafeJIP study to interpret the piezocone data. This was required as no high quality laboratory shear strength data (CAUc, DSS, CAUe) were available to “tie back” the in situ piezocone tip resistance. Utilising data of another method, such as the UU tests, would have rendered the two methods dependent, which would have precluded their use in the statistical analysis. A further point, which applies to all methods of in situ and laboratory strength measurement, concerns deposits that have significant structure, such as fissuring, at a larger spacing than the measurement device. In such cases CPTU data and laboratory strength measurements will tend to overestimate the soil mass strength and spudcan penetration resistance. © 2011 by Taylor & Francis Group, LLC

664

Excluding values thought to be influenced by silt or cemented seams. 1 Phoon & Kulhawy (1999). 2 Lacasse & Nadim (1996). 3 OSHA (1999). 4 Low et al. (2010).

Figure 3. Bias of different shear strength characterisation methods relative to the mean of all methds used in that soil layer.

Methods with high CoV attributable to measurement errors (as opposed to cemented seams, for instance) are clearly undesirable as the intrinsic low accuracy of the obtained data limits the accuracy of the spudcan load-penetration prediction. The bias values derived from the 14 statistically analysed cases of the InSafeJIP data base are shown in Figure 3. The PP and UU generally provide undrained shear strengths that lie about 20% below the average of all methods at low shear strengths. The negative bias of two methods tends to weaken with increasing shear strength.The trend observed for the MotV opposes that of the PP and the UU test in the InSafeJIP data base, overestimating the average by about 25% at low shear strengths. The poor performance of the CPTU, which continued to overestimate the mean, is attributed to the choice of the non-site specific Nkt factor. Note that the bias indicates the difference between a shear strength characterisation method relative to all the method used in this soil layer, not relative to an absolute, correct undrained shear strength. Intrinsic in the calculated bias value is further the mode of shearing compared to the average of all methods used. 3.4

Refinement within the InSafeJIP data base

Conversion factors were derived from the statistical analysis performed within the InSafeJIP data base as

Table 2.

Conversion factors.

Method

Factor

UU MV MotV TV PP CPTU

1.00 0.73 0.67 0.90 1.12 0.72

strongly with depth than evident from the factored TV and MV data. The site investigation data should be scrutinised for an explanation of the differing trends. If this method is employed in practice, not only will the data base continuously be added to, but the conversion factors will be refined in the process. This will allow the development of regional (or soil-specific) conversion factors as well as easier detection of “outlying” data points. Note that these should be scrutinised instead of simply being discarded as erroneous as they may identify a thin layer or drainage having taken place during the measurement due to the presence of silt pockets. While the refined derivation of the undrained shear strength profile for the example site in this paper relates all shear strength measurements to the UU test, this may not be shear strength characteristic of spudcan penetration. The average of CAUc, DSS and CAUe is considered appropriate for assessing spudcan penetration (Osborne et al., 2009b). This is the same strength measurement that is recommended for evaluating sitespecific piezocone Nkt factors (Randolph 2004, Lunne et al. 2005). 3.5

Figure 4. Converted shear strength data and mean undrained shear strength profiles.

discussed in Section 2.2. These factors are provided in Table 2 for the shear strength band of interest to spudcan installation, a range of su from 35 to 80 kPa (calculated as the average range of preload pressure divided by an approximate spudcan bearing capacity factor). The conversion factors were used in the example given in this paper to convert the MV and TV shear strength measurements into equivalent UU strengths. The factored data, their trends as well as the mean profile, are shown in Figure 4. Also included in the Figure is the mean undrained shear strength profile as derived from the initial statistical analysis (see Fig. 1). As expected, the data points obtained by the different methods now fall into a narrower band. As a result, the profiles with depth also fall in a narrow range. However, the scatter between data points obtained by the same method remains unaffected. The trend of the UU data points suggests the soil strength increasing more

© 2011 by Taylor & Francis Group, LLC

Ideal way forward

As discussed above, the statistical approach is a powerful tool that removes much of the subjectivity associated with the choice of design shear strength profiles. By continuously adding to the data base, the statistical results become more robust and the deduced conversion factors more reliable. As the piezocone measurements should be related to the average of CAUc, DSS and CAUe data due to their relevance to spudcan penetration, conversion factors of other methods of shear strength characterisation to the piezocone should be developed. Care should, of course, be taken with all methods to obtain high quality shear strength measurements for which the drainage regime is known with certainty. The equivalent shear strength data thus obtained are related to the characteristic strength sought. Proportions of the standard deviation can be used to create meaningful confidence bands, as mentioned above. Utilising characteristics shear strength profiles as input, the current bearing capacity methods can then be scrutinised regarding their predictive performance when compared with the actual measured spudcan load-penetration profiles offshore. This, of course, requires the premise that continuous measurements are recorded during the jack-up rig installation. One or two data points do not allow for a proper comparison with the predicted behaviour. While it may be tempting to apply conservatism by converting all site investigation data to what is perceived to be the most conservative method and choosing a design undrained shear strength profile that is conservative compared to these data, this “lower bound” merely masks several areas of uncertainty. A better approach would be the derivation of an appropriate design shear strength profile (with meaningful confidence bands) as outlined above.

665

The approach described, and the values derived from the InSafeJIP database, provide valuable statistical information to be used in probabilistic assessments of the prediction of spudcan penetrations. Such a calculation method has been proposed in Houlsby (2010). 4

CONCLUDING REMARKS

The paper employed statistics as a useful approach not only to investigate the characteristics of site investigation data measured by different methods, but also to the derivation of “design” undrained shear strength profiles for foundation analysis in the context of jack-up drilling rigs. The methodology and possible refinements were discussed using the InSafeJIP data base as an example. ACKNOWLEDGEMENTS The work presented here was carried out as part of the InSafeJIP (see http://insafe.woking.rpsplc.co.uk) that aims at the development of “Improved Guidelines for the Prediction of Geotechnical Performance of Spudcan Foundations during Installation and Removal of Jack-up Units”. The support and permission to publish are gratefully acknowledged. However, the methods described here do not form part of the formal recommendations. The authors appreciate the fruitful discussion with the project team, especially from Julian Osborne (RPSE) and Professor Guy Houlsby (Oxford University). REFERENCES Christian, J.T. 2004. Geotechnical engineering reliability: How well do we know what we are doing? J. Geotechnical and Geoenvironmental Eng. (ASCE), 130(10), 985–1003. Det Norske Veritas (DNV). 2007. Statistical representation of soil data. Recommended practice, DNV-RP-C207. Houlsby, G.T. 2010. A probabilistic approach to the prediction of spudcan penetration of jack-up units. Proc. 2nd

© 2011 by Taylor & Francis Group, LLC

Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth. Lacasse, S., Gutteromsen, T., Nadim, F., Rahim, A., Lunne, T. 2007. Use of statistical methods for selecting design soil parameters. Proc. 6th Int. Offshore Site Investigation and Geotechnics Conference: Confronting new challenges and sharing knowledge, London, UK, 449–460. Lacasse, S. & Nadim, F. 1996. Uncertainties in characterising soil properties. Plenary paper for ASCE Conference “Uncertainties ’96”, Madison, Wisconsin, USA. Low, H.E., Lunne,T.,Andersen, K.H., Sjursen, M.A., Li, X. & Randolph, M.F. 2010. Estimation of intact and remoulded undrained shear strength from penetration tests in soft clays. Géotechnique, accepted. Lunne, T., Randolph, M.F., Chung, S.F., Andersen, K.H. & Sjursen, M. 2005. Comparison of cone and T-bar factors in two onshore and one offshore clay sediments. Proc. Int. Symp. on Frontiers in Offshore Geotechnics: ISFOG, Perth, 981–989. Osborne, J.J., Pelley, D., Nelson, C. & Hunt, R. 2006. Unpredicted jack-up foundation performance. Proc. Jack-Up Asia Conference and Exhibition, PetroMin, Singapore. Osborne, J.J., Teh, K.L., Leung, C.F, Cassidy, M.J., Houlsby, G.T., Chan, N., Devoy, D., Handidjaja, P., Wong, P. & Foo, K.S. 2008. An introduction to the InSafeJIP. Proc. 2nd Jack-Up Asia Conference and Exhibition, PetroMin, Singapore. Osborne, J.J., Houlsby, G.T., Teh, K.L., Leung, C.F., Bienen, B., Cassidy, M.J. & Randolph, M.F. 2009a. Improved guidelines for the prediction of geotechnical performance of spudcan foundations during installation and removal of jack-up units. Offshore Technology Conference (OTC), Houston, Texas, OTC 20291. Osborne, J.J., Houlsby, G.T., Cassidy, M.J., Randolph, M.F., Leung, C.F., Teh, K.L., Bienen, B. & Hossain, M.S. 2009b. InSafeJIP, First Year Progress Report. RPS Energy. OSHA 1999. Excavations: hazard recognition in trenching and shoring. OSHA Technical Manual, 5th Edition, U.S. Department of Labor, Occupational Safety and Health Administration. Phoon, K.K. & Kulhawy, F.W. 1999. Characterisation of geotechnical variability. Canadian Geotechnical Journal, 36, 612–624. Randolph, M.F. 2004. Characterisation of soft sediments for offshore applications. Proc. 2nd International Site Characterisation Conference, Porto, Portugal, 1, 209–232. Teh, C.I. & Houlsby, G.T. 1991. An analytical study of the cone penetration test in clay. Géotechnique, 41(1), 17–34.

666

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Centrifuge modelling of spudcan deep penetration in multi-layered soils M.S. Hossain, M.F. Randolph & Y.N. Saunier Centre for Offshore Foundation Systems, The University of Western Australia, Perth, Australia

ABSTRACT: This paper reports centrifuge modelling of spudcan foundations penetrating through multilayered soils with interbedded stiff clay layers. The model tests were carried out at 200 g in a drum centrifuge. Spudcan models of three different sizes, 30, 40 and 60 mm diameter, were used in order to achieve different ratios of the stiff layer thickness to the spudcan diameter for the prepared layered profiles. Test specimens were prepared, mostly mimicking strength profiles from reported case histories particularly those where punch-through failure occurred, with similar normalised layer soil properties and geometry. A range of the stronger layer thickness, relative to the spudcan diameter, of 0.1 to 0.6 was explored. Punch-through and rapid leg penetration (for strong-over-weak) and squeezing (for the reverse) were demonstrated by the penetration resistance profiles and formed the basis for developing design calculations. The results were compared directly with field records of spudcan load-penetration response to evaluate their accuracy in predicting the shape of the spudcan resistance profile and final penetration depths, with excellent agreement obtained. Critical factor for developing a direct correlation between penetrometer and spudcan was discussed. 1

INTRODUCTION

1.1 Evolution of jack-up rig operations and in situ seabed profiles Most offshore drilling in shallow to moderate water depths throughout the world is performed from independent three leg jack-up rigs with proven flexibility, mobility, and cost-effectiveness. Depletion of known reserves in the traditional regions and in shallow waters, is resulting in exploration in deeper, unexplored and undeveloped environments with more complex seabed soil conditions. In emerging provinces

Figure 1. Multi-layered seabed profile susceptible to punch-through (Raya B, offshore Malaysia; see Kostelnik et al. 2007).

© 2011 by Taylor & Francis Group, LLC

667

and fields, highly layered soils are prevalent, such as at the Sunda Shelf (see Fig. 1), Southeast Asia, Bass Strait (SE of Australia) or offshore India. Layered soil profiles result from various geological processes, including previous crustal desiccation, sand channelling and evolving depositional environments associated with changing sea level (Castleberry II & Prebaharan 1985). When a strong layer (sand or stiff clay) is underlain by a soft soil layer (commonly soft clay), the foundation may experience a peak resistance during penetration through the strong layer (Fig. 2). After this peak resistance is reached, a sudden penetration

Figure 2. Punch-through during jack-up preloading at Belida B (see Brennan et al. 2006).

may occur without any additional loading. This sudden penetration is termed ‘punch-through’ failure. 1.2

Punch-through

Installing and preloading a mobile offshore drilling (jack-up) rig in stratified deposits, where an interbedded strong layer overlays a weak layer (Fig. 1), remains a challenge for the offshore industry, with the potential for severe ‘punch-through’ failure under the loadcontrolled conditions. Uncontrolled rapid leg penetration may lead to buckling of the leg, effectively decommissioning the platform, or may even result in toppling of the unit (Brennan et al. 2006, Kostelnik et al. 2007). Although the potential hazard of strong crustal features is well documented (SNAME 2008), jack-up rigs frequently undergo punch-through failure around the world (Jack et al. 2007). Two recent examples are the failures of the Noble David Tinsley, with reported severe damage to the jack-up and its three legs, in May 2009 off the coast of Qatar, and the Sapphire Driller, with reported minor damage to the bracings of one leg, in October 2009 off Ivory Coast (see Fig. 3). The more frequent use of jack-ups in geological conditions where a clay crust is underlain by a soft clay layer, such as on the Sunda Shelf in Southeast Asia (Castleberry II & Prebaharan 1985, Paisley & Chan 2006), and the increasing annual number of rig moves, has increased the rate of punch-through incidents. In recent years, rigs have suffered 1 or 2 major punch-through events per annum, together with about 10 to 12 unexpected rapid footing penetrations (Paisley & Chan 2006). 1.3

Previous work and objective of present study

A number of investigations have been carried out on foundation response in two layer soils through experimental work and finite element analyses, such as on clay over clay (Brown & Meyerhof 1969, Wang & Carter 2002, Hossain & Randolph 2010) and on sand over clay (Lee et al. 2009). To date, no research has been published that investigates penetration of spudcan foundations in multi-layered sediments and no technique has been developed for effective elimination of punch-through hazards in problematic deposits. This paper reports results from an experimental pilot study to investigate the vertical penetration of spudcan foundations in general multi-layered soils, giving penetration resistance profiles that demonstrate punch-through and rapid leg penetration (for strongover-weak strata) and squeezing (for the reverse). 2 2.1

CENTRIFUGE MODELLING Experimental program

The experimental program comprised centrifuge modelling of spudcan foundations penetrating through three- or four-layered soils with stiff clay or sand layers interbedded in soft clay. The work was carried out at 200 g in the drum centrifuge at the

© 2011 by Taylor & Francis Group, LLC

668

Figure 3. Jack-up rigs after punch-through failure: (a) Noble David Tinsley: offshore Qatar, May 2009, (b) Sapphire Driller: offshore Ivory Coast, October 2009.

Figure 4. Preparation of test specimen for NC clay with an interbedded stiff layer.

Figure 5. Centrifuge spudcan models.

University of Western Australia, which has a 1.2 m diameter annular channel and a maximum acceleration level of 485 g. The soil was confined within a purpose designed strongbox to facilitate producing multi-layered specimens, with the box mounted within the drum channel (see Fig. 4). Strongboxes of two different sizes were used, with internal sizes of 258 (length) × 163 (width) × 160 (depth) mm, and 258 × 80 (half width) × 160 mm. Spudcan penetration tests were performed to measure the load-penetration response using three fullspudcans of 60, 40 and 30 mm diameter (Fig. 5). Spudcan models of three different sizes were used in order to achieve different ratios of the stiff layer thickness to the spudcan diameter, t/D, for the prepared

Table 1.

Summary of centrifuge tests reported.

Test

G-level

D (m)

t1 (m)

t2 (m)

t3 (m)

t4 (m)

I II

200 g

12 8

4.5 5.2

6.9 5

14.6 5.7

– 10.1

layered profiles. The models were made from duraluminium and included a 13◦ shallow conical underside profile (included angle of 154◦ ) and a 76◦ protruding spigot. The shape was chosen similar to the spudcans of the “Marathon LeTourneau Design, Class 82-SDC” jack-up rig, as illustrated by Menzies & Roper (2008). The various scaling relationships for modelling at enhanced accelerations were reported by Schofield (1980). 2.2 Preparation of multi-layered specimen A total of thirteen tests were undertaken to investigate the bearing behaviour of spudcan foundations on (a) normally consolidated (NC) clay with an interbedded stiff layer, (b) NC clay with an interbedded sand layer, and (c) overconsolidated (OC) clay with interbedded stiffer layers. The stratigraphies were based on strength profiles from reported case histories, particularly those where punch-through failure occurred, with similar normalised layer soil properties and geometry. The results for two tests will be presented here, as summarised in Table 1. Clay samples with a uniform strength profile were prepared by consolidating thoroughly mixed, and then de-aired, kaolin slurry at 1 g in three separate cells under three different final pressures in order to obtain comparatively stiff, moderate and soft samples. Test specimens were prepared by cutting the pre-consolidated clays according to the size of the strongbox and planned layer thicknesses so that the total depth became ∼130 mm, and then assembling the layers in the larger of the two strongboxes. To produce NC clay with an interbedded stiff clay layer, NC clay specimens with a linearly increasing strength profile were first prepared by performing consolidation directly in the centrifuge. PVC plates (of unit weight similar to clay) with top and bottom geo-fabric drainage were included in the backbone clay profile somewhere in the consolidation process, depending on the planned top clay layer and second clay layer thicknesses. After completion of consolidation, the channel was stopped and the boxes removed. The PVC plates were gently lifted off and the geofabric mats were peeled off and finally a sliced stiff uniform soil layer (consolidated at 1 g) was placed in the gap (see Fig. 4).

Figure 6. Shear strength profile from T-bar test (Test I, Table 1).

with the undrained shear strength profile during spudcan penetration estimated by linear interpolation in time. The (averaged) shear strength profile, for Test I (Table 1), is plotted in Figure 6, based on a T-bar factor of NT-bar = 10.5, where z is the penetration depth of penetrometer mid-diameter. The strength gradient of the third layer (∼0.7 kPa/m), below the inserted PVC plate, is very low compared to the top layer or the usual gradient of ∼1 kPa/m. Low friction plastic pieces were used between the PVC plate and geo-fabric mat to facilitate free movement of the plate during the consolidation process under the top slurry layer. This worked for the other boxes (see the light profile in the figure) but apparently not for this one. In this particular sample, differences of up to 20% were observed in the bottom layer from two T-bar tests conducted at each end of the box. It appears that the plate must have become partially stuck, reducing the pressure on the bottom layer and hence the strength. Suitable penetration rates for various spudcan foundations and T-bar tests were chosen, balancing rate effects against the preference for undrained penetration in clay. The normalised velocity index V = vDe /cv (where v is the penetration velocity, De the object effective diameter and cv the consolidation coefficient) was around 140, with v/De ∼ 0.003–0.088 s−1 ). 3

LOAD-PENETRATION RESPONSE

The results from spudcan tests are presented in terms of ultimate bearing pressure, qu = P/A (where P is the penetration resistance and A is the largest crosssectional area of the spudcan), as a function of normalised penetration depth, d/D.

2.3 Soil strength determination Site characterisation tests were carried out using a Tbar penetrometer of diameter 5 mm and length 20 mm. In-flight strength assessments were undertaken immediately before and after each spudcan test, © 2011 by Taylor & Francis Group, LLC

3.1

NC clay with interbedded stiff layer

Figure 7 displays the load-penetration response from a spudcan vertical penetration in NC clay with an

669

Figure 7. Load-penetration response on 3-layer clay with an interbedded stiff layer (Test I, Table 1).

interbedded stiff layer (Test I, Table 1). The figure shows three interesting aspects of spudcan penetration in multi-layer soil. The spudcan penetration resistance profile initially increases gradually in the upper layer of NC clay (increasing strength) but rises sharply as the spudcan base approaches the stiff layer (noting that the spudcan tip is approximately 0.23D below the level of maximum diameter). Squeezing dominates the response due to the effect of significantly stiffer (strength ratio of 5) and relatively thicker (thickness ratio of 0.6) second layer. In practice, the question is often raised as to the depth range above the strong layer where squeezing effects start to increase the capacity. Although the limiting depth is a function of strength ratio between the two layers, from Figure 7 the effect of the stiff layer is first detected at about d/D = 0.21. Therefore, the limiting depth range is ∼0.16D (from the surface of the stiff layer). This is consistent with unpublished findings from the authors’ large deformation FE analyses. A punch-through (or rapid leg run) is triggered in the stiff layer at d/D = 0.45 and continues until the rising capacity in the bottom (third) layer, comprising NC clay, re-establishes the failure load at d/D ∼1.32. In practical field, the spudcan will rapidly penetrate this distance of 10.5 m. The figure also includes predictions using (a) SNAME approach for squeezing (1st layer), (b) Hossain & Randolph and SNAME approaches for punch-through (2nd layer), (c) Hossain & Randolph approach for single layer (3rd layer). The SNAME approach gives reasonable estimate for squeezing. For the bottom two layers, all approaches provide overly conservative estimates. In order to match the measured response in the 3rd layer, it is necessary to assume a trapped plug of stronger soil from the middle layer forced down with the spudcan (and hence additional frictional resistance – see Fig. 7). Further detailed study of the observed failure mechanisms is underway

© 2011 by Taylor & Francis Group, LLC

Figure 8. Bearing pressure from spudcan and T-bar penetration tests (Test I, Table 1).

in order to assess accurately the height of the plug carried down with the spudcan. For developing a design approach for spudcan penetration in multi-layered soils, the limiting squeezing depth, the depth at which punch-through is triggered, peak bearing capacity at punch-though and the depth of plunge must all be assessed. 3.2 T-bar to spudcan resistance There is increasing interest in prediction of spudcan performance directly from in situ penetrometer test results, just has been applied successfully in pile design. Figure 8 plots the comparative penetration resistance curves, as presented in Figures 6 and 7, from T-bar and spudcan penetration tests. Marked differences are obvious between the two penetration profiles in terms of the overall form and relevant features, largely associated with the differences in size. For the T-bar, which has a prototype diameter of 1 m, a sharp peak in resistance occurs at a depth of 9 m (z/D = 0.75) in the middle of the strong layer. A simple correlation relating spudcan and T-bar bearing capacity factors, Nc /NT-bar , even taking account of differences in strain rate, cannot be used in stratified deposits where strong changes in strength occur. As discussed by Erbrich (2005), the large diameter of the spudcan relative to the layer thicknesses allows different penetration mechanisms, and a reduced maximum resistance in the stronger layer. This needs to be taken into account when developing penetrometer-spudcan relationships in multi-layer soils with distinctive changes in shear strength. The T-bar (prototype diameter of 1 m) is much smaller than the spudcan (D = 12 m). Hence, the thickness of the stiff layer relative to the object diameter is 12 times that for the T-bar penetration. The thickness ratio is a crucial factor for assessing punch-though (Hossain & Randolph 2010). In the stiff layer of

670

thickness ratio 6.9, the T-bar resistance rises monotonically until it approaches close the lower soft layer. It then drops rapidly forming a sharp peak and continues decreasing for a certain distance into the lower layer. By contrast, the much larger spudcan tends to integrate the behaviour of the different soil layers for some distance above and below the spudcan, but also leads to significant differences in the failure mechanism as it punches through the stronger layer. In the bottom layer, the T-bar resistance is markedly lower than that for the spudcan. This reflects the fact that (a) a soil plug of significant depth was carried down with the spudcan, augmenting the penetration resistance through additional resistance around the plug periphery and (b) the soil flow mechanism around the spudcan was impeded by the stronger layer above the spudcan, again increasing the resistance. 3.3 Comparison with field data

Figure 9. Comparison with recorded field data.

As previously described, test specimens were prepared based on strength profiles from reported case histories, with similar normalised layer soil properties and geometry. As such, the spudcan penetration results can be compared directly with the measured spudcan load-penetration records to evaluate their accuracy in predicting spudcan penetration depths. For this comparison, vertical bearing capacity Qv was calculated according to

and is plotted in Figure 9. Spudcan penetration depth, d, was also translated according to the spudcan diameter ratio (Dfield /Dcentrifuge ). In addition, a depth correction was necessary because of different thickness of the top layer, which was 4.5 m in the centrifuge test and 7.45 m in the field. This has been indicated by the dashed line at the start of the centrifuge penetration-resistance curve. Kostelnik et al. (2007) reported this case history, of a Friede & Goldman L780 Mod II jack-up rig installation at Raya B offshore Peninsular Malaysia. The spudcan foundations were 12.1 m in diameter with a truncated cone extending 2.8 m beneath the circular section. The shear strength profile was as illustrated in Figure 1. A punch-through failure occurred during initial preloading. In order to reinstall the jack-up rig at the same location safely, a perforation drilling operation was carried out, with the relative area removed by drilling being about 15 %. The rig was then installed relatively successfully without any significant hazard, with the recorded load-penetration profiles as shown in Figure 9. Therefore, in calculations using Equation 1, a 15 % reduction in su,field has been assumed (the dashed line in Fig. 1). The load-penetration curve from the centrifuge test matches the field data well, providing confidence that field performance can be simulated in centrifuge tests. The discrepancy in the lower part is due to the difference in strength profiles (see Figs 1 and 6), with a

© 2011 by Taylor & Francis Group, LLC

671

Figure 10. Load-penetration response on 4-layer clay with interbedded stiff layers (Test II, Table 1).

uniform hard layer at 19.5 m in the field responsible for the sharp increase in penetration resistance below about 17 m.The squeezing depth is also consistent with the centrifuge observation (discussed in Section 3.1).

3.4

OC clay with interbedded stiff layers

Figure 10 shows the load-penetration response for the test on a 4-layer soil (Test II, Table 1), together with the (averaged) penetration response from T-bar tests. Four slices of pre-consolidated samples were assembled in soft-stiff-moderate-stiff (top down) manner in the strongbox. A punch-through failure occurs at a depth of d/D = 0.9 in the stiff layer, with a potential leg plunge of about 6.2 m. Squeezing dominates spudcan behaviour when approaching the two stiff layers (2nd and 4th). The overall forms of punch-through failure and squeezing depth (to 2nd layer) are similar to Test I. Differences between T-bar and spudcan

penetration profiles are also consistent. The limiting squeezing depth to the 4th layer is somewhat higher (0.25D) as the trapped soil plug caused the stiff layer to be sensed earlier.

Federation Fellowship programs. This support is gratefully acknowledged, as is the assistance of the drum centrifuge technician, Mr. Bart Thompson. REFERENCES

4

CONCLUDING REMARKS

This paper has reported results of centrifuge model tests investigating the potential for punch-through and squeezing of spudcan foundations during deep penetration through multi-layered deposit with interbedded stronger layers.All tests were carried out in a drum centrifuge at 200 g. Test specimens were prepared, mostly mimicking strength profiles from reported case histories particularly those where punch-through failure occurred. The following conclusions can be drawn from the results presented in the paper. 1. The experimental program successfully produced strength profiles mimicking those reported from multi-layered sediments in the field, with similar normalised layer soil properties and geometry. 2. The results from spudcan penetration tests showed promising ability to capture the load-penetration responses recorded during jack-up installation and preloading. 3. Patterns of punch-through failure (for strongover-weak) and squeezing (for the reverse) on multi-layered soils were identified, with a limiting squeezing depth of about 0.15D above a stronger layer. 4. The relative magnitudes of T-bar and spudcan penetration resistance are strongly affected by geometric differences due to the large diameter of the spudcan, and hence the relative layer thickness (t/D), and the soil plug carried down with the spudcan. The most dangerous punch-through situation occurs for relatively thin and strong (sub /sut ≤ ∼0.6) interbedded stiff layers, which can lead to rapid uncontrolled penetration over a significant depth. ACKNOWLEDGEMENTS The work forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established under the Australian Research Council’s Research Centres Program and currently supported through Centre of Excellence funding from the State Government of Western Australia and with support from the Australian Research Council Discovery and

© 2011 by Taylor & Francis Group, LLC

Brennan, R., Diana, H., Stonor, R.W.P., Hoyle, M.J.R., Cheng, C.-P., Martin, D. & Roper, R. 2006. Installing jackups in punch-through-sensitive clays. Proc. Offshore Technology Conf., Houston, OTC 18268. Castleberry II, J.P. & Prebaharan, N. 1985. Clay crusts of the Sunda Shelf – a hazard to jack-up operations. Proc. 8th Southeast Asian Geotechnical Conf., Kuala Lumpur, pp. 40–48. CLAROM 1993. Design guides for offshore structures. Club des Actions de Recherche sur les Ouvrages en Mer, Eds: Le Tirant, P. & Pérol, C., Paris. Erbrich, C.T. 2005. Australian frontiers – spudcans on the edge. Proc Int. Symp. Frontiers in Offshore Geotechnics, ISFOG, Perth: 49–74. Hossain, M.S. & Randolph, M.F. 2009. New mechanismbased design approach for spudcan foundations on stiffover-soft clay. Proc. Offshore Technology Conf., Houston, OTC 19907. Hossain, M.S. & Randolph, M.F 2010. Deep-penetrating spudcan foundations on layered clays: Centrifuge tests. Géotechnique 60 (in press). Jack, R.L., Hoyle, M.J.R., Hunt, R.J. & Smith, N.P. 2007. Jack-up accident statistics: Lots to learn! Proc. 11th Int. Conf. the Jack-up Platform Design, Construction & Operation, London. Kostelnik, A., Guerra, M., Alford, J., Vazquez, J. & Zhong, J. 2007. Jackup mobilization in hazardous soils. SPE Drilling & Completion 22(1): 4–15. Lee, K.K., Randolph, M.F. & Cassidy, M.J. 2009. New simplified conceptual model for spudcan foundations on sand overlying clay soils. Proc. Offshore Technology Conf., Houston, OTC 20012. Low, H.E., Randolph, M.F., DeJong, J.T., & Yafrate, N.J. (2008). Variable rate full-flow penetration tests in intact and remoulded soil. Proc. 3rd Int. Conf. on Geotech. and Geophys. Site Charact., Taipei, 1087–1092. Menzies, D. & Roper, R. 2008. Comparison of jackup rig spudcan penetration methods in clay. Proc. Offshore Technology Conf., Houston, OTC 19545. Paisley, J.M. & Chan, N. 2006. SE Asia jack-up punchthroughs: Technical guidance note on site assessment. Proc. 1st Jack-up Asia Conf. and Exhibition, Singapore. Schofield, A.N. 1980. Cambridge geotechnical centrifuge operations. Géotechnique 30(3): 227–268. SNAME 2008. Recommended practice for site specific assessment of mobile jack-up units. T&R Bulletin 5-5A, 1st Edition – Rev. 3, Society of Naval Architects and Marine Engineers, New Jersey. Wang, C.X. & Carter, J.P. 2002. Deep penetration of strip and circular footings into layered clays. Int. J. Geomechanics 2(2): 205–232.

672

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

A probabilistic approach to the prediction of spudcan penetration of jack-up units G.T. Houlsby Department of Engineering Science, University of Oxford, Oxford, UK

ABSTRACT: A statistically-based method is described for presenting a range of predictions of spudcan loadpenetration response. Measured data can be compared with a family of percentile curves to assess whether observations are within the range of expected behaviour. Tentative suggestions are made for decision making, based on the quality of comparison between observations and expectations. 1

INTRODUCTION

Prior to the installation of a jack-up unit at any given site, a prediction should be made of the expected penetrations of the spudcans into the seabed. This is necessary for a variety of reasons, for instance (in certain soft soils) to provide reassurance that the available leg length is sufficient. The process of making the prediction also focuses attention on identification of hazards, such as the possible occurrence of punch-through. In order to carry out the prediction, various types of information must be assembled. Site investigation data on the properties of the soils that will be encountered is of course of primary importance. From these data an appropriate (idealised) model of ground conditions must be established. This will usually be in terms of a number of soil horizons, with quantified soil properties (e.g. strength and unit weight) for each horizon. The level of accuracy and detail necessary varies with many factors, including considerations such as experience from nearby sites as well as the actual nature of the ground conditions. In addition to soils data, the geometry of the spudcan must also be known. Again this is idealised, usually by approximation of a polygonal spudcan as axi-symmetric. Finally an estimate of the loads imposed by the rig, up to maximum preload, must be known. Equipped with all the above information, an engineer can make a prediction of the expected loadpenetration response of the spudcan (by load we mean net load applied by the spudcan to the seabed, accounting for buoyancy effects, and by penetration we mean the depth of penetration of the tip of the spudcan below the nominal mudline). The InSafeJIP project, on which this paper is based, is primarily concerned with the reliable prediction of this load-penetration response. Assuming that no hazards are identified that prevent it, the installation will then proceed. It is then prudent for the engineers on site to monitor the actual behaviour. At the very least the penetration of each spudcan at

673 © 2011 by Taylor & Francis Group, LLC

light-ship load (hull out of water but prior to preloading), at full preload and after dumping the preload should be measured. However, best practice at any site where significant penetrations are expected would in fact be to monitor sufficient data to construct a much more detailed measured load-penetration response. Inevitably the actual load-penetration response will differ from the prediction. There are uncertainties attached to the parameters on which the prediction is based, on the calculation methods employed and on the measurements made in the field. The engineer on site is therefore faced with a dilemma: at what stage are the discrepancies between the measured behaviour and the prediction sufficiently large that “alarm bells ring”, and further action is necessary to assure (if possible) the safety of the rig. This problem has usually (not always) been addressed by making multiple predictions: for example predictions may be made on the basis of “expected” and “lower bound” soil strengths, which in turn will lead to “expected” and “maximum” spudcan penetrations at a given load. If a further prediction using “upper bound” strengths is used then “minimum” penetrations would also be obtained. In certain circumstances (e.g. an installation in soft clay with strength increasing with depth) an engineer might be assured of the safety of the rig provided that the “maximum” penetrations are not exceeded. The definition of, for instance, “lower bound” strength is, however, a somewhat inexact science, and almost certainly depends on the subjective judgement of the engineer making the initial prediction (or possibly even a different engineer, who provides interpreted soils data to the predictor). The precise status of the “maximum” penetration prediction is therefore a little uncertain. Consider also the following scenario. The soil profile is expected to consist of a very soft clay overlying a much stronger clay at a certain depth. The rig has fairly flat-based spudcans, and the soft clay is insufficiently strong to carry the preload (whether “lower

bound” or “expected” strengths are used), whilst the stronger clay has adequate capacity. Both the lower bound and expected calculations will predict the same penetration – such that the spudcans rest on the surface of the stronger material. Suppose then that two of the spudcans are observed to penetrate to that depth, whilst the third penetrates say 2 m more. How are we to interpret this observation? A likely reason is that the base of soft clay is 2 m lower at this location than at the other spudcans, but the framework in which the only variation that is considered is the possible variation of strength cannot accommodate this possibility, and the penetration is simply observed as outside the expected range and greater than the expected maximum. Our intention here is to explore a framework that permits a more rational approach to such problems. It is based on the concept that each parameter employed in the calculation is assigned a statistical variability. We combine these in a way that predicts a statistically quantifiable range of possible expected behaviour, and the measurements can then be viewed in this context. This paper does not represent completed work, but work in progress. Firstly the proposed method employs information which is currently not readily available within the industry. Secondly it employs some factors that are intrinsically difficult to quantify. Finally, the results of the calculations, when compared to measured data, raise a further conundrum, discussed later. We make no pretence therefore that this paper presents a “solution” to the problems described above, what it presents is a vehicle for discussion of the issues at stake, and how they can be handled. Note that probabilistic approaches to jack-up assessment have been made before, see Van Der Wal et al. (2007), but the approach presented here is rather different to the previous work.

2 2.1

STATISTICAL APPROACH Overall method

We now set out an approach that attempts to provide a rational basis for predicting a possible range of behaviour, such that the range has a clearly defined meaning. At any given penetration the 5th, 25th, 50th, 75th and 95th percentile loads can be presented (see Figure 1, which shows a real but anonymous example). The 50% line represents the most likely behaviour, but the range of the lines represents the many uncertainties in the problem. It is expected that 50% of observations would lie between the 25% and 75% percentile lines, and 90% between the 5% and 95% lines. The spread of the predictions depends on an understanding of all the sources of uncertainty in the problem. These may be grouped as follows:

Figure 1. Example of predicted load-penetration response.

made. Different tests that purport to measure the “strength” may in fact measure slightly different physical properties. The elevations of samples, and of horizons between soils with different properties, cannot be determined exactly. 2. Definition of geometry. Although in principle it should be possible to obtain accurate information about spudcan geometry, in practice some minor uncertainty will attach to the plan area etc. 3. Calculation uncertainty. The calculation procedures for the load-penetration curve typically involve a number of simplifying assumptions, and employ factors derived from a variety of sources (either theory or correlation from experiments). There is therefore an uncertainty as to how accurately the calculation represents the physical problem. 4. Observation errors. The predicted load-penetration response is compared to field measurements which are themselves subject to error. The weight of the jack-up unit will only be known within a certain margin, and the distribution of load between the spudcans adds a further uncertainty. The penetration of the spudcan into the seabed is not known to any great precision.

1. Measurement of soil properties. The soil properties are typically measured by methods of variable reliability. Soil samples may not be truly representative of the soils at the location of the spudcan. Samples may be disturbed before strength measurements are

674 © 2011 by Taylor & Francis Group, LLC

In the proposed method, the first three sources of uncertainty are all accounted for in the presentation of the predicted set of load-penetration curves. The errors associated with the final group are likely to be better known to those actually operating the rig, and they could be treated by presenting error bars on the measurements of load and penetration (as in fact used on Figure 1, although the bars in this case show a sufficiently small range that they are not apparent). The following discussion relates to the first three sources of uncertainty. In the context of reliability theory, each variable that affects the calculation is assigned a statistical distribution rather than a unique value. For relatively simple calculations it is possible to determine the statistics of

the output result (the load-penetration curve) directly in terms of the statistics of the input parameters. For more complex problems, such as the determination of spudcan resistance, a direct calculation is not possible. The variability of each quantity is most easily represented by assuming that it follows a normal distribution, although more complex distributions could be used. Each variable is therefore represented by its mean and standard deviation. It is commonly found, however, that the magnitude of the standard deviation is roughly proportional to the mean value. For instance, for undrained strength, it may be more meaningful to say that a test would provide values that are reliable to within ±10% than to within ±2 kPa, as the absolute error may increase roughly in proportion with the strength. This leads to the concept of coefficient of variation (CoV) which is defined as the standard deviation divided by the mean, for instance:

between the 190th and 191st load value in the ascending ranked order of 200 loads. • The percentile points at different depths are joined to form the overall percentile lines. Note that, because of the way the variables are distributed, and because each calculation is independent, the individual load-penetration curves often cross each other. The 95th percentile curve (for instance) does not therefore track any single load-penetration curve. It is very important that the percentile lines are correctly interpreted. An experienced engineer should expect that (about) half the observations would lie outside the 25th–75th percentile range, and that (about) one case in ten would be outside the 5th–95th percentiles. Just because one point is outside the range does not mean that the calculation is “wrong” or the observation in error. 2.2

where: Vsu = Coefficient of Variation of su σsu = Standard Deviation of su µsu = Mean of su For a calculation in which the output is simply the product of input values, as in Eq. (2), which is a simplified version of the bearing capacity equation:

the CoV of the output can be determined directly:

(This depends on the fact that variations of the three variables are uncorrelated, which is the case here). However, the calculation of the penetration resistance is much more complex than the simplified form shown in Eq. (2). Although Eq. (3) could be used to estimate the overall CoV, it would not take into account the effects of many other variables, for instance the influence of backflow and buoyancy. A common procedure in these circumstances to determine the statistics of the output is to use “Monte Carlo” methods. A number of calculations are made, with values of the input parameters chosen, based on their statistical distributions. If a large number of calculations (say 200) are carried out, then the statistics of the output can be determined with reasonable accuracy. This is implemented as follows: • Each parameter is assigned a statistical distribution. • 200 separate calculations are made, with each

parameter assigned a value based on the statistics of the distribution. 200 calculations seem to provide a robust estimate of the spread of the results. • At each depth the 200 calculated values of load are ranked in ascending order. • The 5th, 25th, 50th, 75th and 95th percentile values of load are determined at each depth. For instance, the 95th percentile line is calculated as midway © 2011 by Taylor & Francis Group, LLC

Standard deviation and CoV values

For some of the variables in the analysis it is most natural to assign an absolute Standard Deviation, and for others it is more natural to use a CoV (i.e. a Standard Deviation proportional to mean). A pragmatic approach is adopted in the following. For simplicity we only consider spudcan penetration in clay (which may be multilayered, as long as all layers can be classified as clay). In the following, very occasionally the use of a Normal Distribution leads to infeasible values – these are simply rejected and a new value chosen from the distribution. 2.2.1 Spudcan geometry Although one might expect that accurate dimensions for spudcans should be available, in practice there may be uncertainties relating to spudcan geometry. The assumption that the spudcan can be idealised as axisymmetric introduces a further uncertainty. The main result of dimensional uncertainty is the effect it has on the calculated plan area of the spudcan. The uncertainty is introduced by specifying a tolerance T for the linear dimensions of the spudcan, defined as follows: 50% of the true values would be expected to lie within ±T from the nominal value. The Standard Deviation is directly proportional to the tolerance, σ ≈ 1.483T . Example: the dimensional tolerance may be taken as T = 0.05m, resulting in a Standard Deviation on all dimension values of σ = 0.074m. 2.2.2 Soil properties Undrained Strength: The variability of the undrained strength is expressed through a CoV value. Note that approximately 68.3% of measurements should be expected to lie in the range µsu ± σsu = µsu (1 ± Vsu ). The CoV for the undrained strength reflects a number of uncertainties, including: (a) uncertainty about the measurement technique: this is likely to apply systematically across multiple layers at a site. Thus, for instance, if a given tool tends

675

to underestimate soil strength by 10%, it is likely to do so for all soils at the site, so the relative strengths of different layers will be correct, but their absolute magnitudes will be systematically 10% low. (b) variability of the actual strength within the soil layer: this may differ from one soil layer to another. Provision is therefore made to split the CoV into two components, one specified as the overall CoV for all soil layers, and the other applied as an individual CoV to the strength for that layer. When the computation of load penetration curves is carried out, the nominal soil strengths for each layer are multiplied by two factors, the first determined by the overall CoV and the second (determined separately for each layer) determined by the layer CoV. If the layer CoV is set to zero, then all the strengths are varied together, whilst at the opposite extreme, if the overall CoV is set to zero, the strength of each individual layer varies independently. The net total CoV
for any one layer may be calculated 2 2 as Vsu(total) = Vsu(overall) + Vsu(layer) . Example: the overall CoV for undrained strength may be taken as 0.10, and the individual layer CoV taken √ as 0.05, resulting in a total CoV for each layer of 0.12 + 0.052 ≈ 0.112. Buoyant unit weight: A simple overall CoV is applied to the buoyant unit weight γ
.

2.2.3 Positions of soil horizons The elevations of the tops of all soil horizons are specified relative to the mudline. A tolerance is applied to these elevations in the same way as to the spudcan dimensions. The tolerance can be used to reflect both the accuracy of determination of levels within a borehole, as well as any possible variations in layer elevations due to the fact that the borehole may be some distance from the spudcan location. Example: high quality CPTU testing at each spudcan location is used to identify soil horizons and the depth tolerance is set to 0.15m. Example: a single borehole near the centre of the rig location is used. Continuous sampling is used but there is some loss of core length. The depth tolerance is set to 0.5m.

Roughness factors: The capacity factors depend slightly on the spudcan roughness, so a CoV is applied to the roughness factors. Backflow calculation: The method described by Hossain et al. (2006) is used for calculating whether backflow occurs around the spudcan. This method uses a number of semi-empirically determined constants, each should be assigned a CoV, although at present the appropriate values are uncertain. 2.2.5 Implementation The above procedures have been coded in a Microsoft Excel spreadsheet. The calculations are too complex to be expressed simply as spreadsheet formulae. Instead, once the data has been assembled, a calculation programmed in Visual Basic is used. As described above, the calculation is repeated many times, with the parameters adjusted randomly within the ranges specified by their statistical distribution. The values of calculated capacity are ranked and the percentile values determined as shown in Figure 1. 2.2.6 Calibration It is important that the range of response shown is realistic. To calibrate the above process with the chosen CoV values and other statistical parameters, predictions were compared with case records for penetration in clay taken from the InSafeJIP database. Exactly 100 observations of penetration were available. The percentile load represented by each observed data point was obtained from the appropriate predicted loadpenetration curve, and these percentile values ranked in ascending order. If the predicted percentile lines represent (approximately) the real variability in the problem, then (about) 5 data points would be expected to be at percentile values below 5th , 20 between 5th and 25th etc. Figure 2 shows a the “actual percentile” (the ranking position amongst the 100 values) against “predicted percentile” (the value read from the load

2.2.4 Calculation uncertainty CoV’s are also used to reflect the uncertainties in the methods used for the calculations. Bearing capacity factors: a CoV is applied to the calculated bearing capacity factor (taken from Houlsby and Martin, 2003), reflecting the uncertainty of the applicability of the analysis to the spudcan foundation problem. It is difficult to determine the appropriate CoV, but one way of estimating a value is to consider the range of values of bearing capacity factors determined by different techniques (e.g. method of characteristics, finite element analysis, direct correlation from experiments). These considerations indicate that a CoV of the order of about 0.1 may be appropriate on the bearing capacity factor. © 2011 by Taylor & Francis Group, LLC

676

Figure 2. Comparison of actual and predicted percentiles.

penetration curve). All points lying on the 1:1 line would indicate that the variability is predicted correctly. We see that the parameters used have indeed captured most features of the variability, although improvements could be made.

3

COMPARISONS WITH DATA

Equipped with the range of predictions described above, an engineer can now proceed with installation of a jack-up. A prudent engineer would measure (or at best estimate) leg loads and penetrations during installation, and plot these data with the predictions. When plotting the data the engineer should be aware that Table 1.

the measurements, like the predictions, are subject to error. For instance the load on the spudcan may at best be known to perhaps ±100 kN, and the leg penetration to ±0.3 m. Appropriate error bars need to be assigned to each data point. We now consider various possible scenarios. First consider the case where the data follow the trend of the predictions rather closely, and fall entirely within the 25th–75th percentiles. The engineer can be assured that the behaviour is as expected, and presumably takes no further action. Now suppose the trend is still followed, but the data plot between the 5th and 25th percentiles. The engineer should not be too alarmed, since they are expecting (if the percentile lines are at a spacing that genuinely

Categories of observed response, and suggested actions

© 2011 by Taylor & Francis Group, LLC

677

models the real variability in the problem) that in 50% of the cases they observe the response will lie outside the 25th–75th percentile range. What if the trend is still followed, but the data fall just below the 5th % percentile curve? If the engineer has extensive experience, they will know that one time in ten they should expect a response outside the 5th–95th percentile curves, so they should hardly be surprised by this occurrence. At some stage, however, the response will be sufficiently far outside the 5th–95th percentile curves that it represents a genuinely rare event, and our hypothetical engineer should perhaps be alarmed. How far outside? How rare? We have replaced the somewhat arbitrary “lower bound” strength with a rationally-based statistical procedure, but the conundrum is that we may be no nearer identifying the response that should “ring alarm bells”. We have, however, provided a vehicle for the discussion of these issues. It is possible for instance that we can make some rational connection between a statistically unlikely response and a genuinely hazardous one. At the very least, if the response is identified as unusual (in a quantifiable statistical way), then it will alert the engineer to the need for further attention. Furthermore, deviations from the norm above the 95th percentile might justify quite a different response from those below the 5th. One approach might be to relate (at each individual site) the percentile lines to identifiable Limit States. The problem is further complicated because we should compare not just a single data point with an expected distribution, but a whole response curve. It is possible, if for instance the range of predictions is rather wide, that all the data points may be contained within the 25th–75th percentile lines, and yet the response could show a trend very different from the expected one. Whether or not this may be alarming depends very much on the individual circumstances, however, it may well indicate a fundamental flaw in the interpretation of the soil stratigraphy. In spite of the above comments, Table 1 sets out tentative suggestions for decision-making based on the

© 2011 by Taylor & Francis Group, LLC

above approach. The action recommended depends not just on the (quantifiable) deviation from the expected mean, but on a qualitative assessment of how closely the predicted trend is followed. 4

CONCLUSIONS

A statistically-based method is described for establishing a range of predicted load-penetration curves (percentile curves) for spudcans on clay. The implications for comparisons between observed behaviour and these predictions are then discussed. ACKNOWLEDGEMENTS This work was carried out as part of the InSafeJIP project (see http://insafe.woking.rpsplc.co.uk), and the support of the sponsors is gratefully acknowledged. However, the methods described here do not form part of the formal recommendations from that project, because of the speculative nature of some aspects of the approach described here. The author is grateful for very useful feedback from the project team, especially from Julian Osborne (RPSE), Prof. Mark Cassidy (UWA) and Dr Teh Kar-Lu (NUS). REFERENCES Hossain, M.S., Randolph M.F. Hu, Y. and White, D.J. (2006) Cavity Stability and Bearing Capacity of Spudcans on Clay. Offshore Technology Conference, paper OTC 17770. Houlsby, G.T. & Martin, C.M. (2003) Undrained bearing capacity factors for conical footings on clay. Géotechnique, 53 (5): 513–520 Van Der Wal, T, Boumeester, D, Peuchen, J, Vrouwenvelder, A.C.W.M.Van Baars, S and Lagers, G.H.G. (2007)A probabilistic approach for geotechnical assessment of jack-up leg penetration. Proc. 6th Int. Offshore Site Investigation and Geotech. Conf., 11–13 Sept., London, 515–520.

678

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

An assessment of jackup spudcan extraction O.A. Purwana, H. Krisdani, X.Y. Zheng, M. Quah & K.S. Foo Keppel Offshore & Marine, Singapore

ABSTRACT: Extraction of jackup spudcan in clayey seabed is often found challenging and may lead to unfavorable implications if not planned properly. Despite severity of the issue, no studies have been devoted to investigate any interaction of overall factors involved in the leg extraction process. The present paper aims to offer some insight into a potential way of assessing leg extraction problem in a more comprehensive manner. 1

INTRODUCTION

Deployment of jackups in clayey soils poses potential leg extraction challenges, which if not addressed in a planned manner, could result in additional rig move time, affect structural integrity of the unit and lead to higher spread costs due to unintended delays. Among other factors, delays in removing the legs can be attributed to the uncertain effectiveness of the typical water jetting system as well as the apparently lack of understanding on the impacts of certain leg pulling out procedures adopted in the field. Despite the unfavorable implications, to date very little attention is given to investigations of jackup leg extraction problems in a comprehensive manner. Thus far, most of the existing studies related to pullout were developed for anchor-type foundations and are still focused on revealing specific aspects of the breakout phenomenon. Only a limited number of studies have been conducted specifically on extraction of spudcan-type foundation in clay. These studies particularly aimed to quantify the pullout resistance with a more realistic simulation of installation and extraction of a single spudcan as well as to investigate behavior of the water jetting. The present paper attempts to investigate several key factors that apparently have some influence on the overall extraction process. A case study of leg extraction problem in soft clay shall be evaluated from various perspectives and discussed herein.

A several feet draft in excess of the neutral draft, or often called “overdraft”, is typically introduced to the legs simultaneously or in cycle to generate “static” pullout force. The latter is often adopted in order to control the movement and release of a certain leg particularly when the jackup was operating adjacent to other platforms. Despite generating larger overdraft seeming advantageous, it is not clear whether imposing the maximum possible buoyancy force, sometimes even with the risk of inundating the hull, is really effective in easing the extraction resistance. To overcome limited buoyancy of the hull, the leg is typically equipped with a water jetting system at the top and bottom of the spudcan. Supplied with high water pressure of up to 1000 psi, the jetting nozzles are commonly perceived to be able to break suction/adhesion on the spudcan surface and to destroy accumulated sediments above the spudcan. However, very little is known about the actual behavior of the conventional jetting system and to what extent it contributes to accelerating the pullout process. Another factor which has yet to be investigated is the potential contribution of sea state to the pullout force and resistance. The prolonged action of small waves hitting the submerged hull during removal generates a cyclic load which will then be transferred to the spudcans. Apart from providing dynamic uplift force, the cyclic loading combined with potential consolidations might somewhat alter shear strength of the soil surrounding the spudcans.

2

3

LEG EXTRACTION PROCEDURES

Unlike jackup installation process, there have been apparently no standard procedures for pulling out the legs during removal. In general, hull buoyancy and water jetting are utilized to generate uplift force and to disturb the soil in some ways. However, each jackup operator apparently has their own procedural techniques which may be largely derived based on their preference and experience in dealing with leg extraction problems.

3.1

Fundamental understanding

As described in Purwana et al. (2005), the total uplift resistance of spudcan Quplift is composed by three main components, that is,

679 © 2011 by Taylor & Francis Group, LLC

SPUDCAN PULLOUT RESISTANCE

where Qtop = resistance of soil above spudcan (shear resistance + soil weight); Qbase = resistance of soil

below spudcan (base suction); and Weff = submerged weight of leg and spudcan. This equilibrium condition assumes negligible friction force on the spudcan’s side wall. The experiment revealed that the tensile stress change at the spudcan base during uplift under undrained conditions is transmitted essentially by a drop in pore-water pressure. The resulting negative excess pore water pressure with respect to the corresponding hydrostatic pressure at the spudcan base, defined here as “base suction”, constitutes the bottom resistance which is later found to be associated with undrained shear strength of the bottom soil. It was also demonstrated that the ultimate breakout force is achieved with the ultimate base and top shear resistance reached nearly simultaneously. Through Particle Image Velocimetry – PIV technique (White et al. 2003) used in the experiment, the failure mechanism of spudcan extractions in undrained condition with penetration depth equivalent to 1.5 times spudcan diameter can be visualized. As schematically illustrated in Figure 1, the breakout failure is accompanied by the shallow-anchor failuretype mechanism above the spudcan while below the spudcan a reverse bearing capacity phenomenon is observed. The subsequent post-breakout uplift is characterized by the presence of residual base suction and localized soil backflow around the spudcan edge from the spudcan top to below the base. 3.2

Estimate of spudcan pullout resistance

Based on a number of data sets from centrifuge tests on spudcan extraction in clay with various shear strengths profiles and penetration depths, Purwana et al. (2009) described in detail a proposed semi-empirical formula for predicting the ultimate breakout force of spudcan in undrained clay for penetration depths of up to two times diameter. The formula distinguishes the contribution of soil resistance at the top and bottom of the spudcan as follows,

Figure 1. Displacement vectors and simplified failure mode of spudcan extraction with shallow penetrations.

where cu,base = shear strength at spudcan base level after installation; Fg,base = gain in shear strength of soil below spudcan base due to any soil reconsolidation
= average unit weight after spudcan installation; γside of soil alongside the spudcan base; Db = depth of spudcan base (see Fig. 1); and Sb = empirical adjustment factor for overburden stress at spudcan base level. It should be noted that the above formulae has only been calibrated for Db /B ≤ 2.0.

where B = spudcan diameter; Nc,top = breakout factor for top soil resistance and S = shape factor (Merifield et al, 2003), Dt = height of backfill above spudcan top surface (see Fig. 1); cu,top = average shear strength of backfill soil above spudcan top after installation;
= average unit weight of backfill soil above spudγtop can top; and Fg,top = change in shear strength of soil immediately above spudcan top due to soil disturbance and any reconsolidation after spudcan installation.

© 2011 by Taylor & Francis Group, LLC

680

4 WAVE-INDUCED PULLOUT FORCE In order to quantify the level of cyclic load induced by a sea state to the legs during jackup removal, a hydrodynamic analysis is performed to simulate wave actions onto a submerged jackup hull through a single 6-hour time domain simulation. Random small waves with significant heights Hs of 3 ft and 5 ft are passed through the hull with a draft of 16 ft, 18 ft, and 20 ft. To evaluate the effect of peak wave period,Tp is varied from 5 sec to 20 sec despite the limiting sea-state for installation and removal, particularly due to leg touch-down, may be far less than 20 sec.A water depth of 250 ft and penetration depth of 75 ft are assumed with the corresponding total leg horizontal stiffness incorporated in the simulation. Wave incident angle is varied from 0 deg to 180 deg from the stern, with an interval of 30 deg. A hydrodynamic software, WAMIT (WAMIT Inc. 2006), is used for generating the response amplitude operators (RAOs) of a generic three-legged jackuphull submerged with the conditions specified above. RAO herein is the amplitude of wave force produced by a sinusoidal wave of amplitude 1 m (0.3048 ft)

However, in cohesive material where most of the extraction problems arise the fluidization may not occur. In the absence of experimental evidence, hypothetically the water jetting actions may be limited to (i) creating soil fractures which could propagate to the soil surface, (ii) forming local cavities and (iii) filling void left by uplifting spudcan. Potential performance of the water jetting in assisting spudcan extraction is discussed qualitatively below. 5.1 Top jetting Top jetting is often aimed to “destroy” accumulated sediments above spudcan and reduce shear strength. Effective reduction of soil weight or volume through the jetting applications with the nozzles fixed at the spudcan top seems unlikely unless a high-capacity dredge pump is used to transfer soil mass above the spudcan elsewhere. Significant reduction of shear resistance along the potential failure surface may be attainable should the nozzles arrangement and applied pressure allow soil fracturing right at the location of shear zone. Thus far, there is no study investigating the potential use of hydraulic fracturing methods for easing uplift resistance in cohesive soil.

Figure 2. Maximum excitation load at the centroid of hull base.

5.2

It was discussed that the high water pressure applied at the spudcan base through the jetting nozzles tends to create soil channeling or hydraulic fractures. With the fractures created, there is a high possibility that any uplift-induced base suction can escape through the fractures although uniform pressure build-up at the spudcan base is potentially hard to develop. The limited number of nozzles and potential clogging also makes effective pressure build-ups more difficult to achieve. Without any field evidence, eliminating the base suction is probably the best capability that the conventional water jetting can offer.

Figure 3. Variation of maximum vertical and horizontal cyclic loads for various wave incident angles.

at a given period. Assuming high vertical soil stiffness at the onset of extraction, vertical and rotational movement of the hull is constrained for simplicity of the analysis. Flexibility of lateral hull movement is modeled with a spring constant at the hull centroid. However, the hydrodynamic analysis only takes diffraction into account while the lateral hull motion and associated radiation are not included. The analysis also produces the phase lag of wave force required in time-domain simulations. The RAOs in terms of maximum vertical and horizontal forces for the wave incident angle of 0 deg is presented in Figure 2. The wave-induced vertical excitation load substantially increases with wave period while the horizontal load slightly decreases after peaking at the rig natural period. No significant gain in RAOs can be obtained by increasing the hull draft from the neutral buoyancy to the perceivably maximum safe overdraft commonly practiced in the field. Analysis with various wave incident angles indicates that the maximum vertical load occurs at the incident angle of 120 deg from the stern (see Fig. 3).

6

CASE STUDY

An assessment of leg extraction is demonstrated herein using the results derived from the preceding hydrodynamic analysis and an assumed geotechnical condition. The key simulation parameters are summarized in Table 1 below. The resulting time history of dynamic force at the centroid of hull base is presented in Figure 4. Apart from the static and dynamic vertical uplift forces, the horizontal dynamic force acting on the hull may also significantly contribute to the total dynamic uplift force at the spudcan. Assuming a pinned footing, the composition of vertical cyclic force generated at the forward spudcan for a 6-hour sea state is given in Table 2 with a single peak of Vcy,max = 16.2 MN.

5 WATER JETTING EFFECT

6.1

It is commonly perceived that water jetting fluidizes the soil and thus weaken its resistance against pullout.

© 2011 by Taylor & Francis Group, LLC

Bottom jetting

681

Discussion

Combining static uplift force of 10.9 MN and dynamic components of 12.0 MN (average of the one-third

Table 1.

Simulated leg extraction condition.

Parameter Water depth Shear strength, Unit weight Spudcan diameter Spudcan base depth Estimated uplift resistance (from Eqs. 1–6) Significant wave height Hs Peak wave period Tp Wave spectrum Wave incident angle Hull overdraft, Static uplift force Single sea-state duration

250 ft (76.2 m) 1.0 + 1.75 z [kPa], 7 kN/m3 50 ft (15.2 m) 75 ft (22.9 m), Db /B = 1.5 Qtop = 28.6 MN Qbase = 9.7 MN 5 ft (1.52 m) 6 sec JONSWAP with Gamma 3.3 0 deg (from stern) 4 ft, 10.9 MN (per spudcan) 6 hours

Figure 5. Accumulated cyclic strain of the soil above spudcan due to uplift force generated by multiple 6 hour sea-states.

Figure 4. Time histories of cyclic load of a single simulation. Table 2. Vertical cyclic load composition at spudcan. Group

Vcy /Vcy,max (%)

Number of cycles

1 2 3 4 5 6 7 8 9 10 11

0–10 10–20 20–30 30–40 40–50 50–60 60–70 70–80 80–90 90–95 95–100

337 721 879 839 583 312 145 64 18 2 1

Notes: 6 hours sea state Cyclic load Vcy is a single amplitude load, i.e. uplift component, generated at spudcan Total cycles = 3901

largest crests), the maximum uplift force generated at the spudcan is apparently less than the predicted total net uplift resistance of 38.3 MN. If it is assumed that the water jetting applied at the spudcan base is able to eliminate the base suction, there is a remaining 28.6 MN of the top soil resistance to overcome with the soil weight contributing 23.4 MN to the total top resistance. In the calculation of top soil resistance, a reduction factor of 0.67 is applied to the soil shear

© 2011 by Taylor & Francis Group, LLC

strength to account for the disturbance due to spudcan installation. In view of the available uplift force being smaller than the resistance, potential effects of the cyclic loading to the spudcan extraction process are examined. The strain-accumulation procedure (Andersen et al. 1992) is adopted to roughly evaluate the time needed to fail the soil with the characteristic of cyclic loading generated at the spudcan. The failure is defined here as a cyclic shear strain of 15%. The assessment of cyclic load effect is primarily focused on the soil above the spudcan which, from the experiment (Purwana et al. 2008), was shown to experience shearing along the plan directly above the edge of the spudcan with a relatively shallow embedment, i.e. Db /B = 1.5, as simulated in this case. From the observed failure mode, the loading condition along the potential failure surface apparently resembles direct simple shear (DSS) tests. In the absence of field data, DSS tests results on normally consolidated Drammen clay, with plasticity index Ip = 27% (Andersen, 2004), are used in the analysis to assess the cyclic loading effects. The cyclic shear strength is typically defined by,

where τf ,cy = cyclic shear strength, τa = average shear stress and τcy = cyclic shear stress. The average shear stress along the potential failure surface generated by the static uplift force is considered small in view of the top soil weight that has to be overcome. After being offset with the top soil weight, the net cyclic uplift load overcoming the top shear resistance corresponds to a maximum ratio τcy /su of about 0.70. Figure 5 shows the accumulated cyclic strain induced by multiple 6-hour sea states based on τcy – γcy – N relationship for Drammen clay with OCR = 1

682

and τa = 0.0. With the level of vertical cyclic load generated by the simulated sea state on the 20 ft submerged jackup-hull and the soil resistance offered by the assumed geotechnical condition, it is found that a failure of the top soil may occur after six consecutive simulations of a 6-hour sea state. This simplified analysis assumes the soil remains undrained throughout the course of extraction and negligible rotational stiffness offered by the spudcan. 7

POTENTIAL IMPROVEMENTS OF BOTTOM JETTING

It is known that most jackups is unable to apply large tensile load to the legs due to limited hull buoyancy. The maximum uplift force that can be generated by the hull buoyancy to the individual legs is generally about 20% to 50% of the maximum preload base reaction. On the other hand, the net ultimate uplift resistance of a spudcan embedded in normally consolidated clay can be more than half of the installation load even for immediate extraction cases where no consolidation of the soil surrounding the spudcan taking place. This implies that wave-induced cyclic loading and water jetting play an important role in a successful leg extraction. It has been demonstrated in the preceding section that under a certain situation the wave-induced cyclic load is able to generate a significant pullout load augmenting static uplift force provided by the hull overdraft to overcome the top soil resistance. In addition, the cyclic loading may also lead to soil failure due to the combination of accumulated static and cyclic strains. However, the reliance on sea state alone will not be adequate if the potential suction develops at the spudcan base is not minimized or eliminated. 7.1 Importance of effective bottom jetting Gaudin et al. (2009) performed an extensive centrifuge tests primarily to study the effect of jetting flow rate and extraction rate to spudcan pullout resistance in normally consolidated kaolin clay. The study demonstrates that the reduction of base suction is proportional to the flow rate of water supplied to the spudcan base. If an undrained condition can be achieved, a jetting flow rate corresponding to 70% of the cavity volume per unit time left by the uplifting spudcan is required to prevent any development of suction at the spudcan base. On the other hand, in partially drained conditions, high flow rate is found not effective enough to overcome the increase in uplift resistance due to consolidation of the soil at the top of the spudcan. The experiment suggests that the effectiveness of water jetting is apparently governed by the supplied flow rate provided an optimum pressure can be maintained at the spudcan base. 7.2 Alternative spudcan mitigation device Potential limitations of the conventional water jetting system and alternative extraction mitigation technique

© 2011 by Taylor & Francis Group, LLC

Figure 6. Concept of alternative extraction device (Patented).

have been discussed in Purwana et al. (2008) with the latter described in more detail in Purwana (2006). Besides potential hydraulic fractures associated with high jetting pressure, uncertain jetting pressure distribution due to potential line clogging and inherent implications of the nozzles arrangement is deemed one of the main drawbacks of the conventional jetting system. In view of this, potential improvements of the jetting system particularly for the spudcan bottom are worth exploring. The study on alternative extraction mitigation device performed in the National University of Singapore (NUS) demonstrates that building up an optimum pressure over the spudcan-soil interface area is found more effective in reducing the base resistance rather than “disturbing” the soil by applying excessive jetting pressure. The pressure build-up is aimed to keep the excess pore water pressure at the spudcan base positive throughout the pullout. Besides compensating base suction, the excess pressure can also provide additional uplift force to overcome the top soil resistance. The concept is schematically illustrated in Figure 6. A combination of pressure chamber and porous material is used as a jetting outlet system in the spudcan model. All the outlets are connected to a single chamber to help ensure uniform pressure distribution and provision of sufficient volume of water to be transferred to the spudcan base interface particularly at the onset of extraction. The pressurized water-layer created at the interface would then serve as a buffer region

683

Despite at certain situations the sea-state being able to provide significant uplift forces, an effective bottom jetting is deemed essential not only to minimize base suction and thus the total uplift resistance but also to provide additional uplift forces. Building up optimum and well-distributed pressure across the spudcan base is found to be the key element for an effective bottom jetting. ACKNOWLEDGMENTS Figure 7. Variation of uplift resistance components for various levels of pressure build-up at the spudcan base.

to prevent the uplift stress induced by the spudcan from being transmitted to the underlying soil. With this water-filled gap, upward force can also be generated at the spudcan base provided the pressure is kept below the hydraulic fracture limit. The use of porous material for the outlets is aimed to minimize potential line clogging and avoid directionality of the applied pressure. The outlets are equally distributed across the spudcan base to ensure well pressure-spread. With a minimum total outlet area corresponding to 1% of the projected spudcan base area, effective reduction of base resistance can be obtained by applying an excess pressure (with respect to hydrostatic pressure at spudcan base) below the hydraulic fracture limit. In Figure 7, it is shown that the base suction decreases in proportion with the excess pressure built up at the spudcan base. The maximum base suction can be fully negated if a similar amount of excess pressure can be generated. Beyond this level, uplift force will be developed at the spudcan base indicated by the total uplift resistance being smaller than the top soil resistance. However, applying pressure higher than the fracture limit seems to be less effective which may be attributed to some pressure loss through potentially created soil-channels. From the observed correlation between flow rate and uplift rate, it is estimated that 60% to 80% spudcan base area is covered by the water-filled cavity. 8

CONCLUSION

An assessment of jackup leg extraction mainly from geotechnics and hydrodynamics perspectives has been presented in this paper. A case study of leg extraction in normally consolidated soft clay was used to demonstrate the potential contributions of several factors involved in the leg pullout process.

© 2011 by Taylor & Francis Group, LLC

Some of the results presented herein are derived from the Keppel-NUS collaborative study performed in 2004–2006 under the supervision of Prof Leung Chun Fai and Prof Chow Yean Khow. The authors also gratefully acknowledge valuable input from Steve Nowak of Integrity Offshore and Julianto Cahyadi on some practical aspects of leg pullout issues. REFERENCES Andersen, K.H. Dyvik, R., Kikuchi, Y., and Skomedal, E. 1992. Clay behavior under irregular cyclic loading. In Proc. Int. Conf. Behavior of Offshore Structures, London, 7–10 July 1992. London: BPP Technical Service. Andersen, K.H. 2004. Cyclic clay data for foundation design of structures subjected to wave loading. In T. Triantafyllidis (ed.), Proc. Intern. Conf. Cyclic Behavior of Soils and Liquefaction Phenomena, Bochum, 31 March – 2 April 2004. London: Taylor & Francis. Gaudin, C., Bienen, B. & Cassidy, M.J. 2009. Centrifuge experiments investigating the use of jetting in spudcan extraction. In Proc. 19th Intern. Offshore (Ocean) and Polar Engineering Conference & Exhibition (ISOPE), Osaka, Japan. Merifield, R.S., Lyamin, A.V., Sloan, S.W., andYu, H.S. 2003. Three-dimensional lower bound solutions for Stability of Plate Anchors in Clay. ASCE Journal of Geotechnical and Geoenvironmental Engineering 129(3): 243–253. Purwana, O.A., Leung, C.F., Chow, Y.K., and Foo, K.S. 2005. Influence of base suction on extraction of spudcan foundations. Géotechnique 55(10):741–753. Purwana, O.A. 2006. Centrifuge model study on spudcan extractions in soft clay. PhD Thesis. The National University of Singapore. Purwana, O.A., Quah, M., Foo, K.S., Leung, C.F., and Chow, Y.K. 2008. Understanding spudcan extraction problem and mitigation device. In Proc. 2nd Jackup Asia Conf., Singapore, 17–18 November 2008. Purwana, O.A., Quah, M., Foo, K.S., Nowak, S., and Handidjaja, P. 2009. Leg Extraction/Pullout Resistance – Theoretical and Practical Perspectives. In Proc. 12th Jackup Conf. London, 15–16 September 2009. City University: London. WAMIT Incorporated. 2006. WAMIT User Manual 6.3. Massachusetts.

684

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

3D FE analysis of the installation process of spudcan foundations G. Qiu, S. Henke & J. Grabe Institute of Geotechnical and Construction Engineering, Hamburg University of Technology, Germany

ABSTRACT: The risk of rapid uncontrolled penetration (“punch-through”) exposes jack-ups to significant risk during installation in strong over weak layered soils. An example for this is a thin sand layer overlying a weaker stratum of clay. Contact problems and large mesh distortions, which lead to convergence problems in FEsimulations may occur if this problem is simulated using conventional FE-method. In recent years, the Coupled Eulerian-Lagrangian method (CEL) has been used to solve geomechanical boundary value problems involving large deformations. In this paper, three dimensional finite element analyses using the CEL formulation are carried out to simulate the installation process of spudcan foundations. A Lagrangian mesh is used to discretize the spudcan foundation, whereas an Eulerian mesh is used for the subsoil. The clay is modeled as elasto-plastic DRUCKER-PRAGER material and a hypoplastic constitutive law is used to describe the sand. The numerical results are compared with existing analytical solutions and centrifuge model test data. 1

INTRODUCTION

Table 1.

Jack-up rigs are an integral part of the offshore oil and gas fleet. Sometimes the soil conditions are difficult if a thin sand layer overlies a weaker clay stratum. If the bearing capacity of the soil is overestimated rapid uncontrolled penetration (“punch-through”) can happen. This may lead to severe damages or even loss of the jack-up rig. Therefore, it is necessary to predict the bearing capacity of the spudcan footing very accurately. In Meyerhof (1974) an analytical method and design charts are presented to predict the bearing capacity of a footing on sand overlying clay. Further investigations using centrifuge tests are presented in Craig and Chua (1990), Hossain et al. (2005), White et al. (2008), Teh et al. (2008) and Lee (2009). Nowadays, the use of numerical methods may be a helpful tool to predict the bearing capacity of spudcan foundations. First numerical calculations where a Coupled Eulerian-Lagrangian approach is used to model spudcan penetration can be found in Tho et al. (2009). In this paper the CEL method is used to simulate the spudcan penetration into uniform clay or sand respectively and into layered soil with a thin sand layer overlying weaker clay. The numerical results are compared to centrifuge test results and analytical considerations to verify the numerical method and to show its suitability to predict the bearing capacity of spudcan foundations. 2

NUMERICAL MODEL

A three-dimensional finite element model is presented to investigate the installation process of spudcans. The use of classic FE-software often leads to contact problems and distortion of the mesh. To deal with these problems the finite element calculations

685 © 2011 by Taylor & Francis Group, LLC

Summary of numerical simulations.

Case

Numerical Model (Fig. 2)

Clay cu [kPa]

C1 C2 C3 S1 SC1 SC2

(a) (a) (a) (a) (b) (b)

39 63 87 – 42 42

Sand ϕc [ ◦ ]

ID [–] – – –

31.5 31.5 31.5

0.8 0.25 0.8

are carried out using the Coupled Eulerian-Lagrangian method (CEL) implemented in the commercial program Abaqus 6.8-EF. All numerical simulations are summarized in Table 1. 2.1

Coupled Eulerian-Lagrangian Method

If a continuum deforms or flows, the position of the small volumetric elements changes with time. These positions can be described as functions of time in two ways: • Lagrangian description: the movement of the con-

tinuum is specified as a function of the material coordinates and time. • Eulerian description: the movement of the continuum is specified as a function of the spatial position and time. In simulations with Lagrangian formulation the interface between two parts is precisely defined and tracked. In these simulations large deformation of a part often leads to hopeless mesh and element distortions. In Eulerian analyses an Eulerian reference mesh, which remains undistorted and does not move,

is required to trace the motion of the particles. The main advantage of an Eulerian formulation is that no element distortions occur. Disadvantageously, the interface between two parts cannot be described as precisely as if a Lagrangian formulation is used. Numerical diffusion can happen during the simulation. A Coupled Eulerian-Lagrangian (CEL) method, which attempts to capture the strength of both the Lagrangian and the Eulerian method, is implemented in Abaqus/Explicit. For general geotechnical problems, a Lagrangian mesh is used to discretize structures, whereas an Eulerian mesh is used to discretize the subsoil. The interface between structure and subsoil can be represented using the boundary of the Lagrangian domain. The Eulerian mesh, which represents the soil that may experience large deformations, is able to overcome problems like mesh and element distortions in finite element simulations. Several benchmark calculations in Qiu et al. (2009) show that the CEL approach is well suited to solve numerical problems involving large deformations which cannot be solved satisfactorily using classical finite element method. 2.2

Geometry and mesh

The geometry of the spudcan is shown in Fig. 1. A spudcan with the same geometry was used by Yu et al. (2009) to investigate the spudcan penetration into loose sand over uniform clay using the RITSS method. In this paper the spudcan is modeled as a discrete rigid body. The penetration of the spudcan into the soil is simulated displacement controlled with a constant penetration velocity. High penetration velocity results in a noisy solution while low penetration velocity leads to long simulation time. After a set of preliminary calculation with the velocities of 0.5 m/s, 1 m/s and 2 m/s the penetration velocity 1 m/s is selected for the following simulations to balance accuracy and efficiency of the simulation. The depth of penetration d is defined as zero after the cone has completely penetrated into the soil (see Figure 1). The soil is modeled as an Eulerian domain. 3D Eulerian elements (element typ: EC3D8R), which are the only available Eulerian elements in Abaqus, are used to discretize the soil. Thus, the axi-symmetric boundary value problem must be simulated in a 3D model. Due to symmetry only one fourth of the whole model is considered in the three-dimensional analysis. Two numerical models are generated to investigate the penetration process of a spudcan into both uniform and stratified deposits (see Figure 2). 2.3

Figure 1. Geometry of the spudcan; unit [m].

Contact formulation

The contact between the soil material in the Eulerian domain and the spudcan meshed with Lagrangian elements is described using the “general contact” algorithm, which enforces the use of the penalty contact method. This contact method works without discrete contact elements. The “general contact” algorithm

© 2011 by Taylor & Francis Group, LLC

Figure 2. Numerical model for spudcan penetration into (a) a uniform sand or clay bed and (b) a sand layer over uniform clay.

with the finite-sliding formulation, which allows arbitrary motion of the surface, is well suited to simulate highly nonlinear processes involving large deformations.The surface of the spudcan is modeled as smooth. Therefore, the friction coefficient between spudcan and soil is set to zero. 2.4

Constitutive models

The uniform clay layer is simulated using a Drucker-Prager material. The friction angle β and the dilation angle in the p-t plane are set to 0◦ . By matching the plane strain response we can derive the relationship:

where cu is the undrained shear strength of the clay; dc is the shear strength in the p-t plane. The clay is assumed to be undrained and the Poisson’s ratio ν is taken as 0.49. A constant stiffness ratio of E/cu = 500 is assumed for all calculations in this paper, where E is the Young’s modulus of clay. The stiffness ratio is within the range commonly adopted for soft clays. The sand layer is modeled using the hypoplastic constitutive law. The theory of hypoplasticity was

686

Table 2.

Hypoplastic parameters of Mai-Liao sand.

Parameters

Value

ϕc hs n ed0 ec0 ei0 α β

[◦ ] [MPa] [–] [–] [–] [–] [–] [–]

mT mR R βR χ

[–] [–] [–] [–] [–]

31.5 32 0.324 0.57 1.04 1.20 0.40 1.00 2.0 5.0 0.0001 0.5 6.0

developed at the University of Karlsruhe in Germany, especially by Kolymbas (1991) and Gudehus (1996). The calculations in this paper are based on the version of von Wolffersdorff (1996) with the extended concept of intergranular strain (Niemunis and Herle 1997). Hypoplasticity is well suited to model the nonlinear and anelastic behavior of dry granular soils. Typical soil characteristics like dilatancy, contractancy, different stiffnesses for loading and unloading and the dependency of stiffness on pressure and void ratio can be simulated. The hypoplastic parameters for the sand layer are those of Mai-Liao Sand shown in Table 2. The soil density varies between a loose (ID = 0.25) and a dense state (ID = 0.8). 3

Figure 3. Bearing pressure response of a spudcan penetration into uniform clay; comparison of FE and centrifuge test results.

influence of the spudcan geometry is very small after 4 m penetration. Figure 4 shows the velocity field of the FE simulations for case C1 at three different penetration depths. Similar flow mechanisms and cavity formation can also be oberserved in Hossain et al. (2005). For small penetration of the spudcan into the soil (Figure 4(a), d/D = 0.09) the clay around the spudcan edge moves upwards and heaves. After 3.5 m deeper penetration (Figure 4(b), d/D = 0.34), the clay around the spudcan edge begins to flow back onto the exposed top of the spudcan. It is difficult to indicate the movement of the clay near the surface. This mechanism is referred to as “flow failure” (Hossain et al. 2005). After the spudcan penetrates 16.8 m into the clay (Figure 4(c), d/D = 1.2), no cavity “wall failure” is observed here as well. According to Hossain et al. (2005) the stable cavity depth H /D can be calculated to

PENETRATION INTO UNIFORM CLAY AND SAND

In this section centrifuge tests carried out by Craig and Chua (1990) are recalculated with CEL method. To overcome difficulties in the numerical simulations the model spudcan shown in Figure 1 has a sharper cone and smaller side compared to the one used in the centrifuge tests. The numerical results are compared with the test data in Figures 3 and 5. Due to the differences in geometry the numerically received bearing pressures are smaller than the results of the centrifuge tests. 3.1 Penetration into uniform clay The tests of penetration into uniform clay were carried out by Craig and Chua (1990) with three different undrained strengths cu = 39 kPa, 63 kPa and 87 kPa. All tests are carried out in conditions where no free surface water is taken into account. The unit weight of clay is set to γc = 20 kN/m3 in the numerical simulations. The bearing pressures increase with increasing depth of penetration. This matches well with the centrifuge tests. In each case the FEM solutions (case C1, C2 and C3) lie below the test results, especially in the first 3 m of penetration. After 4 m penetration, the numerical results match well the test results. Respectively, the

© 2011 by Taylor & Francis Group, LLC

For case C1 the stable cavity depth H /D = 0.34. This matches well with the FE results where d/D is 0.34 (Figure 4(b)). 3.2

Penetration into uniform sand

The results of a centrifuge test of spudcan penetration into a sand bed are shown in Figure 5. This test was carried out with free water table above the sand surface. The effective unit weight of sand is set to γs
= 11 kN/m3 in the numerical simulation. In this analysis the spudcan penetrates into dense sand with ID = 0.8. After about 1.5 m penetration the bearing

687

Figure 6. Velocity field after 6.8 m of spudcan penetration into uniform sand (case S1).

pressure increases linearly with increasing penetration depth. The initial friction angle of the Mai-Liao sand in the numerical simulation is ϕc = 31.5◦ . The bearing pressures observed in the centrifuge test are higher compared to the numerical results. This can be explained by the higher friction angle (37.5◦ –38.5◦ ) of the sand used in the centrifuge. The velocity field after 6.8 m penetration into uniform sand is shown in Figure 6. Comparing Figure 4(b) and Figure 6 it can be seen that the cavity, which is punched into uniform clay, remains open.This matches very well with observations after (Craig and Chua 1990) in the absence of free water. By contrast, “wall failure” can be observed in Figure 6. After penetration into sand the cavity inside the uniform sand becomes unstable. The sand at the surfaces falls down into the cavity, such that the wall collapses and refills the cavity above the spudcan. Figure 4. Velocity field of spudcan penetration into uniform clay (case C1) (a) d/D = 0.09; (b) d/D = 0.34; (c) d/D = 1.2.

4

PENETRATION INTO SAND OVER UNIFORM CLAY

In this section numerical simulations of a spudcan punching through a loose sand layer (ID = 0.25) or a dense sand layer (ID = 0.8) over clay are described.The undrained shear strength of the clay is uniform, with cu = 42 kPa. The numerical model with its dimensions is depicted in Figures 1 and 2(b). 4.1

Figure 5. Bearing pressure response of a spudcan penetrating in uniform sand (ID = 0.8); comparison of FE results, analytical predictions and centrifuge test results.

688 © 2011 by Taylor & Francis Group, LLC

Numerical results

In Figure 7 the penetration resistance of the spudcan with respect to different soil densities for the overlying sand layer is depicted. After about 2 m of penetration the maximum penetration resistance of about 650 kN/m2 is reached in case SC2 and remains nearly constant with further penetration of the spudcan. Regarding the results of the simulation with a loose sand layer over clay the penetration resistance increases continuously with penetration depth. The penetration resistance is higher regarding the simulation with the dense sand layer (case SC2) compared to the simulation with loose sand layer (case SC1). Nevertheless, the penetration resistances converges for both investigated configurations

Figure 9. Distribution of the materials sand and clay after different penetration depths of the spudcan (case SC2).

Regarding Figure 8 it can be seen that after 1.8 m of penetration the soil mainly moves downward below the spudcan. Lateral movement of the soil can hardly be noticed. After 6.8 m of penetration the spudcan cone passed punches into the clay and the sand beside the spudcan flows over the spudcan. Nevertheless, the main deformation of the soil below the spudcan is a vertical motion. This indicates that a “sand plug” under the spudcan is pushed into the softer clay layer. This can also be seen in Figure 9. In Figure 9 the distribution of the materials sand and clay for different penetration depths is depicted and it can be seen that the “sand plug” can be found directly below the spudcan foundation and moves downward together with it. This result is in good agreement with calculations of Yu et al. (2009) or centrifuge tests in Lee (2009).

Figure 7. Penetration resistance of the spudcan for the different simulated configurations.

4.2

Comparison with conventional analyses

The results of the numerical simulations are compared to analytical results using the proposed procedure in Meyerhof (1974). This approach is suited to estimate the ultimate bearing capacity of a footing punching through a thin sand layer into a thick clay bed. The ultimate bearing capacity qu depends on the thickness of the sand layer below the spudcan H , the penetration depth d, the diameter D of the footing, and the soil parameters ϕ, γ and cu and is depicted in equation 3.

Figure 8. Velocity field inside the soil continuum after 1.8 m and 6.8 m penetration into clay under dense sand layer (case SC2).

with higher penetration depth. This is feasible because the underlying clay is simulated with the same soil parameters in both analyses. In Figure 8 the velocity fields after 1.8 m and 6.8 m of penetration into the dense sand layer are shown for better understanding of the mechanisms during the spudcan penetration process. © 2011 by Taylor & Francis Group, LLC

The friction angle ϕ for the dense sand layer is calculated to be approximately 36.7◦ using a numerical triaxial test simulation with regard to the occuring stresses, such that the coefficient sKs can be estimated to 5.5 according to Meyerhof (1974). The results of equation 3 for the investigated case are compared to the numerical results in Figure 7. With regard to Figure 7 it can be stated that the Coupled Eulerian-Lagrangian solution agrees quite well with the analytical solution after Meyerhof (1974). The approach after Meyerhof (1974) slightly underestimates the maximum bearing capacity received from

689

the numerical study. This might be caused by the different shape of the spudcan compared to a simple circular footing. Furthermore the shape factor s in the analytical considerations is estimated conservatively to be s = 1.0. With further penetration the results are diverging. The Meyerhof curve decreases due to further penetration depth and therefore smaller sand layer thickness below the spudcan. Influence of the lateral embedment are not taken into account such that the numerical results must lead to higher bearing capacities. 5

CONCLUSIONS

In this paper, 3D FE-analyses using the CEL method are carried out to simulate the installation process of spudcan foundations into different ground profiles. In a first numerical study centrifuge tests of spudcan penetration into a uniform clay or sand bed are recalculated. The recalculated bearing pressures using CEL method are in agreement with the results from the centrifuge tests. The holes formed due to the penetration of the spudcan into the two different soils were also observed in centrifuge tests. Two different failure mechanisms (“flow failure” and “wall failure”) are observed between penetration into clay and sand. In further calculations the penetration of spudcans into a layered soil with a small sand layer overlying a weaker clay is investigated. The initial density of the overlying sand is varied in FE model. The results are compared to zentrifuge test and show good agreement. Therefore it can be stated that the presented numerical model is well suited to simulate ”punchthrough” processes. In future research work, this numerical model has to be validated by further measurement data. If successful, it will be a promising tool to investigate different aspects of spudcan penetration. It would be possible to investigate the influence of spudcan shape on penetration resistance for example. Furthermore, it might allow the engineer to receive a deeper insight into the mechanisms that occur inside the soil during spudcan penetration.

© 2011 by Taylor & Francis Group, LLC

REFERENCES Craig, W. and K. Chua (1990). Deep penetration of spudcan foundation on sand and clay. Gèotechnique 10(4), 541–556. Gudehus, G. (1996). A comprehensive constitutive equation for granular materials. Soils and Foundations 36(11), 1–12. Hossain, M., Y. Hu, M. Randolph, and D. White (2005). Limiting cavity depth for spudcan foundations penetrating clay. Gèotechnique 55(9), 679–690. Kolymbas, D. (1991). Computer-aided design of constitutive laws. International Journal for Numerical and Analytical Methods in Geomechanics 15, 593–604. Lee, K. (2009). Investigation of Potential Spudcan PunchThrough Failure on Sand Overlying Clay Soils. PhD Thesis, University of Western Australia, Centre for Offshore Foundation Systems. Meyerhof, G. (1974). Ultimate bearing capacity of footings on sand overlying clay. Canadian Geotecnical Journal 11(2), 223–229. Niemunis, A. and I. Herle (1997). Hypoplastic model for cohesionless soils with elastic strain range. Mechanics of frictional and cohesive materials 2(4), 279–299. Qiu, G., S. Henke, and J. Grabe (2009). Applications of coupled eulerian-lagrangian method to geotechnical problems with large deformations. pp. 420–435. Proceeding of SIMULIA Customer Conference 2009, London, UK. Teh, K., M. Cassidy, C. Leung, Y. Chow, M. Randolph, and C. Quah (2008). Revealing the bearing capacity mechanisms of a penetrating spudcan through sand overlying clay. Gèotechnique 58(10), 793–804. Tho, K., C. Leung,Y. Chow, and S. Swaddiwudhipong (2009). Application of eulerian finite element technique for analysis of spudcan and pipeline penetration into the seabed. 12th International Jack-up Conference, London. von Wolffersdorff, P.-A. (1996). A hypoplastic relation for granular material with a predefined limit state surface. Mechanics of cohesive-fractional materials 1, 251–271. White, D., K. Teh, C. Leung, and Y. Chow (2008). A comparison of the bearing capacity of flat and conical circular foundations on sand. Gèotechnique 58(10), 781–792. Yu, L., Y. Hu, and J. Liu (2009). Spudcan penetration in loose sand over uniform clay. Proc. of 28th Int. Conf. on Ocean, Offshore and Arctic Engineering, Hawaii, USA.

690

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Undrained bearing capacity of deeply embedded foundations under general loading Y. Zhang, B. Bienen, M.J. Cassidy & S. Gourvenec Centre for Offshore Foundation Systems, The University of Western Australia, Perth, Australia

ABSTRACT: The foundations of offshore mobile jack-up drilling rigs are subject to combined vertical (V ), horizontal (H ) and moment (M ) loading conditions. In soft soil, spudcans penetrate deeply (up to 2.5 diameters) and soil flows back on top of the foundations. This is generally believed to provide additional bearing capacity. However, the current industry guideline evaluates bearing capacity based on shallow foundation mechanisms regardless of the spudcan penetration, and the potential benefits associated with deep failure mechanisms are unaccounted for. This results in conservatism within site-specific assessments. In this study, the combined undrained bearing capacity of a deeply embedded spudcan foundation in normally consolidated clay is investigated using small strain finite element analyses. The effects of footing shape and aspect ratio on the failure mechanisms and combined bearing capacity are highlighted. To conclude, a closed-form expression is proposed to describe the shape of the VHM failure envelopes. 1

INTRODUCTION

Mobile jack-up drilling rigs are widely used for offshore drilling activities in water depths up to 150 m. Before a jack-up rig can be mobilized offshore, a sitespecific assessment must demonstrate their spudcan footings’capacity to withstand loading associated with the 50-year return period storm (SNAME 2008). In many offshore areas, such as regions of the Gulf of Mexico, the seabed soil consists of normally consolidated or lightly over-consolidated clay (Menzies & Roper 2008). In these soils, spudcans can penetrate several diameters, with soil flowing back on top of the footing sealing the cavity. At sufficient penetration the failure mechanisms do not extend to the soil surface but remain localized (as shown for vertical loading in Hossain & Randolph 2009). Current practice, however, assesses the footing’s capacity based on shallow foundation mechanisms (SNAME 2008) without any additional bearing capacity due to the deep mechanisms. Shallow mechanisms are also considered in a current state-of-the-art foundation-soil interaction model for spudcans on clay (Martin & Houlsby 2001). This plasticity based model utilizes a displacement hardening yield surface that is written directly in combined vertical (V ), horizontal (H ) and moment (M ) loading and is based on a series of model experiments (Martin & Houlsby 2000). As these tests were performed in heavily over-consolidated clay at 1g, soil back-flow was not observed. Cassidy et al. (2004) performed similar tests in normally consolidated clay in the centrifuge, achieving full back-flow, and found Martin & Houlsby’s (2001) envelope to be conservative in those conditions. However, no alternative

expression was proposed. Templeton et al. (2005) and Templeton (2009) studied the pure moment and horizontal capacities of a buried spudcan using finite element analyses (FEA). While expressions for moment and horizontal capacities were suggested, coupled VHM interaction was not considered. On the other hand, combined loading solutions exist for deeply buried anchors of various plan shapes (O’Neill et al. 2003, Elkhatib 2005). However, applicability of these findings to footing shapes such as spudcans has not been investigated. This paper presents results of a numerical investigation of the combined VHM bearing capacity of a deeply embedded spudcan in undrained normally consolidated (NC) clay by means of small strain FEA. The effects of the spudcan’s shape and aspect ratio on the combined bearing capacity are explored through comparison with circular plate footings of two different aspect ratios. 2

The FEA were carried out with the commercial software Abaqus V6.7 (Dassault Systèmes, 2007). 2.1

Footings

Figure 1 shows the adopted spudcan geometry. Rather than modeling a specific design, it represents a generic spudcan currently used offshore. The spudcan is 18 m in diameter (D), with an overall aspect ratio (T /D) of 0.40. The top and bottom slope angles are 30◦ and 15◦ respectively. The top cone and upstand account for 70% of the total height, with the remaining 30%

691 © 2011 by Taylor & Francis Group, LLC

METHODOLOGY

Figure 2. Close-up of the mesh around the spudcan. Figure 1. Spudcan geometry adopted for FEA.

the bottom cone. The jack-up’s truss work leg is not simulated, with the focus solely on the capacity of the spudcan. Two circular plate footings, with the same footing diameter as the spudcan and aspect ratios of T /D = 0.05 (comparable with the spudcan’s upstand) and T /D = 0.40 (equal to the overall aspect ratio of the spudcan), were also investigated. 2.2

Soil

The soil modeled is normally consolidated (NC) clay with a linearly increasing shear strength profile with depth (k) of 1.2 kPa/m. The mudline strength (sum ) is idealized to be zero (though a small non-zero value, 0.1 kPa, was used for numerical reasons), giving a normalized profile kD/sum approaching infinity. The soil was assigned an effective unit weight (γ
) of 6 kN/m3 . The soil is modeled as linearly elastic, perfectly plastic governed by the Tresca failure criterion, which is implemented in 3D FEA by adopting a rounded yield surface at the vertex in the deviatoric plane. This implementation may result in slightly lower results, with the influence depending on the rounding angle (Taiebat & Cater 2008). In order to reduce computational cost, E/su was chosen to be 10,000. Though this is significantly higher than the values used in other numerical studies in soft clay (Bransby & Randolph 1998, Taiebat & Carter 2000, Gourvenec & Randolph 2003), it did not impact the results as only the ultimate footing capacity was of interest. 2.3

Mesh

Due to symmetry, a semi-cylindrical three dimensional model was used. The vertical and horizontal boundaries were placed 2.5D from the footing edge, sufficiently remote so as not to affect the results. At the horizontal boundaries of the model no horizontal translation was permitted while the model base was fixed in all three coordinate directions. The mesh was initially developed and benchmarked for the thin circular plate as no exact solutions are known for the spudcan. After accuracy and convergence of the solution were shown (as detailed in 2.7 Benchmarking), the footing shape was amended to represent the spudcan while the © 2011 by Taylor & Francis Group, LLC

principles of the mesh construction were upheld. Figure 2 shows the mesh around the spudcan footing on the vertical diametrical face. The model consists of about 43,000 first order hexahedral (C3D8) elements. The footing-soil contact is assumed to be fully rough in shear and fully bonded in tension. 2.4

Load reference points

The footings were modeled as rigid bodies with loads and displacements related to a single load reference point (LRP). The choice of LRP affects the shape of the envelope in the HM plane through cross-coupling of the independent degrees of freedom (as shown in Fig. 4). For the spudcan, the LRP is usually chosen at the lowest maximum diameter (LRP1, Fig. 1). For the plate footings, the LRP is usually chosen at the footing base. For the purpose of making direct comparison of the results of the three footings, which requires a consistent normalization among the footings, another LRP at the footing base (LRP2, Fig 1) is defined for the spudcan. 2.5

Load paths

All footings were ‘wished in place’, i.e. installation was not modeled. The embedment depth was 3.5D (measured from the mudline to footing base) to achieve localized failure mechanisms for all three footings considered. The combined bearing capacity surfaces were established by combinations of constant vertical displacement “swipe” tests and constant displacement ratio “probe” tests. In the VM loading plane (H = 0) swipe tests were used. A vertical displacement was initially applied until ultimate vertical load was achieved. The footing was then rotated while keeping the vertical displacement constant. The load path followed in the swipe test tracks the failure envelope of the footing under VM loading. This method has been successfully used in FEA by, for example, Bransby & Randolph (1998) and Gourvenec & Randolph (2003). However, swipe tests have been found to undercut the failure envelopes in the VH (M = 0) and HM (V = 0) planes (discussed in Gourvenec 2008). In these two planes, constant ratio displacement probes, as described in detail by Bransby & Randolph (1998), were used. In

692

Table 1.

Summary of notations used in this paper.

Load Ultimate load Maximum load Bearing capacity factor

Vertical

Horizontal

Rotational

V Vult – NcV = Vult /Asu0

H Hult Hmax NcH = Hult /Asu0

M Mult Mmax NcM = Mult /ADsu0

Table 2.

Spudcan∗ (Spudcan∗∗ PlateT /D=0.05 PlateT /D=0.40 ∗

A : maximum bearing area, A = 0.25πD2 . su0 : undrained shear strength at the load reference point level.

combinedVHM loading space, a proportion of ultimate vertical load (0, 0.25, 0.50, 0.75, 0.90) was applied and held constant, while the footing was taken to failure under different ratios of horizontal displacement and rotation. These procedures allow a complete VHM envelope to be constructed. 2.6 Sign conventions and notations The sign convention adopted in this paper is shown in Figure 1. Table 1 summarizes the notations used in this paper. Due to the cross-coupling of the footing’s horizontal and rotational degrees-of-freedoms, the footing’s maximum horizontal and moment loads are different from the ultimate loads and are mobilized when accompanied by a moment or horizontal force respectively. 2.7 Benchmarking The accuracy of the FE model was demonstrated through comparison with known solutions in homogeneous soil. Thus, for the purpose of benchmarking su = const. was assumed. For a rough, infinitely thin, deeply buried circular plate in homogeneous soil, Martin & Randolph (2001) gave the exact solution for NcV based on bound theorems of 13.11. The result obtained from this model for the thin plate in homogeneous soil was 13.23. For the moment capacity, Elkhatib (2005) provided an upper bound solution NcM of 1.57 based on a spherical scoop mechanism. The result from this model was 1.63. Because the plate is very thin (T /D = 0.05), the sidewall contribution to vertical or the moment capacity was considered negligible. The differences between the FE results and the theoretical solutions are 0.91% and 3.82% for vertical and moment capacity respectively. The above comparisons demonstrate good accuracy of the FE model compared to known solutions. Note that results in the following were obtained with linearly increasing (NC) soil strength profile. 3 3.1

RESULTS Uniaxial bearing capacity

3.1.1 Uniaxial bearing capacity of the spudcan Table 2 summarizes the values of uniaxial bearing capacity factors for the spudcan as well as the circular

NcV

NcH

NcM

Hmax /Asu0 Mmax /DAsu0

12.99 12.54 13.06 15.50

4.93 4.31 3.41 7.23

1.61 1.55 1.61 1.99

5.21 5.03 3.43 9.61

1.63 1.78) 1.63 3.15

LRP1 (Fig. 1), ∗∗ LRP2 (Fig. 1).

plate footings. The values at the two LRPs are presented for the spudcan. The differences in the values of NcV = Vult / Asu0 , NcM = Mult /DAsu0 and Hmax /Asu0 between the two LRPs are purely caused by the different su0 values used for normalization as their mechanisms are independent of the position of the LRP. The same is not the case for NcH = Hult /Asu0 and Mmax /DAsu0 . The results from the FEA suggest significantly higher vertical bearing capacity than methods recommended based on shallow failure mechanisms in the current guidelines, SNAME (2008). The alternative bearing capacity factor method (later published by Houlsby & Martin, 2003) gives a value of 8.6 for NcV , which is 34% lower than the value in Table 2 (at LRP1). Based on large deformation FEA, Hossain & Randolph (2009) suggested the vertical bearing capacity factor for a deeply penetrated spudcan approaches a limiting value of 11.3 regardless of footing roughness or soil non-homogeneity. This was for a spudcan with very thin upstand and is not too dissimilar to the result of this current study. The reduction may be due to their use of large deformation FEA. Softer soil is carried down under the penetrating footing, producing a lower bearing capacity factor. SNAME (2008) recommends ML0 to be 0.10DVult and HL0 to be the friction on the footing base, including the effect of the laterally projected area below the widest diameter (ML0 and HL0 are defined as peak moment and horizontal capacities respectively in the VM and VH loading planes). Martin & Houlsby (2001) provided values of HL0 /Vult = 0.127 and ML0 /DVult = 0.083 based on 1g tests in heavily over consolidated soil with no soil back-flow. Compared with these values, results of this study (Hult /Vult = 0.380, Mult /DVult = 0.124) show that the capacity recommendations based on shallow foundation mechanisms are conservative. Soil back-flow increases the spudcan’s horizontal and moment capacities significantly. For deeply buried spudcans, Templeton et al. (2005) suggested ML0 /DVult = 0.175 (compared to 0.124 in this study) and related HL0 to the ratio of spudcan’s horizontal to vertical projected areas (Templeton 2009). For the spudcan area ratio of 0.31 (different from the aspect ratio T /D) adopted here, Templeton’s expression gives a value of 0.46, higher than the value of 0.38 obtained in this study. One of the reasons for the differences may be due to the coarse mesh of the FE model on which Templeton’s recommendations were mainly based.

693 © 2011 by Taylor & Francis Group, LLC

Normalised uniaxial bearing capacities.

Figure 3. Comparison of uniaxial mechanisms through displacement contour plots (LRPs for all footings at footing base).

Soil remolding due to the installation process affects the footing’s bearing capacity. The results presented here (and in Templeton et al. 2005, Templeton 2009) apply to undisturbed conditions. 3.1.2 Comparison with circular plate footings As shown inTable 1, the spudcan’s uniaxial vertical and moment capacities are similar to those of the thin plate, whose height is comparable to the spudcan’s upstand. This is due to the similar failure mechanisms mobilized as Figure 3 illustrates. For an infinitely thin footing under uniaxial horizontal load, the mechanism should be pure sliding at the footing soil interface. Therefore, Hult /Asu0 (NcH ) and Hmax /Asu0 should be synonymous and equal to 2. However, as the footing aspect ratio increases, passive soil resistance at the sidewall constitutes a significant proportion of the horizontal capacity. The spudcan’s ultimate and maximum horizontal capacities are larger than those of the thin plate footing because the spudcan is taller overall, but it is much smaller than that of the thick plate of equal aspect ratio. This is due to the shape of the spudcan. The conical shape of the spudcan results in a smaller rotational scoop than associated with the thick plate, and therefore less horizontal capacity (see Fig. 3).

the lowest maximum diameter (LRP1, Fig. 1) and the dashed line to LRP2. Both are shown as LRP1 is most commonly used for spudcan; whereas LRP2 allows direct comparison between the spudcan and the circular plate footings in section 3.2.2. The eccentricity of the envelope in the HM plane increases significantly as the LRP is moved from position LRP1 to LRP2. It is observed that the peak of envelopes in the VH and VM planes are at a vertical load of zero, i.e. Hult = HL0 , Mult = ML0 respectively, different from those reported in Martin & Houlsby (2000). This is because full tensile capacity (due to presence of full soil backflow) and fully rough footing soil contact are assumed for the deeply buried spudcan.

3.2 Combined VH, VM and HM bearing capacity 3.2.1 Failure envelopes for the spudcan footing Figure 4 presents the failure envelopes derived for the spudcan footing in the VH, VM and HM planes. The bold lines correspond to the typical LRP location at

© 2011 by Taylor & Francis Group, LLC

694

3.2.2 Comparison with circular plate footings Figure 5 compares the envelopes of the spudcan and plate footings of aspect ratios of 0.05 and 0.40. For a consistent comparison, the spudcan envelopes shown in Figure 5 are for LRP2 (Fig. 1). The spudcan behaves significantly differently to a plate footing of the same aspect ratio (T /D = 0.40). In all planes, the envelopes for the spudcan lies considerably inside of those for the T /D = 0.40 plate and closer to the envelopes for the thin plate (T /D = 0.05). In the VM plane, the envelope of the spudcan falls slightly within that of the thin plate. The two curves also have roughly the same shape. In the HM plane, both the size and the shape of the envelope of the spudcan differ from those of the plate footings. While the envelope for the thin plate is almost symmetric about the M axis (it would be fully symmetric, with no crosscoupling, for an infinitely thin plate), the envelope for

Figure 4. Failure envelopes of the spudcan footing.

Figure 5. Comparison of failure envelopes between spudcan (LRP2) and circular plates.

the thick plate is highly eccentric. The eccentricity of the spudcan footing’s envelope lies in between.

(Martin & Houlsby 2000, 2001) as well as for deeply embedded anchors (O’Neill et al. 2003, Elkhatib 2005). Written directly in terms of the loads on the footing, the complete expression has the following form:

3.3 Combined VHM bearing capacity 3.3.1 Failure envelopes Figure 6 presents the spudcan’s bearing capacity under combined VHM loading in the form of failure envelopes in the HM plane at different vertical load levels. Figure 6a shows the envelopes in the form of bearing capacity factors, while Figure 6b shows the envelopes normalized by Vult (for comparison with similar envelopes of spudcan footings, as discussed in 3.1). It can be seen that the envelopes are slightly eccentric. At low vertical load, the envelope size remains fairly constant. As V /Vult increases beyond 0.50, however, the envelope reduces in size. 3.3.2 Curve fitting An expression to describe the failure envelopes for the spudcan as well as the plate footings under VHM combined loading has been developed. This follows similar expressions for surface flat footing (Taiebat & Carter 2000), and for spudcans shallowly embedded in clay

where variables ah and am are the exponents of the H and M terms. The parameter e determines the eccentricity of the envelopes in the HM plane and has the form e = e1 + e2 v 2 , v = V /Vult , e1 , e2 vary with the footing. The factors c1 and c2 are employed to adjust the shape in the VH and VM planes, c2 = 1 − c3 (v 3 − v4 ), c1 = c2 for spudcan, c1 = 1 for plate footings. Compared to the expression for spudcans by Martin & Houlsby (2000, 2001) the vertical load term of Eq. 1 is simplified and raised to the power of 2. The fitting parameters for all footings are summarized in Table 3. Figure 6 shows the fit of the yield surface for the spudcan footing compared to the numerical results. The quality of fitting is considered good.

695 © 2011 by Taylor & Francis Group, LLC

acknowledged. The third author acknowledges the support of the Australian Research Council through his Future Fellowship. REFERENCES

Figure 6. Failure envelopes under VHM loading (LRP1). Table 3.

Fitting parameters all three footings.

Spudcan PlateT /D=0.05 PlateT /D=0.40

ah

am

c1

c3

e1

e2

2.5 4 2

1.5 1.5 1.5

c2 1 1

1.5 1.5 3.0

0.10 0.02 0.63

0.10 0 0.10

As c2 is not an independent variable (c2 = 1 − c3 (v3 − v4 )), it is not listed in the table.

4

CONCLUSIONS

This paper presents a comprehensive suite of solutions for a deeply embedded spudcan in normally consolidated soil with complete back-flow under undrained general loading conditions. The results of this study highlight the additional bearing capacity available of a deeply embedded spudcan compared to shallow foundations. The effects of the spudcan shape and aspect ratio on combined load bearing capacity are discussed. Expressions that fit the normalized envelopes in VHM loading space for the spudcan and circular plate footings are proposed. ACKNOWLEDGEMENTS The funding support from the second author’s UWA Research Development Award is gratefully

Bransby, M.F. & Randolph, M.F. 1998. Combined loading of skirted foundations. Géotechnique, 48(5), 637–655. Cassidy, M.J., Byrne, B.W. & Randolph, M.F. 2004. A comparison of the combined load behaviour of spudcan and caisson foundations on soft normally consolidated clay. Géotechnique, 54(2), 91–106. Dassault Systèmes, D. 2007. Abaqus analysis user’s manual, Simula Corp, Providence, RI, USA. Elkhatib, S. 2005. The behaviour of drag-in plate anchors in soft cohesive soils. PhD thesis, Uni. of Western Australia. Gourvenec, S. 2008. Effect of embedment on the undrained capacity of shallow foundations under general loading. Geotechnique, 58(3), 177–185. Gourvenec, S. & Randolph, M.F. 2003. Effect of strength non-homogeneity on the shape of failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique, 53(6), 575–586. Hossain, M.S. & Randolph, M.F. 2009. New mechanismbased design approach for spudcan foundations on single layer clay. Journal of Geotechnical and Geoenvironmental Engineering, 135(9), 1264–1274. Houlsby, G.T. & Martin, C.M. 2003. Undrained bearing capacity factors for conical footings on clay. Géotechnique, 53(5), 513–520. Martin, C.M. & Houlsby, G.T. 2000. Combined loading of spudcan foundations on clay: laboratory tests. Géotechnique, 50(4), 325–338. Martin, C.M. & Houlsby, G.T. 2001. Combined loading of spudcan foundations on clay: numerical modelling. Géotechnique, 51(8), 687–699. Martin, C.M. & Randolph, M.F. 2001. Applications of lower and upper bound theorems of plasticity to collapse of circular foundations. Proc. 10th Int. Conf. Ass. Comp. Meth. Adv. Geomech. (IACMAG), 1417–1428. Menzies, D. & Roper, R. 2008. Comparison of jackup rig spudcan penetration methods in clay. Offshore Technology Conference (OTC), Houston, Texas, OTC 19545. O’Neill, M.P., Bransby, M.F. & Randolph, M.F. 2003. Drag anchor fluke-soil interaction in clays. Canadian Geotechnical Journal, 40(1), 78–94. SNAME (Society of Naval Architects and Marine Engineers) (2008). Site specific assessment of mobile jack-up units, SNAME Technical and Research Bulletin 5-5A. 1st Ed., 3rd Revision, New Jersey. Taiebat, H.A. & Carter, J.P. 2000. Numerical studies of the bearing capacity of shallow foundations on cohesive soil subjected to combined loading. Géotechnique, 50(4), 409–418. Taiebat, H.A. & Carter, J.P. 2008. Flow rule effects in the Tresca model. Computers and Geotechnics, 35(3), 500– 503. Templeton, J.S., Brekke, J.N. & Lewis, D.R. 2005. Spud can fixity in clay, final findings of a study for IADC. 10th Jack-Up Conference. Templeton, J.S. 2009. Spudcan fixity in clay, further results from a study for IADC. 12th Jack-Up Conference.

696 © 2011 by Taylor & Francis Group, LLC

9 Anchoring systems

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Trajectory prediction for drag embedment anchors under out of plane loading C.P. Aubeny & C.-M. Chi Texas A&M University, College Station, Texas, USA

ABSTRACT: Partial failure of mooring systems for floating structures will subject drag anchors to loads having an appreciable component outside of the intended plane of loading. Partial failures of mooring systems during hurricanes in recent years have generated an interest in understanding drag anchor performance under these conditions. This paper presents a formulation for prediction of three dimensional trajectories of an anchor system subjected to an out-of-plane load component. The analysis shows that, if the self-weight of the anchor chain is neglected, the anchor chain configuration will lie always within a plane, although for general conditions of out-of-plane loading, the anchor chain will lie in an oblique plane defined by the direction of the anchor chain at the pad-eye and the mudline. Therefore, existing formulations for predicting anchor chain configuration and anchor trajectory during drag embedment for in-plane loading can readily be adapted to out-of-plane loading provided that appropriate adjustments are made to account for rotation of the coordinate frame. Illustrative examples of three-dimensional trajectories for anchor systems subjected to out-of-plane loading are presented in the paper. For the conditions simulated in the example analyses, the anchor experienced a modest amount of continued embedment following partial failure of the mooring system; however, the ultimate embedment and capacity of the anchor are much less than what would have developed if the anchor had continued in its original trajectory within the plane of intended loading.

1

2 ANCHOR CHAIN IN AN OBLIQUE PLANE

INTRODUCTION

Mooring systems for offshore floating structures, such as the semi-submersible mobile offshore drilling units (MODUs) commonly used for hydrocarbon exploration, typically comprise eight to twelve mooring lines positioned about the structure. Caissons or plate anchors may be used to secure the mooring lines; this paper focuses on situations where drag embedment anchors (DEAs) are used, but some of the general concepts presented will apply to any anchor. When a DEA is loaded as intended, the following conditions are assumed to hold: (1) the anchor line load acts in the plane containing the shank of the anchor, a condition referred to as “in-plane” loading, and (2) the anchor chain (or anchor line if a wire line is used) lies in a vertical plane. During extreme storm events, MODUs will sometimes experience a partial failure of the mooring system. The consequent movement of the MODU following a partial failure will generally invalidate both assumptions stated above. Firstly, the anchor line force will no longer lie in the plane of the shank, and the anchor will be subjected to “out-ofplane” loading. Yang et al. (2010) analyze the effect of such out-of-plane loading on anchor load capacity. Secondly, the anchor chain will no longer lie in a vertical plane; the effect of this condition on the ultimate pullout capacity of a DEA is the primary focus of this paper.

© 2011 by Taylor & Francis Group, LLC

Figure 1 shows a plan view of the problem under consideration. A global coordinate system, xg −yg −zg , may be established such that zg is the vertical coordinate, xg is in the plane of intended loading, and yg is normal to the plane of intended loading. Prior to the storm event, an in-plane condition of loading exists, φag = φ0g = 0, where the angle φ denotes the departure of the anchor chain from an in-plane condition.

699

Figure 1. Definition sketch for out-of-plane loading.

Figure 3. Anchor in plane of chain.

Figure 2. Force equilibrium for chain element.

The angles θ ag and θ 0g denote the direction of the anchor chain measured from a horizontal plane at the pad-eye and mudline, respectively. The mudline angles φ0g and θ 0g are arbitrarily prescribed boundary conditions for the problem. The pad-eye angles φag and θ ag will be computed by a procedure to be described subsequently, but for the present we will assume they have been defined. Unit vectors describing the direction of the anchor chain at the shackle and mudline can then be computed using Equations 1 and 2, respectively, as shown below:

can similarly be adapted to calculations outside the vertical plane. A coordinate system (xc , yc , zc ) can now be defined for which the chain lies in the xc −zc plane, subsequently termed the c-frame. It should be noted that the unit normal n to the plane of the chain acts in the yc direction in c-frame system. The transformation is accomplished by first rotating the xg −yg axes an angle δh about the zg axis, and then rotating an angle δv about the xc axis to create an xc −yc −zc frame. The appropriate matrix relating the global g-frame to the plane of the chain c-frame is then as follows:

Further, a plane containing the direction vectors d ag and d0g , which shall be termed the ‘c-plane’, can be defined in terms of the following cross product:

The angles δh and δv are defined in term of the unit normal vector n (components nxg ,nyg ,nzg in the global frame) as follows:

where n is a unit vector normal to the c-plane. Figure 2 shows the system of forces acting on a chain element in the c-plane, considering the equilibrium of a chain element ds, where T is line tension, Q is soil resistance acting normal to chain, F is friction, and W is chain weight per unit length. Except at very shallow depths, W is small relative to the shearing resistance of the soil and may be reasonably neglected. By virtue of Equation 3, the tensions at the pad-eye and mudline, Ta and T0 , act within the c-plane. If the self-weight W of the chain is neglected, the integrated resultants of F and Q must also act within the c-plane. Therefore, the chain can be considered to lie in the c-plane, an oblique plane normal to the unit vector n, so long as the anchor self-weight is small relative to the shearing resistance of the soil. Accordingly, the anchor line equations developed for anchor chains in a vertical plane (Neubecker & Randolph, 1995) can be adapted to the out-of-plane problem, provided appropriate coordinate transformations are made. DEA trajectory calculations derived from the Neubecker-Randolph equations (Aubeny & Chi, 2010)

Subsequent subscripts “g” and “c” in this paper refer to the global frame and the frame oriented about the plane of the chain, respectively. It should be noted that the direction of the angles follow the right-hand rule. Having defined the frame rotation angles [Rgc ], transformations of key variables from the g-frame to the c-frame can proceed. The anchor line angle at the pad-eye θ ac measured in the c-plane can be evaluated by using Equations. 4–6 to determine the vector components of da in the c-frame, dac = − [daxc dayc dazc ], from which θ ac in Figure 3 can be calculated:

700 © 2011 by Taylor & Francis Group, LLC

A similar transformation is possible for the anchor line angle at the mudline θ 0c . A revised form of the Neubecker and Randolph (1995) relationship between line tension and line angles can now be derived along the same lines as

Figure 4. Depth parameters in the plane of chain.

the original formulation for the case of an anchor line lying in an oblique plane:

where Ta is anchor line tension at shackle point, En is multiplier applied to chain bar diameter, Nc = bearing factor for wire anchor line, b = chain bar or wire diameter, su0 is the soil undrained shear strength at mudline, k is soil strength gradient with depth, and Zc is the distance from the anchor shackle to mudline in the zc direction. From Figure 4 it is apparent that Zc = zc / cos2 δv . Finally, for trajectory calculations the rate of change of θ ac with respect to the change in depth can calculated in the manner described byAubeny & Chi (2010). When expressed in terms of zc rather than Zc , the rate equation takes the following form:

where zˆc is normalized depth of shackle (= zc /b), Tˆ a is normalized tension at shackle (= Ta /sua b2 ), sua is soil shear strength at the shackle, dθ0c /dθ ac is the rate of change of mudline angle, ηc is a strength gradient parameter (= bkcosδv /su0 ). Therefore, the rate equation for computing the change in θ ac with increasing embedment in the three-dimensional case assumes a nearly identical form to that of an anchor contained within a vertical plane. 3

are computed using the procedure of Aubeny and Chi (2010) developed for vertical in-plane motion. Anchor line angles at the end of this installation stage are θ agi and θ0gi . 2. At this point, a partial mooring failure is simulated in which the anchor line angle at the mudline deviates from the intended plane of loading (the xg direction in Figure 1) by an angle φ0g . The anchor line angle at the pad-eye is assumed to be initially unaffected by this change, such that φag = 0 and θ ag = θ agi . From Equations 1–7, the anchor line angles, θ aci and θ 0ci measured in the c-plane, can be computed. Inserting these values into Equation 8 permits the current line tension T a to be calculated. The calculations will show that the current pad-eye tension is less than the original installation tension, Ta < Tai . Therefore, as continued loading occurs, the anchor will remain stationary and the angle θ ac decreases until Ta = Tai . During this “line tautening” process, the orientation of the c-plane changes continuously as θ ac changes. 3. When the line tension reaches the point at which Ta = Tai , further loading will be accompanied by a resumption of the drag embedment process, now in the c-plane. Experience with simulations of twodimensional trajectories (Aubeny & Chi, 2010) implied that a drag anchor rapidly aligns itself with the anchor line such that the line of action of Ta acts through the center of the anchor. A similar process is assumed to occur under three-dimensional loading conditions. That is, upon resumption of the drag embedment process, the anchor is assumed to rapidly re-align itself such that the shank lies in the c-plane. 4. Considering the orientation of the anchor in the cplane (Figure 3), the orientation of the fluke will therefore be θ fc = θ fs -θ ac . If the fluke advances an increment of distance s in the direction of the fluke, then the components of displacement in the c-plane become:

The coordinates of the shackle in the global frame are accomplished through the use of the inverse of matrix [Rgc ]−1 :

COMPUTATION SEQUENCE

Having formulated the necessary transformations to describe the anchor chain in an oblique plane, trajectory prediction can proceed according to the sequence outlined below. 1. Initial installation to an arbitrarily selected depth zi and the tension Tai corresponding to this depth © 2011 by Taylor & Francis Group, LLC

701

As occurs with drag embedment in a vertical plane (e.g. O’Neill et al. 1999), the anchor reaches an ultimate embedment depth when the centerline of the fluke is parallel to the ground surface and shackle tension converges to a maximum value.

Figure 6. Predicted effect of out-of-plane loading on anchor load capacity.

Figure 5. Predicted effect of out-of-plane loading on anchor trajectory.

4

EXAMPLE ANALYSIS

For the process described above, two variables with respect to out-of-plane loading will affect the trajectory and capacity of the anchor following partial failure of the mooring system, the first being the magnitude of the out-of-plane orientation angle of the chain at the mudline. The second is the installation depth zi which failure occurs. For parametric studies, an installation penetration ratio Rp may be defined as follows:

where zmax is the ultimate embedment depth under conditions of purely in-plane loading. As an illustrative parametric study of drag anchor trajectory predictions for out-of-plane loading, we consider the case of drag embedment of an anchor with 12-m2 anchor in a soft clay having zero strength at the mudline and a strength gradient k = 1.57 kPa/m. The anchor line in this example is a wire line with diameter b = 0.089 m. Pad-eye tension was related to soil strength at the pad-eye sua and fluke area Af by the relationship Ta = Ne sua Af , where is Ne is the anchor bearing factor.A bearing factor Ne = 5 was selected for this example. Trajectory predictions were performed for φ0g = 30 degrees, and Rp was set to 0.25, 0.5, and 0.75. The anchor line angle at the mudline was maintained in a horizontal orientation at all stages of the simulation, φ0g = 0. Figure 5 shows the projections of the predicted trajectories onto the plane of intended loading (xg –zg plane) for the various cases. In all cases in which Rp is less than unity, continued embedment occurs following the occurrence of out-of-plane loading. However, the ultimate embedment depth, and therefore anchor load capacity, is considerably less than what would develop had the anchor continued its trajectory in a vertical plane. Figure 6 shows the anchor load capacities corresponding to the trajectory predictions in Figure 5. Consistent with the trajectory predictions, after partial failure of the mooring system, there is a modest

© 2011 by Taylor & Francis Group, LLC

Figure 7. Predicted trajectory out of plane of intended loading.

gain in anchor load capacity under sustained dragging. However, the ultimate capacity is much less than that which would develop under continued dragging within the original vertical plane of intended loading. Figure 7 shows the predicted trajectory in a plan view; i.e. a projection of the anchor trajectory onto the xg –yg plane. The out-of-plane path exhibits a slight curvature, with greater curvature corresponding to lower Rp values. The scale is exaggerated in the yg direction, but a close inspection of the plot shows that the average angle of the deflected anchor path is roughly comparable to the out-of-plane anchor line angle φ0g imposed at the mudline. It is noted that Step 2 of the analysis procedure makes a fairly significant assumption in maintaining φag = 0 during the process which defines the re-configuration of the chain following a partial mooring failure. If, for example, φag were actually greater than zero, the reduction in ultimate embedment depth and load capacity would be somewhat less. In the extreme (unlikely) case of φag equaling φ0g , the anchor would simply resume its trajectory in a vertical plane oriented an angle φag from the xg -axis, with no loss of ultimate embedment or capacity. Thus, from this perspective, the assumed behavior in Step 2 provides a conservative prediction of the effects of out-of-plane

702

loading due to a partial mooring failure. A planned program of laboratory model tests are expected to shed light on the validity of the assumptions made in formulating the analysis procedure, particularly those in Step 2. 5

the embedment process within an oblique plane. For the example analysis considered in this paper, continued embedment occurred after imposition of outof-plane loading; however, the additional embedment beyond the initial installation depth was relatively modest. While the analyses did not indicate a loss of the original installation load capacity due to a mooring system failure – at least for the scenario considered by the parametric study – they did predict a substantial reduction in the ultimate, or reserve, load capacity of the anchor.

CONCLUSIONS

This paper presents an analysis of an anchor which, after an initial installation stage occurring in the vertical plane of intended loading, is subjected to an out-of-plane load. Except at very shallow depths, the soil shearing resistance is large relative to the weight of the anchor chain, in which case the chain will lie in a plane, even under conditions of out-of-plane loading. The orientation of the chain at the pad-eye and mudline determine the orientation of the oblique plane. With appropriate rotation of the coordinate system, it is possible to formulate tension-angle and trajectory prediction equations that are very similar to those previously developed for anchor chains contained within a vertical plane. Using the modified equations, a calculation sequence has been developed for predicting drag anchor performance for the following scenario: (1) initial anchor installation within the intended vertical plane of loading to an arbitrary installation depth, (2) the occurrence of out-of-plane loading applied at the mudline due to partial failure of the mooring system accompanied by the re-configuration of the anchor chain, and (3) continued loading with resumption of

© 2011 by Taylor & Francis Group, LLC

REFERENCES Aubeny, C. P., Murff, J. D., and Kim, B. M. (2008), “Prediction of Anchor Trajectory during Drag Embedment in Soft Clay”, International Journal of and Offshore Polar Engineering, 18(4), 314–319. Aubeny, C. P., and Chi, C. (2010), “Mechanics of Drag Embedment Anchor in a Soft Seabed”, Journal of Geotechnical and Geoenvironmental Engineering, ASCE. Neubecker, S. R., and Randolph, M. F. (1995). “Profile and frictional capacity of embedded anchor chain,” Journal of Geotechnical Engineering Division, ASCE, 121, 11, 787–803. O’Neill, M.P., Randolph, M.F., and House, A. R. (1999) “The Behavior of DragAnchors in Layered Soils,” International Journal of Offshore and Polar Engineering, Vol. 9, No. 1, March 1999. Yang, M., Murff, J.D. and Aubeny, C.P. (2010) “Undrained Capacity of Plate Anchors under General Loading,” accepted for publication Journal of Geotechnical and Geoenvironmental Engineering, ASCE.

703

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Setup following keying of plate anchors assessed through centrifuge tests in kaolin clay A.P. Blake & C.D. O’Loughlin Department of Civil Engineering & Construction, Institute of Technology Sligo, Sligo, Ireland

C. Gaudin Centre for Offshore Foundation Systems, University of Western Australia, Perth, Australia

ABSTRACT: This paper describes a series of centrifuge tests carried out at 100 g on square plate anchors to assess the potential capacity increase due to reconsolidation following the keying process. Numerous post-keying consolidation periods were adopted before reloading the anchor and measuring the subsequent peak capacity. The tests with long term reconsolidation periods gave an average bearing capacity factor, Nc = 13.5, which is in good agreement with experimental and numerical bearing capacity factors in the literature. The tests with the shortest reconsolidation periods yielded Nc values that were ∼75% of the long term values. The rate at which reconsolidation takes place, as assessed through the increase in Nc with time, is much slower for plate anchors than for piles, suction caissons and dynamically installed (torpedo) anchors. It is suggested that the maximum practical reconsolidation is 50% for which Nc = 11.8.

1

INTRODUCTION

The continuous depletion of oil and gas deposits in shallow waters has made it necessary to exploit reserves in deep waters (500–1500 m) and ultra deep waters (>1500 m) in order to ensure future supply. In deep waters it is not feasible to use fixed drilling and production platforms and instead floating structures are employed. Associated with the move to deeper waters, has been a transition from catenary to taut and semi-taut leg mooring systems (Eltaher et al. 2003). Taut and semi-taut mooring lines are designed to have high angles of inclination (40–45◦ ) with the mudline and thus require anchors that can sustain significant vertical components of load (Ehlers et al. 2004). One such anchoring system is the ‘Suction Embedded PlateAnchor’(SEPLA) (see Figure 1).The SEPLA uses a suction caisson to embed a plate anchor that is slotted vertically into its base. The caisson is lowered to the seabed, where it is allowed to penetrate under self-weight. Water is then pumped from the interior of the caisson to allow the system to reach the design embedment depth. The plate anchor mooring line is then disengaged from the caisson and the pump flow is reversed, water being forced back into the caisson, causing it to move upwards while leaving the plate anchor in place at the required depth. At this stage the plate anchor and the mooring line are embedded vertically in the seabed. The mooring line attached to the plate anchor is then tensioned, causing the plate anchor to rotate or ‘key’ to an orientation that is perpendicular to the direction of loading. In this way, the maximum

© 2011 by Taylor & Francis Group, LLC

projected area is presented to the direction of loading, maximising the bearing resistance of the anchor. The installation and keying processes are summarised schematically in Figure 2. The anchor keying process gives rise to two effects. Firstly, the anchor embedment depth will reduce as the plate rotates, and secondly, the clay in the immediate vicinity of the plate will be remoulded (Randolph et al. 2005). The reduced anchor embedment depth has been addressed both experimentally and numerically (e.g. O’Loughlin et al. 2006, Gaudin et al. 2009, Song et al. 2009, Yu et al. 2009). These studies suggest that the loss in embedment is mainly a function of the eccentricity of the padeye and that the keying induced loss in embedment may be minimised by ensuring that the anchor padeye is located at a minimum distance

705

Figure 1. SEPLA plate and caisson.

Figure 2. SEPLA concept: 1. Suction installation, 2. Caisson retrieval, 3. Anchor keying, 4. Mobilised anchor. Figure 3. 1:100 reduced scale model square plate anchor.

from the anchor plate (e.g. at least a full anchor height as suggested by O’Loughlin et al. 2006). It is likely that the reduction in soil strength due to remoulding during keying will be recovered in due course through consolidation. However the magnitude of the associated short term reduction in anchor capacity and the rate at which this capacity is regained has not yet been addressed. To this end this paper investigates ‘setup’ effects for directly embedded plate anchors through a series of centrifuge tests on square plate anchors in normally consolidated kaolin clay. 2 2.1

EXPERIMENTAL PROGRAM Geotechnical centrifuge

Centrifuge tests were carried out at an acceleration level of 100 g using the geotechnical drum centrifuge at the University of Western Australia. The drum channel has an internal diameter of 1200 mm, allowing for a soil sample that is 300 mm wide (vertical in-flight) and 200 mm deep (radial in-flight). The maximum operational speed of the channel is 850 rpm which equates to a maximum acceleration level of 485 g at the base of the channel. The tool table is capable of vertical, radial and circumferential actuation. The drum centrifuge facility is described in detail by Stewart et al. (1998). 2.2

Model anchor

The square plate anchor was modelled at a reduced scale of 1:100 and fabricated from stainless steel (Figure 3). The plate breadth (B) and length (L) was 25 mm (2.5 m in prototype scale) with a thickness of 1 mm (0.1 m in prototype scale). The padeye eccentricity (e) was 15 mm (1.5 m in prototype scale), so that the eccentricity ratio, e/B = 0.6. The anchor mooring line was modelled using uncoated stainless steel fishing wire (0.9 mm diameter) due to its flexibility, high tensile strength and resistance to stretching. © 2011 by Taylor & Francis Group, LLC

Table 1.

Kaolin clay properties (after Stewart 1992).

Property

Symbol

Units

Value

Liquid limit Plastic limit Specific gravity Angle of internal friction Void ratio (at p
= 1 kPa on critical state line) Slope of normal consolidation line Slope of swelling line Strength ratio Coefficient of vertical consolidation (at OCR = 1 and σv
= 75 kPa)

LL PL Gs φ ecs

% %

61 27 2.6 23 2.14

2.3

◦

0.205

λ κ su /σv
cv

2

m /yr

0.044 0.19 2.6

Sample preparation and soil properties

Samples were prepared by mixing commercially available kaolin clay powder with water to form a slurry at 120% water content (twice the liquid limit, see Table 1). The slurry was deposited within the drum channel at 20 g using a gantry crane, hopper and hose line. During consolidation additional slurry was added to achieve the target sample depth of 155 mm and a 5–10 mm layer of water was maintained at the sample surface to ensure saturation. A drainage blanket at the base and along the sides of the sample allowed for 2 way drainage during consolidation. Table 1 provides a summary of the characteristics of kaolin clay. Soil characterisation tests were conducted using a T-bar penetrometer (Stewart & Randolph 1991) throughout the testing program to quantify any variation in undrained shear strength over the course of the tests. The T-bar penetrometer was inserted and extracted at a rate of 1 mm/s, which corresponds with a normalised velocity, V = vd/cv > 30 (where v is the T-bar penetration velocity, d is the T-bar diameter and cv the coefficient of vertical consolidation), thus ensuring undrained conditions (Lehane et al. 2009). Typical

706

Figure 5. Anchor installation arrangement.

Figure 4. Typical undrained shear strength profiles.

undrained shear strength profiles are provided in Figure 4, where the undrained shear strength is derived from the penetration resistance using the commonly adopted T-bar factor, Nt−bar = 10.5 as suggested by Stewart and Randolph (1991). The dashed lines on Figure 4 represent theoretical su profiles that account for the varying acceleration level in the centrifuge sample (93 to 118 g) and the variation in the effective unit weight of the sample (determined from a core sample after the centrifuge testing). At T-bar penetrations greater than ∼6 T-bar diameters, the measured profiles correspond with undrained strength ratios in the range su /σv
= 0.14 to 0.17.The latter value, su /σv
= 0.17, is representative of conditions close to the end of testing, and implies a prototype undrained shear strength gradient, k = 1.1 kPa/m (assuming an average effective unit weight of 6.5 kN/m3 ) which is typical of that reported for kaolin clay (e.g. Gaudin et al. 2006).

2.4

Figure 6. Anchor keying and pullout arrangement.

Load was measured using load cells at the actuator and anchor padeye, although the padeye load cell proved unreliable over the testing program and consequently its test data has not been considered in the analysis. The load-displacement response of the actuator load cell was monitored in real time until keying was considered to be complete (discussed later).At this point loading stopped and a post-keying consolidation period was permitted before subsequently loading the anchor to failure. Tests with no post-keying consolidation period were also undertaken to quantify the capacity of the plate anchor when no reconsolidation is permitted.

Experimental setup and procedure

To simplify the testing procedure the plate anchor was installed in the sample at 1 g using an installation ‘fork’ driven by an actuator (see Figure 5). The mid anchor breadth embedment depth, H = 112.5 mm, corresponds to a normalised embedment ratio, H/B = 4.5, which is sufficient to ensure a deep failure mechanism during pull-out (Merifield et al. 2003). Following installation, the installation fork was removed and the mooring cable was connected to the actuator. The drum tool table was lowered by 15 mm to position the actuator centrally above the anchor padeye such that the load inclination was vertical (see Figure 6). Prior to anchor keying, a reconsolidation period of 30 minutes (at 100 g) was permitted to allow for regain of potential strength loss of the sample due to swelling at 1 g. Following this reconsolidation period the anchor was loaded via the mooring cable at an actuator velocity, v = 0.15 mm/s, which corresponds with a normalised velocity, V = vB/cv in excess of 30, thus ensuring undrained conditions (House et al. 2001). © 2011 by Taylor & Francis Group, LLC

3

EXPERIMENTAL RESULTS AND DISCUSSION

3.1 Test programme summary The testing took place over a period of two weeks and included 23 individual anchor tests having reconsolidation periods in the range t = 0 to ∼66 hours. The test database is summarised in Table 2.

707

Table 2.

Summary of plate anchor test results.

Consolidation Peak Bearing time, t Dimensionless capacity, capacity Fv (N) factor, Nc Test (hh:mm:ss) time, T∗ 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 ∗

00:00:00 00:25:55 00:52:34 13:08:24 00:01:01 00:04:19 00:12:58 01:45:07 13:08:24 00:01:01 00:25:55 00:00:00 04:22:48 00:12:58 00:25:55 04:22:48 00:52:34 01:45:07 65:42:00 00:02:01 00:08:38 00:52:34 08:13:54

0.00 3.10 6.28 94.5 0.12 0.52 1.56 12.57 94.45 0.12 3.13 0.00 31.84 1.56 3.15 31.57 6.31 12.61 474.05 0.24 1.01 6.11 57.44

91.64 98.91 99.28 111.36 89.18 91.59 96.71 102.20 111.39 92.16 95.88 94.78 117.11 99.59 104.40 109.32 109.63 108.22 118.31 93.94 95.04 101.21 106.96

10.95 11.96 12.01 13.44 10.47 10.65 11.41 12.28 13.42 10.40 10.68 10.91 12.94 11.33 11.46 12.63 12.67 12.52 13.67 10.38 10.49 11.78 12.49

Figure 8. Dependence of bearing capacity factor on post keying reconsolidation.

observed at approximately 20 mm displacement in both tests is typical for a vertical load inclination and represents the majority of the anchor rotation during keying. In tests where no post-keying reconsolidation was permitted (e.g. Test 1), the actuator movement was continued until a clear peak capacity was observed. In tests where a period of post-keying reconsolidation was permitted (e.g. Test 19), the actuator was stopped and the load allowed to relax after the ‘keying plateau’ was observed. At the end of the reconsolidation period the actuator was restarted at the same velocity until a distinct peak capacity was observed.

See Equation 2

3.3 Anchor capacity As the load inclination and hence the mooring line in the clay is vertical, the final anchor embedment depth after keying may be readily determined from Figure 7. The undrained shear strength can be obtained using this final anchor embedment depth and the most representative T-bar profile. In this way experimental bearing capacity factors (Nc ) can be calculated using:

Figure 7. Typical load-displacement responses during keying and pull-out.

3.2

Load response during keying and pullout

Figure 7 represents the load-displacement response for Test 1 having no post-keying consolidation time (t = 0), and Test 19 with the highest post-keying consolidation time (t = 2.7 days). The different stages of the load response during keying and pull-out of a plate anchor are explained in detail by Gaudin et al. (2006). After recovering the slack in the mooring line, the load starts to increase as the anchor starts to key. The plateau © 2011 by Taylor & Francis Group, LLC

708

where Fv is the peak anchor capacity less the effective weight of the anchor in clay, A is the projected area of the anchor (L × B) and su is the most representative undrained shear strength of the soil at the anchor’s post-keying embedment depth. The experimental bearing capacity factors are plotted in Figure 8 against the non-dimensional time (T) expressed as:

where t is the post-keying reconsolidation time, ch is the horizontal coefficient of consolidation, and d

is the drainage path length assumed to be B/2. The horizontal coefficient of consolidation was estimated using Equation 3 (Lehane et al. 2009):

where kh0 is the horizontal coefficient of permeability (assumed to be kh0 = 1.49e2.03 × 10−3 as established by Al-Tabbaa & Wood (1987) for kaolin clay at 20◦ C), e0 is the void ratio at the final anchor embedment depth (determined from post centrifuge test sampling to be e0 = 1.37), σh
is the horizontal effective stress and γw is the unit weight of water. Figure 8 shows the Nc values (also summarised on Table 2) to be in the range 10.4 to 13.7, with higher values corresponding to the longest reconsolidation periods. These values are in good agreement with Nc = 12.3 to 13.5 as reported by Gaudin et al. (2006) for ‘jacked-in’ square plate anchors in normally consolidated kaolin clay. The upper bound of the experimental range (this study and Gaudin et al. 2006) is in excellent agreement with Nc = 13.5 as determined numerically for a deeply embedded square plate anchor (Song et al. 2006). Normalised capacity ratios, Nc /Nc,max , are plotted against non-dimensional time, T, in Figure 9. Nc,max = 13.7 at T = 474 was taken from Test 19 which had the longest post-keying reconsolidation period (∼66 hours = 75 years in prototype scale) and represents the long term ‘maximum’ anchor capacity. Also shown on Figure 9 is the following curve fitting function which was used by Richardson et al. (2009) in analysing setup effects for dynamically installed anchors:

where A1 is the initial normalised capacity ratio; A2 is the final normalised capacity ratio; T0 is the value of T at the mid-point between A1 and A2 and p is a fitting parameter governing the slope of the curve. The best fit parameters for the normalised test data are A1 = 0.75, A2 = 1.0, T0 = 8.0 and p = 0.75. The initial short term capacity at T ≤ 0.01 is considered to represent short term capacity as it corresponds with a prototype reconsolidation period of ∼14 hours (assuming a prototype plate dimension, B = 2.5 m and ch = 10 m2 /yr). Nc /Nc,max ∼ 0.75 at T = 0.01 infers that the short term capacity is ∼75% of the long term capacity. This is considerably higher than corresponding values for piles and suction caissons, where the short term capacity is typically 25 to 45% of the long term maximum capacity (Esrig et al. 1977, Bogard and Matlock 1990, Chen and Randolph, 2007). In contrast the short term capacity of dynamically installed (torpedo) anchors has been demonstrated to be as low as 6% of the long term capacity (Richardson et al. 2009). Unlike the quoted capacity reduction ratios for piles, suction caissons and dynamically installed anchors, the tests reported here do not consider installation effects. However the agreement with work reported by Gaudin et al. (2006) which modelled (jacked and suction) installation and keying, suggests that the dominant reduction in strength is due to keying, not installation.

3.4

Practical implications

The time required for 90% consolidation following keying of a square plate anchor with a prototype dimension, B = 2.5 m, and ch = 10 m2 /yr is t90 = 15 years. This value is much greater than t90 values established from centrifuge testing for dynamically installed anchors (t90 = 35–350 days) (Richardson et al. 2009), and for suction caissons (t90 = 90 days or less) reported by Jeanjean (2006). A more realistic degree of reconsolidation for plate anchors might be 50%, which for B = 2.5 m, and ch = 10 m2 /yr, requires a reconsolidation period, t50 ∼ 1 year at which point Nc = 11.8. 4

CONCLUSIONS

This paper presents results from a series of drum centrifuge tests in normally consolidated clay that were devised to quantify the magnitude of the short term reduction in anchor capacity due to anchor keying, and the rate at which this capacity is regained through consolidation. The main findings from this study are summarised in the following:

Figure 9. Dependence of normalised capacity ratio on post keying consolidation.

© 2011 by Taylor & Francis Group, LLC

1. Bearing capacity factors for deeply embedded square plate anchors in clay are in the range 10.4 to 13.7. The lower bound of this range indicates the reduction in short term anchor capacity due to soil remoulding during anchor keying (∼25%),

709

whilst the upper bound is in excellent agreement with other published bearing capacity factors determined both experimentally and numerically. 2. The ratio of short to long term capacity for plate anchors is much higher than those for piles, suction caissons and dynamically installed (torpedo) anchors. 3. The time for 90% reconsolidation is very long for plate anchors and not practical. For the conditions considered here, a consolidation period of 1 year would lead to reconsolidation of ∼50%, for which the bearing capacity factor is 11.8. ACKNOWLEDGEMENTS The technical support of Mr. Bart Thompson, Mr. Phil Hortin and Mr. Dave Jones of the Centre for Offshore Research Foundation Systems is gratefully acknowledged. REFERENCES Al-Tabbaa, A. & Wood, D.M. 1987. Some measurements of the permeability of kaolin. Géotechnique 37(4): 499–503. Bogard, D. & Matlock, H. 1990. Application of model pile tests to axial pile design. Proc. 22nd Offshore Technology Conf., Houston, Texas, USA, May 7–10 1990. OTC 6376. Chen, W. & Randolph, M.F. 2007. External radial stress changes and axial capacity for suction caissons in soft clay. Géotechnique 57(6): 499–511. Ehlers, C.J., Young, A.G. & Chen, J.H. 2004. Technology assessment of deepwater anchors. Proc. 36th Offshore Technology Conf., Houston, Texas, USA, 3–6 May 2004. OTC 16840. Eltaher, A., Rajapaksa, Y. Chang, K.T. 2003. Industry trends for design of anchoring systems for deepwater offshore structures. Proc. 35th Offshore Technology Conf., Houston, Texas, USA, 5–8 May 2003. OTC 15265. Esrig, M.I., Kirby, R.C., Bea, R.G. & Murphy, B.S. 1977. Initial development of a general effective stress method for the prediction of axial capacity for driven piles. Proc. 9th Offshore Technology Conf., Houston, Texas, USA, 2–5 May 1977. OTC 2943. Gaudin, C., O’Loughlin, C.D., Randolph, M.F. & Lowmass, A.C. 2006. Influence of the installation process on the performance of suction embedded plate anchors. Géotechnique 56(6): 381–391. Gaudin, C., Tham, K.H. & Quashine, S. 2009. Keying of plate anchors in NC clay under inclined loading. Int. J. Offshore and Polar Engrg. 19(2): 135–142.

© 2011 by Taylor & Francis Group, LLC

House, A.R., Oliveria, J.R.M.S. & Randolph, M.F. 2001. Evaluating the coefficient of consolidation using penetration tests. Int. J. Physical Modelling in Geotechnics 1(3): 17–26. Jeanjean, P. 2006. Setup characteristics of suction anchors for soft Gulf of Mexico clays: experience from field installation and retrieval. Proc. 38th offshore Technology Conf., Houston, Texas, USA, 1–4 May 2006. OTC 18005. Lehane, B.M., O’Loughlin, C.D., Gaudin, C. & Randolph, M.F. 2009. Rate effects on penetrometer resistance in kaolin. Géotechnique 59(1): 41–52. Merifield, R.S., Lyamin, A.V., Sloan, S.W. & Yu, H.S. 2003. Three-dimensional lower bound solutions for stability of plate anchors in clay. J. Geotech. and Geoenvir. Engrg. 129(3): 243–253. O’Loughlin, C.D., Lowmass, A.C., Gaudin, C. & Randolph, M.F. 2006. Physical modeling to assess keying characteristics of plate anchors. Proc. 6th Int. Conf. in Physical Modelling in Geotechnics, Hong Kong, 4–6 August 2006, 1: 659–666. Randolph, M.F., Cassidy, M.J., Gourvenec, S. & Erbrich, C.T. 2005. Challenges of offshore geotechnical engineering. Proc. 16th Int. Conf. on Soil Mech. and Geotech, Engrg., Osaka, Japan, 12–16 September 2005, 1: 123–176. Richardson, M.D., O’Loughlin, C.D., Randolph, M.F. & Gaudin, C. 2009. Setup following installation of dynamic anchors in normally consolidated clay. J. Geotech. and Geoenvir. Engrg. 135(4): 487–496. Song, Z., Hu, Y., Wang, D. & O’Loughlin, C.D. 2006. Pullout capacity and rotational behaviour of square anchors in kaolin clay and transparent soil. Proc. 6th Int. Conf. on Physical Modelling in Geotechnics, Hong Kong, 4–6 August 2006, 2: 1325–1331. Song, Z., Hu, Y., O’Loughlin, C.D. & Randolph, M.F. 2009. Loss in anchor embedment during plate anchor keying in clay. J. Geotech. & Geoenvir. Engrg. 135(10): 1475–1485. Stewart, D.P. & Randolph, M.F. 1991, A new site investigation tool for the centrifuge. Proc. Centrifuge 91, Boulder, Colorado, 13–14 June 1991, 1:531–538. Stewart, D.P, 1992. Lateral loading of piled bridge abutments due to embankment construction. PhD thesis, The University of Western Australia. Stewart, D.P, Boyle, R.S. & Randolph, M.F. 1998. Experience with a new drum centrifuge. Proc. Centrifuge 98, Tokyo, Japan, 23–25 September 1998. 1: 35–40. Yu, L., Liu, J., Kong, X. & Hu, Y. 2009. Three-dimensional numerical analysis of the keying of the vertically installed plate anchors in clay. Computers and Geotechnics 36(4): 558–567.

710

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Seismically-induced displacements of a suction caisson in soft clay A.J. Brennan University of Dundee, UK

S.P.G. Madabhushi University of Cambridge, UK

P. Cooper Intecsea (UK) Ltd.

ABSTRACT: The seismic design for offshore foundations is based predominantly on experience onshore. This paper describes the results of dynamic centrifuge tests performed to validate the performance of a suction caisson installed in normally consolidated clay. The main objective is to evaluate the likely plastic displacement under different shaking levels. Permanent displacement results indicate that the displacements experienced are well within the allowable movement for the foundation considered, even though a strength based design approach would consider this to be a failure. Larger earthquakes are seen to produce comparatively smaller displacements. It is concluded that the when designing for seismic loading, if some displacement is permissible then a performance-based approach allowing some displacement proves significantly less conservative than a purely strength-based design. It is also concluded that dynamic response analyses should consider the strength of soil, as this can act as a fuse against large amplitude shear waves. 1

INTRODUCTION

Whilst much work is ongoing into the static and small strain cyclic capacity of offshore foundations, there is little specific research into the effects of seismic loading on this infrastructure. There is consequently a limited amount of appropriate guidance for seismic analysis of offshore structures, as conventional earthquake engineering considers primarily onshore infrastructure and incorporates a significant degree of conservatism. Although such conservatism is often necessary due to the unpredictable nature of earthquake loads, it is a major limitation on offshore design. One area where conservatism can be reduced is in the design philosophy. Peak forces from earthquakes may be large but their limited duration may make the ensuing displacement manageable. It therefore become important to evaluate the displacements resulting from a design earthquake as well as the potential forces. Structural earthquake engineering is moving away from conventional force-based designs to displacementbased (or performance-based) designs (e.g. Priestley et al., 2007), but geotechnical earthquake engineering has been slow to follow, with only simple “Newmark sliding blocks” (Newmark, 1965, and subsequent upgrades e.g. Torisu et al., 2010) making a significant contribution. This research was undertaken to assess the seismic behaviour of a manifold foundation on a normally consolidated clay. The preferred foundation for this

© 2011 by Taylor & Francis Group, LLC

manifold was a suction caisson. Both foundation type and soil are rare in conventional earthquake engineering, and a force-based seismic design resulted in a very long caisson. 1.1 Aims and objectives The aim of the research was primarily to evaluate the seismic displacements that might accrue if the caisson were rather shorter, using dynamic centrifuge modelling. Results presented in the current paper will examine: 1. Permanent displacements obtained under different levels of seismic loading. 2. Accelerations responsible for these displacements. 3. Any role played by excess pore pressures. 4. A discussion on the influences on caisson response. Complementary results of static response and dynamic transfer functions are presented in a sister paper (Brennan et al., 2006a). 2 2.1

METHOD Dynamic centrifuge modelling

Accurate reproduction of in-situ self weight stresses in a small-scale model may be achieved by increasing gravity. The Cambridge University 10 m diameter beam centrifuge was used for this purpose. In this

711

Parameter

Value

into the model container (see below), trying to keep air-entrapment to a minimum. The properties of this soil have been well-documented, and the following parameter values in Table 2 are obtained from Barker (1998). In-flight consolidation was not a viable option as the Cambridge centrifuge cannot run unattended, so another solution for rapid consolidation had to be found. At the target prototype condition of 20 m deep clay at buoyant unit weight 5 kN/m3 , the base effective stress is 100 kPa. With a sheet of porous plastic across the entire base, a vacuum was applied to the ports that are normally used to saturate models through the base. The vacuum pump was connected via a chamber for collecting displaced water. The model surface was free to atmosphere and kept submerged to prevent air entrapment. Thus, the target effective stress distribution was applied to the model on the laboratory floor, and this was reapplied whenever the model was not under centrifugally applied high gravity. A surcharge of 4.6 kPa was then applied to the surface for 24 hours, in order to provide a surface undrained shear strength of approximately 1 kPa in line with field conditions. The structure was installed under self weight as far as possible. A hole enabled air trapped within the caisson to escape through the floor of the superstructure. When self-weight installation had achieved 140 mm penetration, model scale, additional suction was required to continue installation. Once the superstructure was in contact with the soil, the drainage hole was plugged to provide a realistic suction condition. Instrumentation was retrospectively added to the model, by auguring holes, installing the instruments and backfilling with clay slurry.

Specific gravity Gs Permeability k Coefficient of consolidation Cv normal compression isotropic rebound Slope of virgin compression line λ Slope of unload-reload κ line Plastic limit Liquid limit

2.60 5 × 10−8

2.3

Figure 1. Dimensions (in prototype scale) and salient instrument locations. Table 1. Salient positions in Figure 1, relative to lower left corner. All dimensions in metres, prototype scale. Point A2 A6 A9 P1 H1 V1 V2 CoG

accelerometer accelerometer accelerometer pore pressure transducer displacement transducer displacement transducer displacement transducer structure centre of gravity

Table 2.

x

z

4.0 17.5 18.4 26.5 18.4 20.7 23.0 21.6

0.0 12.3 20.8 9.7 20.4 21.8 21.8 14.3

Parameters for E-grade kaolin (after Barker, 1998).

1.0 mm2 /s 5.0 mm2 /s 0.124 0.02 30% 51%

case, caisson and superstructure were modelled at a length scale of 1:50, requiring an increase in gravity to 50 g. The full scale prototype modelled is a manifold superstructure of mass 306 tons supported by a 6.3 m outer diameter 10 m long caisson in a 17 m thick layer of normally consolidated clay (soil data in Table 2). A schematic diagram showing the dimensions and instrumentation for the caisson is shown in Figure 1. Instrument positions and structure centre of gravity are given in Table 1, relative to an origin in the container lower left corner. 2.2

Model construction

The model is constructed from E-grade kaolin clay, mixed at 100% water content and deposited by hand

© 2011 by Taylor & Francis Group, LLC

Model containment and loading system

Boundary conditions for dynamic physical modelling present a difficult problem, particularly in soft (and highly non-linear) soils. The tests presented were tested in an equivalent shear beam container designed by Brennan et al., 2002) consisting of laminations of stiff metal and flexible rubber as shown in Figure 1. This container is designed to match the dynamic amplification of soil layers of a given stiffness; however, the design soil was a medium dense sand that is likely to be somewhat stiffer than the soil used in the tests presented. Accelerometers, particularly near the container boundary, consequently recorded loading frequencies that appeared to have been amplified by the container rather than the soil (Brennan et al., 2006a). This must therefore be considered when analysing the results, and none of conclusions drawn will have been significantly affected by this influence. It would be prudent to recommend that future tests performed on soft soil, where timescale allows, should use a more flexible container such as a free-moving laminar stack (Brennan et al., 2006b). The earthquake motion was applied by means of a Stored Angular Momentum actuator designed by Madabhushi et al. (1998). This produces

712

Table 3.

Shaking events considered.

Earthquake

Peak accel. (g)

Duration (s)

Freq. (Hz)

Small Moderate Large

0.030 0.088 0.175

25 25 15

0.6 1 1

approximately sinusoidal loading at frequency f0 , with significant harmonics at 3f0 and 5f0 as well as some other harmonics. For this work, such vagaries were accommodated and even taken advantage of (for dynamic transfer functions, see Brennan et al., 2006a) but it is worth noting that recent developments in centrifuge-mounted earthquake actuation have resulted in a number of actuators worldwide that are capable of superior sinusoidal loading as well as broadband frequency motions if required. 2.4 Dynamic and added mass Structure inertia in water is greater due to the added mass effect. The superstructure was constructed with this added mass included in its static weight. To ensure that this did not add excessive vertical loads, a 2 kg counterweight was designed to reduce the weight force of the heavy structure. Thus, both the vertical weight force of the structure (56 tons) and apparent mass in horizontal inertia (306 tons) were modelled. 3

Figure 2. Small earthquake: a) horizontal displacement H1; b) vertical displacement V1; c) vertical displacement V2; d) input acceleration A2.

DYNAMIC RESPONSE

In this section, the effect of three shaking events is considered, as outlined in Table 3, in prototype scale. In practice, these were earthquakes number two, one and five from a series of six carried out on the same model, the effect of which is considered below. 3.1 Permanent displacements Figures 2–4 show the permanent prototype displacements measured by the one horizontal and two vertical displacement transducers shown in Figure 1. Measurements are sampled at 40 Hz prototype scale and filtered at 5 Hz to reduce noise. The only displacements seen during the small earthquake are a small horizontal movement, indicating a purely translational oscillatory response (no significant rocking) with a peak amplitude of 20 mm each The peak displacement of the input motion will be approximately 21 mm (obtained by |d| ≈ |a|/(2πf)2 ) implying that the displacement is predominantly elastic, and this is verified by the absence of permanent displacement seen at the end of the test. Figure 3, presenting the displacements measured under a moderate earthquake (peak acceleration 0.088 g), shows a different response. The oscillatory horizontal displacement is superimposed on a drift of 62 mm permanent displacement.This is well within the

© 2011 by Taylor & Francis Group, LLC

713

Figure 3. Moderate earthquake: a) H1; b) V1; c) V2; d) A2.

Figure 5. Small earthquake: a) structure acceleration A9; b) soil acceleration A6; c) input acceleration A2.

Figure 4. Large earthquake: a) H1; b) V1; c) V2; d) A2.

1 m permanent displacement that this system can tolerate, proving the value of allowing some displacement to occur. This movement is also largely translational, as there is very little vertical oscillation measured. A small amount of tilting occurs as shown by the slight permanent vertical displacement. This represents a tilt of 0.30◦ relative to the initial position. For the larger earthquake shown in Figure 4, a much smaller permanent displacement is seen, and the oscillatory component of displacement appears to reduce in amplitude through the duration of shaking. Displacement at input should peak at 43 mm, but most caisson displacement is somewhat less than this amount. This is examined more below. A further point to observe here is the vertical response. Instrument v2 appears to record no movement whereas v1 shows some oscillation at amplitudes of almost 10 mm. This implies that noticeable rocking is occurring in this case, about a point on the axis of instrument v2. 3.2 Acceleration response Horizontal accelerations recorded at positions A2 (base of model), A6 (soil mid-depth) and A9 (top of structure) are shown for the three earthquakes (table 3). These reveal a key feature of system behaviour. In Figure 5, for the small earthquake, accelerations appear to remain uniform through the soil column, with small amplification at structure level. In Figure 6, the moderate earthquake, base accelerations A2 are

© 2011 by Taylor & Francis Group, LLC

714

Figure 6. Moderate earthquake: a) A9; b) A6; c) A2.

followed by the structure A9 (with some soil filtering of higher frequencies) but it is noticeable that in the soil (A6), accelerations are much smaller, and at comparable levels with the smaller earthquake. This reduced acceleration is even more pronounced in the larger earthquake (Figure 7) where soil accelerations appear to be limited to a value of 0.05g. This is an important effect rarely considered in onshore earthquake engineering due to a lack of sufficiently soft soils. The applied shear stresses are too large for the soil to transmit, and soil therefore fails. In

Figure 7. Large earthquake: a) A9; b) A6; c) A2. Figure 8. Excess pore pressures recorded at P1 during: a) small earthquake; b) moderate earthquake; c) large earthquake.

effect, the soft clay is behaving like a fuse and limiting accelerations experienced nearer the surface. This reduced acceleration helps to explain the permanent displacements observed in Figure 4.

4 3.3 Pore pressure response

4.1

A number of readings of excess pore pressure were taken within the soil, enabling the equilibrium of the soil to be monitored. It was seen that the pore pressures recorded by each instrument read an identical static pore pressure prior to each earthquake. It was not anticipated that large pore pressures should be generated during shaking. The model as tested was 17 m deep (Figure 1), 55% moisture content and 6.7 kN/m3 buoyant unit weight, causing an actual fully-consolidated value of effective stress of 114 kPa at the base. It was previously suggested (Brennan et al., 2006) that small excess pore pressures of the order of 14% of initial vertical effective stress might be present during testing. However, inspection of the pore pressure transducer records show that prior to each earthquake, pore pressure readings returned to a constant value, implying hydrostatic conditions. As an example of the typical excess pore pressures observed, measurements from point P1 (see Figure 1) are plotted in Figure 8 for each of the three earthquakes in Table 3. Increasing amounts of transient pressures are observed in each earthquake. This is likely to be due to the vibration of the soil container and the caisson causing pressure waves through the soil. In addition, a rise in pore pressure of 8 kPa in the moderate earthquake and 15 kPa in the large earthquake are observed. As the instrument is 7.3 m deep where effective stress is 49 kPa, this rise in pore pressure corresponds to 30% of initial vertical effective stress. © 2011 by Taylor & Francis Group, LLC

DISCUSSION Sequential loading

The moderate earthquake shown in Figure 3 was the first to occur in this test series, and the only one to cause permanent displacement greater than 10 mm in any direction. As the movement was predominantly translational, it was unlikely to be purely due the correction of initial imperfections in the model. This displacement was not symmetric. Nearsymmetric loading was applied as this was a limitation of the actuator. As non-symmetric responses were possible, this loading symmetry is not thought to be responsible for the symmetric (and barely plastic) response during the large earthquake. The symmetry of loading is a variable requiring further study, and developments in actuator technology since these tests were performed will enable that in the near future.

715

4.2

Plastic displacement

Critical state soil mechanics suggests that the soil around the caisson is being yielded by the first cycles of this moderate earthquake. This causes the yield surface of the soil to expand. However, as subsequent cycling is no larger than the initial shaking (Figure 3) then the yield surface is no longer being expanded and decreasing amounts of plastic deformation occur each time. Models of similar behaviour have been used already in the non-seismic analysis of caissons in clay (e.g. Cassidy et al., 2006) as well as other foundation systems such as spudcans

(e.g. Martin & Houlsby, 2000) and on-bottom pipelines (e.g. Randolph & White, 2008; Morrow & Bransby, 2009) for example. This not only explains the shape of the horizontal displacement in Figure 3 but also the fact that the small earthquake (occurring at a later time) induces almost exclusively elastic response. Once the large strain of the large earthquake is applied, plasticity theory would suggest that, as such large loads had not previously been applied to the system, further plastic displacement would accumulate. Figure 4 shows that, in this test, it does not. There are two possible explanations for this. The first possibility relates to the observed accelerations. Although accelerations at input were seen to be greater in Figure 4 than Figure 3, the fuse-like behaviour of the soft soil has reduced accelerations at the structure in the larger case. Structure accelerations are still 60% larger than during the moderate earthquake, but extensive sub-yield plasticity can cause yield surfaces to expand hardening. The second possible explanation for this lack of plastic deformation during strong cycling concerns the data that has built the yield surface models. These models have been built using small displacements, and it is acknowledged that accommodating large strain deformation is a step yet to be taken. It is therefore possible that the nonlinearity of the soil as it becomes very deformed and distorted changes has a significant effect on the response that is outwith the expectation of existing models. 5

CONCLUSIONS

Based on consideration of the results presented, the following can be concluded: 1. Allowing some displacement to occur enables a structure to withstand earthquakes even though a conventional force-based analysis might suggest forces could exceed maximum design values. 2. Performance based design is therefore a significant consideration for seismic loads offshore. However, there is still little guidance as to how this might be achieved. 3. Although the structure tested could tolerate a large amount of displacement, very little permanent displacement was observed despite a series of six earthquakes being applied. 4. It was seen that his could be due to both the stresslimiting effect of the very soft soil, that limited measured accelerations, and also to an expansion of the yield envelope for the clay.

© 2011 by Taylor & Francis Group, LLC

It would therefore be useful for future work to examine the potential for yield-based constitutive models for the large-strain loading of an earthquake, with a view to developing useful methods for performancebased design. The improvement in earthquake actuation methods in recent years allows superior validation of such methods. REFERENCES Brennan, A.J. & Madabhushi, S.P.G. 2002. Design and performance of a new deep model container for dynamic centrifuge testing. In R. Phillips, P.J. Guo & R. Popescu (eds.), Proc. International Conference on Physical Modelling in Geotechnics, 183–188. Balkema, Rotterdam. Brennan, A.J., Madabhushi, S.P.G. & Cooper, P. 2006a. Dynamic centrifuge testing of suction caissons in soft clay. In C.W.W. Ng, L.M. Zhang & Y.H. Wang (eds), Proc. 2nd International Conference on Physical Modelling in Geotechnics: 625–630, Balkema, Rotterdam. Brennan, A.J., Madabhushi, S.P.G. & Houghton, N.E. 2006b. Comparing laminar and shear beam containers for dynamic centrifuge modelling. In C.W.W. Ng, L.M. Zhang & Y.H. Wang (eds), Proc. 2nd International Conference on Physical Modelling in Geotechnics, 171–176. Balkema, Rotterdam Cassidy, M.J., Randolph, M. & Byrne, B.W. 2006. A plasticity model describing caisson behaviour in clay. Applied Ocean Research 28(5): 345–358. Madabhushi, S.P.G., Schofield, A.N. & Lesley, S. 1998. A new stored angular momentum (SAM) based earthquake actuator. In T. Kimura, O. Kusakabe & J. Takemura (eds.) Centrifuge 98, 111–116. Balkema, Rotterdam Martin, C.M. & Houlsby, G.T. 2000. Combined loading of spudcan foundations on clay: laboratory tests. Geotechnique 50(4): 325–338. Morrow, D.R. & Bransby, M.F. 2009. The influence of slope on the stability of pipelines subjected to horizontal and vertical loading on clay seabeds. In Proc. 28th Int. Conf. on Ocean, Offshore andArctic Engineering, Honolulu, HI. Paper no. OMAE2009-79050. Newmark, N. 1965. Effects of earthquakes on dams and embankments. Geotechnique 15(2): 139–160. Priestley, M.J.N., Calvi, G.M., & Kowalsky, M.J. 2007. Displacement-Based Seismic Design of Structures. Pavia: IUSS Press. Randolph, M.F. & White, D.J. 2008. Upper-bound yield envelopes for pipelines at shallow embedment in clay. Geotechnique 58(4): 213–229. Schofield, A.N. 1980. Cambridge geotechnical centrifuge operations. Géotechnique. 30(3): 227–268. Torisu, S.S., Sato, J., Towhata, I. & Honda, T. 2010. 1g model tests and hollow cylindrical torsional shear experiments on seismic residual displacements of fill dams from the viewpoint of seismic performance-based design. Soil Dynamics and Earthquake Engineering (ahead of print). doi: 10.1016/j.soildyn.2009.12.016.

716

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

SEPLA keying prediction method based on full-scale offshore tests R.P. Brown, P.C. Wong & J.M. Audibert ExxonMobil Development Company, Houston, Texas, USA

ABSTRACT: Suction Embedded Plate Anchors (SEPLAs) were introduced to the anchoring market in 1998 in response to the growing industry need for anchors capable of resisting high vertical mooring loads. The concept of the SEPLA was to combine the advantages of suction piles (i.e., known penetration depth and geographical location) with those of VLAs (i.e., geotechnical efficiency and lower cost), while trying to avoid their drawbacks (i.e., large, heavy and costly to handle for suction caissons, and imprecise positioning for VLAs). In August 2003, two SEPLAs (Suction Embedded PLate Anchors) were load tested at the Kizomba A drill center Hungo West in Angola Block 15 (AB 15). The tested anchors were instrumented with inclinometers to measure the pitch of the fluke during keying, and then loaded using the bollard pull of the installation vessel M/V Dove. From the results of these load tests, a simple analytical method has been developed to predict SEPLA keying using only the geometry of the anchor and local in situ vane shear test measurements as inputs. Further development and verification of the keying model are expected as new centrifuge and field test data become available in the near future. It is hoped that the published test results can be useful to other researchers in providing data against which they can test their model predictions.

1 THE SEPLA CONCEPT The Suction Embedded Plate Anchor (SEPLA) was introduced to the anchoring market in 1998 in response to the growing industry need for anchors capable of resisting high vertical mooring loads. The concept of the SEPLA was to combine the advantages of suction piles (i.e., known penetration depth and geographical location) with those of VLAs (i.e., geotechnical efficiency and lower cost), while trying to avoid their drawbacks (i.e., large, heavy and costly to handle suction caissons, and imprecisely positioned VLAs). A SEPLA is a simple plate anchor that has a rectangular fluke, with a keying flap running the full length along its top edge (Figures 1 and 2). Figure 1. SEPLA general arrangement.

2

SEPLA DEPLOYMENT AND KEYING SEQUENCE

A suction caisson, called a suction follower, is used to implant the plate anchor to its design embedment. The SEPLA is mounted in two vertical slots cut at the bottom of the follower and is retained in that position by the mooring line and recovery bridle (Figure 3). The suction follower and the SEPLA are lowered to the seafloor, allowed to penetrate under their self-weight, and suction then is applied to finish embedment to the target depth, as is done for the installation of a suction caisson. Next, the mooring line and retrieval bridle that hold the SEPLA secure in the bottom of the follower are released by the installation ROV, and the follower is extracted by pumping it out

717 © 2011 by Taylor & Francis Group, LLC

Figure 2. Typical MODU SEPLA.

Table 1. Use of SEPLAs during the period 2000–2008 (data courtesy of Bob Wilde, InterMoor).

Year

Rig or Field name/ Client

Water depth (m)

2000

Homer Ferrington/BP

1400

2001

Ocean Confidence/BP

2134

2001

Deepwater Horizon/BP

760

2003

1384

2004

Xikomba/Kizomba A/ Esso Angola Kizomba B/Esso Angola

1097

2005

Gomez/ATP

1311

2006

Cajun Express/Chevron

2195

2006– 2007 2007

Kizomba C/Esso Angola

686

Transocean Marianas/BP

2073

Kikeh Alt. Temp. Moorings/ Murphy Tombua Landana/ Chevron

1400

2007– 2008 2008

4 Figure 3. SEPLA installation offshore Angola (Photos and illustrations reproduced with kind permission from Bob Wilde, InterMoor).

of the seafloor, leaving the SEPLA in place. The overboarding operation sequence is shown in a series of photos and sketches reproduced in Figure 3. The follower is then recovered to the installation vessel for deployment of the next SEPLA. The SEPLA keying flap is mounted with an offset hinge such that, during keying of the SEPLA, the soil pressures acting along its top edge force the flap to rotate with respect to the fluke, preventing it from travelling back up the disturbed installation track when the mooring line is tensioned. The mooring line is attached to the fluke by means of a twin plate steel shank.

3

SEPLA USE FOR MODU ANCHORING

Since their introduction to market in 2000, many SEPLAs have been installed, as listed in the table below: Typically, for MODU (Mobile Offshore Drilling Unit) applications, the fluke is a solid steel plate 2.5 to 3.0 m wide and 6 to 7.3 m long (Figure 2). For permanent installations, a SEPLA would be quite a bit larger, with the fluke being typically a double plate hollow fluke, 4.5 m × 10 m in size. © 2011 by Taylor & Francis Group, LLC

381

Number and type of SEPLAs installed 2 MODU Anchors 8 MODU Anchors 4 MODU Anchors 19 MODU Anchors 3 MODU Anchors 12 MODU Anchors 8 MODU Anchors 9 MODU Anchors 16 MODU Anchors 8 Temporary Anchors 8 Permanent Anchors

FULL-SCALE FIELD TESTS

Potential vertical pullout during keying is considered as a critical consideration in the design and installation of direct-embedment anchors like the SEPLA. The keying flap is one measure, which has been incorporated into the SEPLA to mitigate this risk. Although vertical pullout has not been experienced with SEPLAs, the benefit of the keying flap has never been quantified. In the design of the MODU SEPLAs for Angola Block 15 (AB15), which were installed in 2003, the designers lengthened the shank in an effort to maximize the torque or rotational driving-moment developed from the padeye load and, in this way, gained confidence that the fluke would rotate, resulting in an increase in the normal (to the fluke) component of the padeye load and decreasing the tangential component, which could generate pullout. However, to gain added confidence, the designers also chose to instrument and pre-key two SEPLAs in August 2003 at the Kizomba A drill center Hungo West in AB15. The tested anchors were instrumented with inclinometers to measure the pitch (and roll) of the fluke during keying, and then loaded using the bollard pull of the installation vessel M/V Dove.

5 TEST SEPLA GEOMETRY The test SEPLA geometry and the pitch and roll directions are shown in Figure 4. For reference, a

718

Table 3.

Hungo West undrained shear strength.

Depth below mudline (m)

Undrained shear strength (kPa)

0.0 2.5 2.5 6.7 14.8 14.8 25 25 40

3 5 9 9 15 20 30 35 48

Figure 4. Kizomba A SEPLA geometry.

Figure 6. ODBC Pull Test #1 using MV Dove, 28 Aug 2003. Figure 5. Transferring SEPLA from M/V Dove to supply vessel. Table 2.

Kizomba A SEPLA dimensions.

Fluke width Fluke height Horiz. distance from fluke face to padeye Keying flap height Vertical distance from anchor tip to padeye Padeye depth below mudline (as-installed) Anchor weight in air

6.71 m (22 ft) 3.05 m (10 ft) 2.59 m (8.5 ft) 1.22 m (4 ft) 1.52 m (5 ft) 26.3 m (86.1 ft) 235 kN (24 mt)

Figure 7. ODBC Pull Test #8 using MV Dove, 28 Aug 2003.

photograph of the SEPLA in this same orientation is provided in Figure 5. The SEPLA dimensions are also provided in Table 2. 6

However, the stage at 100% BP had to be terminated after a period of only 10 min. to prevent overheating of the vessel’s engines. The second test was done at No. 8. In this test, the engines were throttled to 100% BP as quickly as possible and held at 100% BP for a period of 100 min. The inclinometer measurements from these two tests are provided in Figures 8 and 9.

BOLLARD PULL TESTS

The two anchors tested were No. 1 and No. 8 at the Hungo West drill center. The water depths at No. 1 and No. 8 recorded at installation were 1217 m (3993 ft) and 1224 m (4015 ft), respectively. Figures 6 and 7 are plots of vessel displacement vs. time in each of the two tests. These plots are color-coded by percent of maximum bollard pull applied (% BP). The first test was done at No. 1. This was a staged test with 70, 80, 90 and 100% BP held for an average duration of 30 min. © 2011 by Taylor & Francis Group, LLC

7 ANALYSIS OF TEST RESULTS The sustained pulls at 70, 80, 90 and 100% BP were each treated as constant-load (constant-moment) experiments, with four experiments from the staged

719

Figure 8. Inclinometer data for Hungo West No. 1.

Figure 10. Example of rectangular hyperbola fitted to incremental pitch vs. time data. Table 4. pitches.

© 2011 by Taylor & Francis Group, LLC

Measured Pitch (deg.)

Extrapolated Pitch (deg.)

Anchor No.

% BP

Cumul.

Incremtl.

Cumul.

Incremtl.

1 1 1 1 8

70 80 90 100 100

0.94 1.37 3.05 14.47 9.57

0.84 0.45 1.68 8.56 9.57

0.95 1.42 5.14 14.97 –

0.85 0.50 3.77 9.06 –

Figure 9. Inclinometer data for Hungo West No. 8.

test at No. 1 and one experiment from the test at No. 8. The first step in the analysis of results was to determine whether or not rotational equilibrium (pitch constant vs. time) had been reached in each experiment. To make this determination, the pitch increment (from one load level to the next) vs. time was plotted for each experiment. It was found that none of the experiments achieved rotational equilibrium. The 10 to 30-min. durations of the experiments at No. 1 were not long enough for equilibrium to be reached or for steady-state (secondary) rotational creep to be observed. In the single experiment at No. 8, it appears tertiary creep was actually occurring at the time that the test was terminated. This behavior is interpreted to represent post-peak rotation of a rotation-softening system, similar to that observed in vane shear tests (VST) in clay. This analog will be pursued in greater depth later in this paper. For No. 1, an attempt was made to extrapolate the results of each experiment to equilibrium by modeling pitch vs. time under constant-moment using the equation for a rectangular hyperbola, assuming the potential steady-state rotational creep rate to be very small. An example of this rectangular hyperbola fitted to the measured data is shown in Figure 10. The end-of-experiment pitches, both measured and extrapolated, for all five experiments are provided in Table 4.

End-of-experiment measured and extrapolated

In the staged test at No. 1, the incremental pitch generally tends to increase with each increase in bollard pull. The exception to this trend is the 80%-BP experiment, for which the measured incremental pitch is only 0.45◦ , while that in the 70%-BP experiment was 0.84◦ . The higher incremental pitch in the 70%BP experiment is considered to be composed of three components: (1) initial seating rotation, (2) elastic rotation, and (3) plastic rotation. In contrast, the incremental pitch in the 80%-BP experiment is considered to be composed of only plastic rotation plus a small component of incremental elastic rotation. Following the termination of both tests, a counterrotation of approximately 0.5◦ was measured. This counter-rotation is considered to represent recovery of the elastic rotation associated with 100% BP loading. Based on these two consistent observations, the elastic rotation for the 70% BP is likely to be approximately 0.35◦ (70% of 0.5◦ ), while the incremental elastic rotation in the 80%-BP is probably only 0.05◦ (10% of 0.5◦ ). It is expected that, if the initial seating rotation and the majority of the elastic rotation were accounted for and subtracted from the incremental pitch in the 70%-BP experiment, the adjusted incremental pitch in the 70%-BP experiment would be less than that in the 80%-BP experiment.

720

Table 5.

Calculated padeye loads and angles.

Percentage of maximum bollard pull (%)

Calculated padeye load (kips)

Angle above horizontal (deg.)

70 80 90 100

222.4 271.9 316.4 361

88 82.3 78.6 75.7

8

INTERPRETATION

To compare the observed behavior of the anchors to the modeled behavior assumed in design, the measured (and extrapolated) rotations were compared to the predicted rotations. The AB15 SEPLA designers’ predicted rotations were based on their own proprietary model developed by the Norwegian Geotechnical Institute (NGI). For the range of undrained shear strength profiles across the Xikomba and Kizomba A drill centers, NGI’s model yielded predicted rotations of 19◦ to 41◦ for the 100%-BP case. The undrained shear strength profile for the Hungo West drill center (Table 3), at which the experiments were conducted, is approximately 10% stronger than the lower-bound profile on which the predicted rotation of 41◦ was based. The measured (and extrapolated) rotations were significantly smaller than this prediction, with extrapolated rotations of 9◦ and 15◦ in the two experiments. However, conventional wisdom at the time of the experiments was that throughout the keying process the fluke would rotate to be normal to the chain angle at the padeye.Although the chain angle was not measured in the experiments, the chain angle for the 100%-BP case was estimated (through reverse catenary modeling) to be 75.7◦ above horizontal (Table 5). A fluke normal to this estimated chain angle would have experienced a rotation of 14.3◦ , a value very close to the rotation observed at the end of the staged test at No. 1.

9 VANE SHEAR ANALOG MODEL As previously mentioned and as will be further explored in this section, the load (or shear stress)rotation behavior of the SEPLA is believed to be similar to what is observed during an in situ vane shear test (VST). The results of three such VSTs performed at the KizombaA drill center Hungo West inAngola Block15 (AB 15) are presented in Figure 11, as plots of normalized shear stress acting on a cylindrical shear surface as in a VST. Similarly, the results of the SEPLA keying have been plotted on Figure 12, as a plot of shear stress on a cylindrical shear surface with a diameter and a length equal to those of the fluke alone (i.e., ignoring the keying flap for these fluke rotations of less than

Figure 11. Vane shear tests at KizombaA drill center, Hungo West, Angola block 15.

Figure 12. Application of vane shear analog to SEPLA keying rotation.

20◦ , the maximum offset angle of the flap). The similarity in behavior, particularly the post peak softening, is notable, but could be a coincidence resulting from scatter in the limited number of data points. We have also plotted on Figure 12 the stress-strain (shear stress vs. rotation) curve of the most representative VST (red curve). As can be seen, the VST test reaches its peak at a slightly smaller rotation than for the SEPLA, probably due to the fact that the SEPLA’s behavior is somewhat affected by the remolding zone created by the insertion of the SEPLA (and follower) into the soil. To improve the match between the VST and SEPLA curves, we applied an empirical adjustment factor of 1.2 (i.e., 20% increase) on the VST rotation values. The vane shear analog is an improvement on the earlier NGI model. In the absence of a better model, the vane shear analog provides a simple and useful model, but it is sensitive to the input undrained shear strength and has not been calibrated to a sufficiently large database of tests, due to the limited data available from both field tests and model tests (i.e., centrifuge tests). The second and third authors are currently engaged

721 © 2011 by Taylor & Francis Group, LLC

The test results obtained during keying of the SEPLAs, in terms of torque vs. rotation, showed a behavior very similar to that observed in vane shear tests (VST). This prompted the development of a simple keying model that emulates the shear stress vs. rotation behavior observed during a VST test. In the absence of a better model, this vane shear analog model provides a simple and useful model. The second and third authors are currently engaged in an experimental and analytical effort to develop a more sophisticated and better calibrated model. In the interim, the test results have been reported in this paper to create a database against which other researchers can check the predictive capability of their own keying models.

in an experimental and analytical effort to develop a more sophisticated and better calibrated model, which will incorporate a model for keying flap behavior and which cannot be adequately addressed by the vane shear analog model. 10

SUMMARY AND CONCLUSIONS

During the past decade, SEPLAs have proven to be efficient and cost effective way to anchor MODUs in deep waters around the world. However, aspects of SEPLA behavior still warrant further investigation, particularly if SEPLAs are to be used in permanent mooring systems for floating producing facilities. Vertical pullout during keying is potentially a critical consideration in the design and installation of direct-embedment anchors like the SEPLA. Although vertical pullout has not been experienced with SEPLAs, the behavior of SEPLAs during keying, particularly the rotation of the SEPLA and the behavior of the keying flap are not well understood. To investigate the rotational of the SEPLA, two SEPLAs were instrumented and tested in August 2003 at the Kizomba A drill center Hungo West in AB15.

ACKNOWLEDGEMENTS The authors are grateful to the management of ExxonMobil Development Company, our affiliate Esso Exploration Angola (Block 15) Ltd. and its partners Sonangol, BP, ENI and Statoil for supporting the publication of this paper.

722 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Set-up of suction piles in deepwater Gulf of Guinea clays J.L. Colliat & D. Colliard Total, Pau and Paris, France

ABSTRACT: Suction piles were installed at several deepwater oil fields operated by Total in water depth ranging from 700 m to 1300 m offshore Angola and Congo. The increase in friction resistance was tested in-situ by extraction of piles at various set-up times, ranging between 1 day and 3 ½ years. The paper presents the installation behaviour and set-up resistance of these suction piles, with design consequences in term of average interface friction factor.

1

INTRODUCTION

Table 1.

Since 2001, suction piles have been used extensively in the Gulf of Guinea, notably at several deepwater oil fields operated byTotal offshoreAngola and Congo.At these sites, suction piles were installed for the moorings of two large FPSOs and off-loading buoys, one FPU, four riser towers, and a number of manifolds. With several more applications to come, the lessons learned may have some implications for future design studies. In particular, when considering the minimum consolidation time required between installation and hook-up or loading of suction piles with significant pull-out loading component (for TLM moorings, or for riser tower anchors under constant tension, for example), the pile shaft friction resistance and its increase with time are key design issues. The topic of the paper is the increase in pile friction resistance with time (“set-up” due to thixotropy and/or consolidation after installation) in soft deepwater West Africa clays, as determined from field extraction testing of suction piles at various set-up times ranging from one day to 3 ½ years. The database presented in the paper is for extraction of six suction piles installed at three different sites in water depth ranging from 700 m to 1300 m offshore Angola and Congo, for the moorings of one FPU and two FPSO vessels, and for a large RTA pile. The data were available from retrieval of suction piles related to operational constraints, and the estimate of the setup effect is a direct by-product. However, some of the installation or retrieval data might be incomplete and therefore involve an inevitable uncertainty.

2

SOIL CONDITIONS AND INSTALLATION BEHAVIOUR

The clays present similar general characteristics at the three sites under consideration, and the applications presented in the paper cover a range of pile sizes, © 2011 by Taylor & Francis Group, LLC

Characteristics of suction piles.

Water depth Dimensions – diameter – thickness∗ Penetration Subm. weight Set-up (days)

Site A FPSO

Site A RTA

Site B FPSO

Site C FPU

1300 m

1300 m

1300 m

700 m

4.5 m 20 mm 16.5 m 55 T 1&2

8.0 m 25 mm 19.0 m 175 T 8 & 35

4.9 m 25 mm 20.5 m 105 T 1260

3.8 m 20 mm 17.0 m 53 T 7

∗ Increased to 40–50 mm in the mooring padeye area for the FPSO and FPU piles.

from 3.8 m to 8.0 m in diameter and up to 20.5 m of penetration (Table 1). The piles were installed by a combination of self-weight and suction penetration. The self-weight penetration ranged between 50% and 67% of the final embedment, as is typically the case for suction piles in soft deepwater clays (Andersen et al. 2005, Colliat 2006). 2.1

Soil conditions

The soil conditions at the three sites are relatively uniform and composed of high plasticity clays (plasticity index IP equal to or over 100%), with a low submerged unit weight (γ
ranging between 3 and 5 kN/m3 ), typical of deepwater West Africa clays (Puech et al. 2005, Colliat et al. 2010). The undrained shear strength increases linearly with depth, with an apparent over-consolidation ratio equal to 1.5 down to about 15 m, which is attributed to ageing and possibly chemical bonding without any pre-consolidation stress from past overburden pressure. The clay sensitivity (i.e. the ratio of the intact to remoulded undrained shear strength obtained from laboratory tests) ranges between 2.5 and 4. For sites A and B, the former is considered as a lower bound value since the observed self-weight penetrations (generally larger than the

723

Figure 2. Site C CPT cone resistance profiles. Figure 1. Site B CPT cone resistance profiles. Table 2.

calculated ones) suggest a higher sensitivity of about 4 to 5. Typical CPT cone resistance profiles from Site B are given in Figure 1 (also representative of Site A), where the gradient of net cone resistance (i.e. the cone resistance corrected for hydrostatic water pressure effect, qnet ) is equal to about 21 kPa/m. At both Sites A and B, a 1 m thick stiffer seafloor formation is present, with a shear strength ranging between 7 and 15 kPa. The CPT cone resistance profiles from Site C are given in Figure 2, showing a slightly larger qnet gradient of about 23 kPa/m at this shallower depth site. The design undrained shear strength profiles are summarized in Table 2. These profiles are based on the direct simple shear strength su,DSS with reference to a reasonably conservative characteristic strength for the holding capacity analysis, and to an upper bound profile for the penetration and retrieval analyses. Note that, despite slightly higher measured CPT cone resistances, the design strength profile for Site C (about 22 kPa at 20 m depth) falls below those of Sites A and B (26–27 kPa at 20 m depth), which is related to a more conservative choice by the designer. 2.2 Installation behaviour The installation behaviour of Sites A and B FPSO piles is shown in Figure 3, giving the suction versus

© 2011 by Taylor & Francis Group, LLC

724

Site A Site B Site C

Design undrained shear strength profiles (su,DSS ). Holding capacity analysis

Penetration/Retrieval analyses

su = 6 kPa [0–3 m] su = 6 + 1.26(z-3) su = 4 kPa [0–2.5 m] su = 4 + 1.31(z-2.5) su = 1.6 + 1.05z

su = 8.5 kPa [0–3 m] su = 8.5 + 1.4(z-3) su = 4.6 kPa [0–2.5 m] su = 4.6 + 1.51(z-2.5) su = 3.2 + 1.27z

penetration curves. At final penetration depth, the installation suction is in the range 70–96 kPa and 102–135 kPa for Site A and Site B, respectively. The variation of about 20% in suction (corresponding in first approximation to the “as-built” penetration resistance) is similar to the scatter in measured CPT cone resistance (see Figure 1). With the Site A piles having a smaller diameter and half the weight of the Site B piles, a higher suction would have been predicted for these piles, when similar suction values are obtained between 14 m and 16.5 m depth at both sites. This lower than expected installation resistance at Site A is explained by two characteristics of the suction piles, i.e.: – Site A piles have two 400mm wide ring stiffeners in the mooring padeye area (at 4.5 m above pile tip), which is known to reduce the inside friction

Figure 3. Installation results for Site A and Site B FPSO piles.

(Erbrich & Hefer 2002, Dendani 2003, Andersen et al. 2005); – Initially painted and then sand-blasted, but with about 20% of their external wall remaining painted, a reduced outside friction is obtained at the soil-pile interface (see details in Colliat et al. 2007). The installation results for the Site C FPU piles are given in Figure 4. At this site, the scatter in suction response is significantly larger than the variation in CPTU cone resistance (see Figure 2), which is not fully understood to date. When excluding the two extreme values, the final installation suction ranges between 120 and 175 kPa. Compared to Sites A and B, the larger installation suction measured for the Site C piles is related to: (a) the combination of a slightly stiffer clay and smaller pile diameter, and (b) friction acting on a significantly larger surface, with Site C piles including a vertical 15 mm thick web plate over the whole pile length, giving an increase of 64% in inside friction area. The installation results for the three Site A riser tower anchors are described in detail in Colliat et al. (2007) and are not repeated here. It is only recalled that a very low penetration resistance was observed, which was explained by the following two main causes: (a) a dramatic decrease of the outside soil-steel interface friction because the RTA external surface was fully painted (contractor’s choice for protection against corrosion), and (b) an increased remoulding effect and reduction of the inside friction by 265 mm large ring stiffeners (last one 2 m above pile tip) down-dragging a mixture of seafloor soil and water. © 2011 by Taylor & Francis Group, LLC

Figure 4. Installation results for Site C FPU piles.

Moreover, the installation of Site A RTA that has been field tested was a bit hectic and deserves some additional explanations. Landed on the seafloor too quickly, the RTA initially self-penetrated to 15.5 m depth with a tilt of about 11◦ . After partial retrieval and re-penetration, the tilt was reduced to 4◦ (i.e. below the installation tolerance of +/−5◦ ), and the pile reached its final depth with a suction of 20 kPa. In addition to the paint effect, it is believed that this extremely low suction is also related to the rough selfpenetration phase, with a gap that opened along the pile side to an unknown depth. A second RTA, more properly installed close to the previous one, obtained a similar self-weight penetration (14.5 m), but with a significantly higher suction of 30 kPa at final depth, which is probably more representative of the actual soil-pile interface friction, and therefore is used in the analysis below. 3 3.1

RETRIEVAL AND SET-UP DATABASE Set-up database

The set-up database is summarized in Table 3. All retrieval cases were performed by over-pressurizing after the piles had reached their final penetration depth (i.e. short duration stoppages at intermediate depths of penetration are not considered here). The increase in friction resistance was field tested over various setup times, i.e.: (a) one to seven days, for suction piles that were re-located after being installed out of tolerances at Sites A and C, (b) eight days and about one month, for the field testing of a RTA pile at Site A, and

725

Table 3.

Site A FPSO Site A FPSO Site C FPU Site A RTA Site B FPSO ∗

Set-up database. Set-up time (days)

Final inst. & Initial retr. pressure∗

Initial retr./ Final inst. resistance

α factor

0 1 0 2 0 7 0 8 35 0 1260

−92 kPa +180 kPa −87 kPa +165 kPa −160 kPa +400 kPa −30 kPa +95 kPa +110 kPa −115 kPa +347 kPa

– 1.42 – 1.36 – 1.93 – 1.47 1.70 – 2.03

0.25 0.43 0.23 0.38 0.36 0.56 0.18 0.24 0.30 0.33 0.55

Negative value corresponds to installation suction.

3 ½ years, for the replacement of a FPSO anchor and mooring line at Site B. Figure 5 shows the results of the installation and retrieval of the two Site A FPSO piles that were retrieved one and two days after installation. The ratio of retrieval to installation resistance is obtained by dividing the retrieval resistance (equal to the initial retrieval pressure acting over the internal pile area, counteracting soil resistance and pile weight minus any pull-out crane load) by the installation resistance (equal to the sum of the pile submerged weight and the final installation suction acting on the internal cross section). With the pile submerged weight not always known accurately, due to some added weight from the rigging and the length of ground chain hanging on the pile side, the largest uncertainty in the database lies with the installation suction or retrieval pressure when the measurement is not done directly inside the pile and not corrected for venturi effects and hydraulic losses in the pumping system. The results given in Table 3 suggest a rapid increase in resistance, from 35–45% in one week to 70% in one month (with the exception of the Site C pile showing 93% of increase in one week), but very little gain between one month and 3 ½ years. A larger database is obviously needed, but it is worth mentioning that this result is in relatively good agreement with: (a) the results of thixotropy tests in high plasticity clays published by Andersen & Jostad (2002), Dendani (2003) or Colliat et al. (2010), and (b) similar data obtained at deepwater Gulf of Mexico sites, with Jeanjean (2006) obtaining similar results, and Dupal et al. (2000) mentioning a pump-out pressure equal to 2.5 times the pump-in suction for piles which were in place over six months. 3.2 Calculated average interface friction factor The following major assumptions were made for calculating the average interface friction factor α: – Same friction resistance for self-weight penetration and suction assistance zones, and friction adhesion constant with depth (i.e. no degradation effect)

© 2011 by Taylor & Francis Group, LLC

726

Figure 5. Installation and retrieval of two Site A FPSO piles.

– Similar friction resistance along both internal and external pile surfaces (which may be questionable for Site A FPSO and RTA piles with internal ring stiffeners). An average interface friction factor is then obtained from the following equation:

where Q = installation resistance (or retrieval resistance when the pile weight is subtracted); Aside = total pile shaft area (inside + outside); Atip = total pile tip area (tip + stiffeners); α = average interface friction factor; z = penetration depth; su = characteristic undrained shear strength; p
z = effective vertical stress; and Nc = end bearing capacity factor (7.5). When excluding the lowest values obtained logically for the fully painted Site A RTA pile, Table 3 suggests that the average interface friction factor is in the range 0.25–0.35 at time of installation, and increases to 0.40–0.55 in the longer term. The latter value is greater than the design average friction factor for the Site B FPSO piles. A cautious design method was applied for these piles, by considering an outer friction factor equal to 0.65 (from Andersen & Jostad, 2002), but an interface friction factor of 0.45 below the padeye stiffeners along the internal pile wall and no soil friction above the padeye inside the pile (based on the possibility of having internal ring stiffeners, which was actually not the case).

4

CONCLUSION

Although based on a too scarce database to draw firm conclusions, the set-up behaviour of suction piles in highly plastic Gulf of Guinea deepwater clays appears to be relatively similar to that of suction piles from Gulf of Mexico sites. In both cases, the increase in friction with time is rapid in the short term, i.e. 100% of increase in about one month, as compared to two months (proposed by Andersen & Jostad 2002) or about three months generally considered as a conservative design assumption. This rapid set-up effect is in good agreement with the results of thixotropy tests carried out on Gulf of Guinea deepwater clays (Colliat et al. 2010). However, the lack of significant increase in long term friction resistance is a design issue that would require further verification by the performance of specific field tests, possibly allowing separate measurement of inside and outside friction. The safety margin for the piles presented is not adversely affected since a cautious design method was applied, i.e. (a) in friction resistance, by using reduced internal friction adhesion based on the possibility of using ring stiffeners (Site B), and (b) in bearing capacity, either by not considering reverse end bearing (Sites A and C) or by limiting it to about 30% (Site B), as was required by the certifying authority (Colliat et al. 2007). 5 ABBREVIATIONS CPT DSS FPSO FPU TLM REB RTA

Cone Penetration Test Direct Simple Shear Floating Production Storage and Off-loading Floating Production Unit Taut Leg Mooring Reverse End Bearing Riser Tower Anchor

ACKNOWLEDGEMENTS The authors thank Total EP for the permission to publish this paper. Special acknowledgement is made of

the design studies made by NGI and Fugro France, and of the installation and retrieval operations performed by the respective crews of Acergy “Polaris”, Saipem “FDS”, Technip “Constructor” and Solstad “Normand Progress” (sub-contracted by Doris Engineering). REFERENCES Andersen, K.H. & Jostad, H.P. 2002. Shear strength along outside wall of suction anchors in clay after installation. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Kyushu. Andersen, K.H., Murff, J.D., Randolph, M.F., Clukey, E.C., Erbrich, C., Jostad, H.P., Hansen, B., Aubeny, C., Sharma, P. & Supachawarote, C. 2005. Suction anchors for deepwater applications. Proc. 1st Intern. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Colliat, J.L. 2006. Evaluation of suction piles and plate anchors from current deepwater mooring applications. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, San Francisco. Colliat, J.L., Dendani, H., Puech, A. & Nauroy, J.F. 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences. Proc. 2nd Intern. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Colliat, J.L., Dendani, H. & Schroeder, K. 2007. Installation of suction piles at deepwater sites in Angola. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Dendani, H. 2003. Suction anchors: some critical aspects for their design and installation in clayey soils. Proc. Offshore Technology Conference, OTC paper 15376, Houston. Dupal, K., Von Eberstein, B., Loeb, D., Xu, H., Grant, J. & Bergeron, B. 2000. Shell’s experience with deepwater mooring systems for MODU’. Proc. Int. Conference on Deep Offshore Technology, DOT, New Orleans. Erbrich, C. & Hefer, P. 2002. Installation of the Laminaria suction piles – A case history. Proc. Offshore Technology Conference, OTC paper 14240, Houston. Jeanjean, P. 2006. Set-up characteristics of suction anchors for soft Gulf of Mexico clays: experience from field installation and retrieval. Proc. Offshore Technology Conference, OTC paper 18005, Houston. Puech, A., Colliat, J.L., Nauroy, J.F. & Meunier, J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments. Proc. 1st Intern. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth.

727 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners and set-up time J.L. Colliat & H. Dendani Total, Pau and Paris, France

H.P. Jostad & K.H. Andersen NGI, Oslo, Norway

L. Thorel, J. Garnier & G. Rault LCPC, Nantes, France

ABSTRACT: The paper presents the results of pull-out centrifuge tests carried out on suction piles in soft deepwater Nigeria clay. The general characteristics of the centrifuge clay samples are presented, including inflight strength measurement by a miniature CPT probe. The suction piles have a diameter of 8 m and are 24 m long at 100 g prototype scale. Both un-stiffened and stiffened suction piles are tested under monotonic pull-out, and the effect of set-up is investigated by testing suction piles at various prototype times after installation, equal to 10 days, 6 months and 1 year.

1

2 TEST PROGRAM AND PROCEDURE

INTRODUCTION

The topic of the paper is a centrifuge testing program, part of an internal deepwater R&D Project at Total, performed by LCPC and interpreted by NGI, with the objective to explore the pull-out capacity of suction piles submitted to monotonic tension loading. With respect to current industry practice for the design of suction piles, the two main goals are: (i) to study the effect of large internal ring stiffeners on the pullout capacity, and (ii) to verify the variation in tension capacity with set-up time. As compared to most centrifuge tests carried out with Speswhite kaolin, one key feature is that natural clay is used in the current tests, namely reconstituted dried clay from a deepwater site offshore Nigeria. The paper presents the results of monotonic vertical tension loading tests, performed at 100 g on 1/100th scale model piles, 8 m diameter and 24 m long in prototype dimension. Two types of piles are tested in the same centrifuge bin, i.e. one smooth-walled pile (unstiffened) and the other pile with ring stiffeners. In each bin, the pull-out tests are carried out after the same set-up time equal to about ten days, six months and one year. The piles are installed by a combination of selfweight and suction penetration, simulating prototype conditions. The centrifuge clay sample is highly instrumented, in particular with pore pressure sensors in the pile tip area, to provide insight on the undrained capacity from skirt friction and reverse end bearing (REB, or passive suction) in rapid (i.e. undrained) and slow (partly drained) pull-out tests.

© 2011 by Taylor & Francis Group, LLC

The general background and scaling principles of centrifuge testing in offshore geotechnical engineering are described in Murff (1996) and are not repeated here. The description of the LCPC centrifuge and details about the general testing procedure and sample preparations are given by Thorel et al. (2010) or Raines et al. (2005), and only key points are summarised in the following sections. Unless noted, the data presented in the paper correspond to the prototype values. 2.1

Soil characteristics

The clay sample is reconstituted with dried clay coming from a deepwater Nigeria site, with zero salinity water and at a water content of 110% (as compared to about 130% in-situ in 1300 m of water). The mixture, placed in cylindrical containers, is consolidated in four layers by hydraulic jacking pressure, lasting about two months and giving 375 mm high clay samples. The container is then placed in the centrifuge, and final consolidation is achieved under 100 g for a period of about five hours just before testing. After this, the clay thickness has reduced to about 360 mm and it is checked that a minimum of 90% degree of consolidation is obtained. At the end of centrifuge consolidation, in-flight miniature CPT testing (12 mm diameter model CPT probe) is performed to measure the undrained shear strength of the clay su , and compare it to the target strength profile. The undrained shear strength is determined based on correlations between CPT and vane

729

The weight of the model piles, equal to about 3500 kN and 5400 kN for the U and S pile, is much larger than equivalent prototype piles weighing around 2000 kN. Therefore, the pile weight is counter-balanced in the centrifuge, and it is installed by a combination of self-weight and suction to simulate actual site conditions. The model piles and centrifuge clay sample are highly instrumented, and selected measurements are recorded continuously during installation and load testing (Thorel et al. 2010). The following main parameters are recorded: – Clay sample settlement and water level in the centrifuge sample – Pore pressures at several locations inside and outside the pile, and in the clay sample at some distance from the pile (see Fig. 3 inset) – Pile displacements and rotation – Clay plug displacement inside the pile – Load transferred to the pile and clay during installation – Vertical tension load during pull-out testing. 2.3 Test procedure Figure 1. Shear strength profiles from in-flight CPT tests.

tests carried out at 1 g in the clay samples, giving a cone factor Nk equal to 18.5. The cone resistance and corresponding shear strength profiles from five centrifuge samples are given in Figure 1. The miniature CPT has a maximum stroke of 25 m, and the shear strength below the suction pile is determined by linear extrapolation of the cone resistance profiles. The undrained shear strength increases more or less linearly with depth, with su (kPa) ranging between 1.08 z and 1.52 z (where z is the depth in meter), and the average strength gradient is close to the specified value of 1.3 kPa/m which is typical of deepwater conditions. The scatter in clay strength within the different centrifuge samples highlights the difficulty to obtain consistent and repeatable shear strength profiles with natural clay, clearly more difficult than with Speswhite kaolin.

2.2

Model piles and instrumentation

The model piles are made of stainless steel, 8 m in diameter and 24 m long in prototype dimension, which is similar to actual piles from deepwater Angola sites (Colliat et al. 2007). The pile wall thickness is equal to 50 mm and two types of piles are used, i.e. one smooth-walled or un-stiffened (U pile) and the other one stiffened (S pile) with five 500 mm wide internal ring-stiffeners equally separated by 4 m (see Fig. 1 inset). Such large ring stiffeners, i.e. 2 to 3 times larger than generally used in actual situations, corresponding to an area ratio of 0.256, were chosen for enhancing their consequence on the pile installation behaviour and pull-out capacity (the corresponding skirt tip area over base area of the U pile is equal to 0.025). © 2011 by Taylor & Francis Group, LLC

Ten pull-out tests were performed in five centrifuge bins. The test procedure, summarised in Table 1, gives more emphasis to the measurement of the undrained pile capacity from skirt friction and reverse end bearing (REB) in rapid (i.e. undrained) pull-out tests. Each time, one U pile and one S pile are installed and load tested in the same centrifuge bin. The distance between the piles and between each pile and the centrifuge cylinder wall is minimum 30 m (i.e. 3.75 pile diameters), thus ensuring no interaction effect. The piles are penetrated by self-weight to 10 m, after which suction is applied by pumping-out water from inside the piles. Details on the pile installation are given in Thorel et al. (2010). Both U and S piles are submitted to monotonic vertical pull-out testing by continuous extraction. Rapid pull-out tests are performed at a rate of 0.4 mm/s up to large vertical displacement, i.e. 4.0 m to 5.8 m or 50% to 70% of the pile diameter. One slow pull-out test was also carried-out (at a rate of 0.001 mm/s in an attempt to achieving drained conditions), and partial results are presented only, comprising the pile penetration phase and ultimate capacity (given in Figs. 2 and 4). The effect of set-up on the undrained capacity is explored by performing rapid pull-out tests with three different set-up times, approximately one year, six months and ten days. The test program is summarised in Table 1. 3 TEST RESULTS 3.1

Penetration phase

After self-weight penetration to 10 m, the suction phase is completed with a pressure gradient of 5 kPa/m

730

Table 1.

Summary of centrifuge test program.

U pile S pile Set-up time (days) Penetration U & S pile Test type ∗

Bin 2∗

Bin 3

Bin 4

Bin 5

Bin 6

A2U A2S 382

A3U A3S 368

B1U B1S 208

B2U B2S 31

B3U B3S 9.7

23.0 m 22.6 m Rapid

23.9 m 23.2 m Slow

23.0 m 22.8 m Rapid

23.4 m 23.2 m Rapid

23.3 m 23.2 m Rapid

Repeat of Bin 1, after soil plug failure during suction phase.

Figure 3. Rapid (undrained) pull-out tests B1U and B1S.

Figure 2. Measured penetration resistance for U and S piles.

and at a penetration rate of 0.2 m/min, with the objective to reach a maximum value of about 140 kPa at final penetration which is representative of actual prototype installations (Colliat et al. 2007). The applied suction ranged between 65–130 kPa (U piles) and 60– 160 kPa (S piles) at the final penetration of about 23 m. At this stage, the internal soil plug is in contact with the caisson top cover. The results of the penetration phase are given in Figure 2, showing that the measured penetration resistance is 30–45% higher in the tests with stiffeners (S piles) than in the tests with no stiffeners (U piles). There is also significant variation in the measured penetration resistance from one test to another, which may be expected when taking into account the variation in CPT resistance and inferred shear strength between the different centrifuge samples in Figure 1. 3.2 Pull-out capacity The results of the two rapid pull-out tests performed in bin 4 after a set-up time of about six months are presented in Figure 3, giving the comparison between the un-stiffened pile B1U and the stiffened pile B1S. © 2011 by Taylor & Francis Group, LLC

The load-displacement curves represent the net pullout pile capacity (i.e. gross tension load minus pile weight), and show a significantly larger capacity for the stiffened pile B1S. The maximum capacity is obtained at a vertical displacement between 0.8 m and 1.3 m (10–15% of the pile diameter), with only minor loss in capacity (softening) at large displacements. Similar pore pressure measurements are obtained in both the U and S piles, with the largest suction underpressures measured at the level of the pile tip. The mobilization of the REB capacity is well measured by the four pore pressure sensors located below the pile head, at the pile tip, and at a distance of half a diameter and one diameter below the tip. The suction measured at the pile tip when the peak resistance is mobilised is equal to 334 kPa and 406 kPa for the U and S pile, respectively, and the maximum suction (375 kPa and 500 kPa for the U and S pile) is obtained at a much larger pile displacement of 4–5 m. Similar results were obtained with the other tests. The mobilization of the REB capacity at the pile tip is also in agreement with the internal soil plug measurement sensors which suggest that the plug is lifted with the piles in all rapid pull-out tests.

4 4.1

INTERPRETATION OF TEST RESULTS Clay parameters

After completion of the centrifuge test, the measurements done in the clay samples give water contents and unit weights of 80% and 15.2 kN/m3 at the surface and of 65% and 16 kN/m3 at 24 m depth. Note that this is significantly different from the equivalent in-situ clay parameters, where the water content ranges between 160% (at seabed) and 120% (at 24 m depth),

731

and the unit weight between 12.5 kN/m3 (seabed) and 13.5 kN/m3 (24 m depth). This might be related to the use of zero salinity water to reconstitute the clay, but it is not known how much the use of zero salinity water and reconsolidation may have modified the properties of the reconstituted clay, in particular its sensitivity which is a key factor in the interpretation of the tests.
The normalized shear strength ratio sDSS u /p0 for the centrifuge clay is equal to 0.24. This is rather low when compared to the normalized shear strength ratio of 0.32 obtained from laboratory tests on the original Nigeria clay, or to equivalent Gulf of Guinea clays (Colliat et al. 2010). Another important aspect of the interpretation of centrifuge tests is the rate effect on the undrained shear strength of the clay, where the shear strength estimated from laboratory tests must be corrected for the much higher rates of shear strain applied in the centrifuge, both during the penetration phase and the pull-out tests (Andersen et al. 2005). Triaxial and DSS tests are normally run in the laboratory with a rate of shear strain of 3 to 5 %/hr. Prototype suction piles are installed with a penetration rate of about 2 m/hr, and the centrifuge model piles are penetrated at a rate of about 0.2 m/min (in model dimensions). It is uncertain what the rate of shear strain is, but if one assumes that the shear zone along the pile skirt has a thickness equal to the skirt thickness, i.e. 50 mm for the prototype and 0.5 mm for the centrifuge model, the rate of shear strain becomes about 4000 %/hr for the prototype and 600 times faster for the model pile. In the rapid pull-out tests, the model piles are extracted with a rate of vertical displacement of 0.4 mm/s = 0.024 m/min (model dimensions) or 1/8th of the penetration rate. Using the above strain rates and the correlation proposed by Lunne & Andersen (2007), the following rate corrections apply to the interpretation of the centrifuge tests: – Penetration phase: The shear strength from the laboratory tests should be multiplied by a factor of 1.4 for the prototype and 1.9 for the centrifuge model. However, the assumption that the shearing occurs in a thin zone with a thickness equal to the skirt wall thickness is uncertain. The zone may be wider or smaller, and some shear strain will also occur outside this zone or as sliding at the interface. The experience from prototype installations is that the skirt penetration resistance in some cases has agreed reasonably well with penetration resistance calculated based on DSS tests with standard rate of shearing divided by the sensitivity. Based on this, the strength for skirt penetration should be the standard DSS shear strength multiplied by 1.35 = 1.9/1.4 to account for the 600 times faster penetration in the centrifuge than in the prototype condition. – Undrained pull-out capacity: If the rate factor is 1.35 for penetration, the correction factor for a rate that is eight times smaller should be about 1.2. The strain rate effect is expected to be lower below the pile tip than along the skirt.

© 2011 by Taylor & Francis Group, LLC

Therefore, the definition of the required parameters for the interpretation of the centrifuge test results will inevitably involve some uncertainty. 4.2

Penetration phase

In the centrifuge tests, the large width of the ring stiffeners significantly increases the penetration resistance of the S piles, with the stiffeners causing more of the soil displaced to go outside the pile. Assuming that the clay plug will deform back to the pile wall shortly after passing a ring stiffener, the calculated effect of the ring stiffeners increases the penetration resistance by about 40% in the first 15 m and by about 15% at final penetration. The test measurements, however, give a relatively constant ratio between the penetration resistances for the S and U piles (Fig. 2). The variation in the measured penetration resistance from one test to another could not be related solely to the variation in CPT resistance in the centrifuge bins. When considering the shear strength given by the CPT, the average ratio between calculated and measured penetration resistances varies between 1.2 for tests A2U-A2S, and about 2.3 for tests B3U-B3S. If the rate effect for the penetration phase is assumed to be 1.35, it comes that the CPT-based shear strength profiles should be corrected by a factor ranging between 0.89 (bin 2) and 1.67 (bin 6). 4.3

Pull-out capacity

The interpretation of the centrifuge tests is based on: (i) the evaluation of the measured results, (ii) back-calculations with the NGI limit equilibrium program HVCap (Andersen & Jostad 1999) using different roughness factors, (iii) back-calculations where the outer friction is based on different factors times the measured maximum penetration resistance of the smooth-walled U piles, and (iv) back-calculations using the FEM program Plaxis (www.plaxis.nl). The key input data are the interface shear strength along the outer pile skirt which is expected to increase with set-up time, and the average undrained shear strength below the pile tip which also may change with time due to dissipation of the excess pore pressures generated by the penetration and the pile weight. The measured net pull-out capacities show the same degree of scatter as the measured penetration resistances. By normalizing the results by the measured penetration resistance of the U pile in the same centrifuge bin, a clearer picture is obtained. These normalized capacities are given in Figure 4 as function of the set-up time. It appears that:

732

– The net rapid pull-out capacity of the S piles is between 10 to 35% higher than for the U piles, and this factor increases with increasing set-up time after installation; – The normalized rapid capacity of the S piles increases by about 40% when the set-up period increases from ten days to one year. For U piles, the corresponding increase is lower;

Figure 5. Measured and calculated load-displacement curves. Figure 4. Normalized pull-out capacity versus set-up time.

– The capacities of the slow pull-out tests are 30–40% lower than the capacities of the rapid tests. This reduction is primarily assumed to be due to reduced undrained shear strength caused by the lower deformation rate, even if some drainage may occur. The measured excess pore pressures appear to be somewhat inconsistent, with the excess pore pressure at the top of the soil plug corresponding to a load carried by the soil plug between 60% and 95% (unrealistically high) of the net peak capacity (based on general trends for the U piles, roughly 75–80% of the net peak load is taken by the soil plug), and the excess pore pressure at the pile tip level being significantly higher than that at the top of the soil plug. Such a large REB capacity would give an unrealistically low contribution from outside friction resistance, i.e. lower friction during pull-out than during penetration or unrealistically low shear strength profile below the pile tip. The reason for these inconsistent excess pore pressures is not clear, but it might be related to local stress concentrations at the probe due to relative displacements between the soil and the probe (still to be confirmed). The measured peak pull-out capacities were backcalculated on the basis of the measured penetration resistance of the U piles, where the outer friction is calculated as a multiple of the measured maximum penetration resistance of the un-stiffened piles, and the shear strength below the pile tip is then selected in order to fit the measured capacities (using a bearing capacity factor of 9). In order to obtain a ratio between the outer friction and the net vertical capacity of about 20% for the un-stiffened U piles (with the 20% ratio based on the pore pressure measurements), the outer friction during the pull-out test is about equal to the friction during the penetration phase, but corrected for the different strain rate. Most consistent results between calculated and measured results are obtained by using a sensitivity of about 2. This is lower than the sensitivity of 3 to 4 for the original Nigeria clay but not unrealistic for the reconstituted soil. Due to a possibly lower sensitivity of 2, the effect of set-up

© 2011 by Taylor & Francis Group, LLC

733

Figure 6. Measured and calculated excess pore pressures versus applied net pull-out load (U pile, test B1U). The calculated results are shown as lines.

Figure 7. Measured and calculated excess pore pressures versus applied net pull-out load (S pile, test B1S). The calculated results are shown as lines.

for the U piles is assumed to be small, which fits with the measured results (see Fig. 4). On the other hand, the set-up effect for the S piles with ring stiffeners is significantly higher. The centrifuge tests were also analysed by using the FEM program Plaxis. The linearly elastic – perfectly plastic Mohr Coulomb model was used to model the undrained shear strength profiles and the reduced shear strength along the outer pile wall (defined by a roughness factor) taken from the previous back-calculations based on the measured penetration resistance of the U piles. As shown by Figures 5–7 corresponding to tests B1U-B1S after six months of set-up time, the calculated load-displacement curves

and excess pore pressure versus applied net vertical load agree rather well with the measured test results. It shows that the assumed ratios between the interface roughness for the piles without (U) and with (S) ring stiffeners and the intact shear strength give good agreement with the measured response, even though a simple linearly elastic – perfectly plastic material model with no shear-induced pore pressure is used. It is possible that a better agreement could be obtained by using a more advanced constitutive model. Finally, based on an undrained simulation with Plaxis, it is also found that the most likely reason for the higher measured pull-out capacities of the stiffened S piles is that a larger amount of soil displaced during penetration is moved outside the pile which leads to higher normal stresses along the outer pile wall. These higher normal stresses along the pile will give higher effective stresses and higher friction resistance along the outer pile wall with increasing set-up time. 5

CONCLUSIONS

The series of centrifuge tests presented had the objective to study the behaviour of 8 m diameter and 24 m high prototype piles installed by self-weight and suction in soft clay reconstituted from a deepwater Nigeria site. The paper describes the results of rapid pullout tests performed on both piles with and without ring stiffeners submitted to monotonic tension loading, with set-up times ranging between ten days and one year. The scatter in the test results is larger than expected and reduces the possibility to draw clear conclusions. In addition to the variation in shear strength profile in the different centrifuge clay samples, the water content of the reconstituted clay is significantly lower than that of the natural clay, and it is not known how far the reconstitution of the clay may have modified the soil properties. The measured penetration resistance is 30–45% higher in the tests with stiffened piles (almost constant with penetration depth), and the measured net rapid pull-out capacity of the same stiffened piles is 10–35% higher than for the piles without stiffeners (increasing pull-out capacity with set-up time). The higher pullout capacity of the stiffened piles is attributed to the

© 2011 by Taylor & Francis Group, LLC

larger amount of soil displaced out of the pile during penetration. However, the results obtained should be used with caution in design cases since the situation may be different for other ring stiffener configuration and other clay strength profiles.

ACKNOWLEDGEMENTS The centrifuge test program was supported as part of an internal deepwater R&D Project at Total, and permission to publish this paper is acknowledged. Special acknowledgement is made of the outstanding work in the performance of the model tests by the whole LCPC centrifuge staff in Nantes. REFERENCES Andersen, K.H., Jeanjean, P., Luger, D. & Jostad, H.P. 2005. Centrifuge tests on installation of suction anchors in soft clay. Ocean Engineering, 32, 845–863. Andersen, K.H. & Jostad, H.P. 1999. Foundation design of skirted foundations and anchors in clay. Offshore Technology Conference, OTC paper 10824, Houston. Colliat, J.L., Dendani, H. & Schroeder, K. 2007. Installation of suction piles at deepwater sites in Angola. Proc. Int. Offshore Site Investigation and Geotechnics Conference, SUT OSIG, London. Colliat, J.L., Dendani, H., Puech, A. & Nauroy, J.F. 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences. Proc. 2nd Intern. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Lunne, T. & Andersen, K.H. 2007. Soft clay shear strength parameters for deepwater geotechnical design. Proc. Int. Offshore Site Investigation and Geotechnics Conference, SUT OSIG, London. Murff, J.D. 1996. The geotechnical centrifuge in offshore engineering. Proc. Offshore Technology Conference, OTC paper 8286, Houston. Raines, R.D., Ugaz, O. & Garnier, J. 2005. Centrifuge modeling of suction piles in clay. Proc. 1st Intern. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Thorel, L., Dendani, H., Garnier, J., Colliat, J.L. & Rault, G. 2010. Installation process of suction anchors with and without stiffeners in Gulf of Guinea clay: centrifuge modelling. Proc. Intern. Conference on Physical Modelling in Geotechnics, ICPMG, Zurich.

734

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Numerical FEM and laboratory study of the bearing capacity factor Nc for plate anchors L.N. Equihua-Anguiano & M. Orozco-Calderón GPEMSO-STE Pemex, Exploration & Production, Mexico

P. Foray & M. Boulon Grenoble Institute of Technology, Laboratoire 3S-R, Grenoble, France

ABSTRACT: A study of the bearing capacity factor Nc for plate anchors is presented in this paper. Comparisons among FEM analysis and laboratory results were done. A soil corresponding to a soft normally consolidated clay was considered. Numerical FEM analyses were performed with the Plaxis® code, using an elasto-plastic model with a Mohr-Coulomb criterion. Undrained soil parameters and an adhesion factor α = 1 were used. Values of Nc factor for axi-symmetric and 2D FEM were obtained. Two geometries considering a perpendicular load applied in the anchor area and horizontal anchor position to different depths were studied. In the same way, anchor plates to a reduced scale were tested in a tank containing a soft soil. It was verified that the factors Nc reach constant values beyond a determined depth of the soil. Finally, the experimental and numerical values were compared with previous results reported in the literature.

1

INTRODUCTION

2

Plate anchors are frequently used as foundations solutions for offshore structures to transmit forces to surrounding soils at various depths. The capacity estimation of the anchor is nevertheless uncertain insofar as there are factors that have a large influence in its behavior such as the installation process, soil characteristics, geometry of plates, large numbers of methods used to calculate the holding capacity of the anchors, among others. Nowadays, to reduce uncertainties of design, plate anchors behavior is studied with the aim of standardizing the values employed, like Nc factor among others. Tools used in the practical design of this kind of anchor are the finite element analyses (Merifield et al. 2001, 2003; Equihua-Anguiano 2008), analytical models and laboratory tests (Gaudin et al. 2006). In this context, this article presents a study of Nc factor values, obtained from axi-symmetric and plane deformation FEM simulations done in Plaxis® code, as well as, results of scale down plate anchors tested in the laboratory. In both cases, the soil considered was soft clay with deep water sediment characteristics. The description of specific FEM analyses and laboratory test characteristics taken into account are described below. Comparisons with expressions found in literature were made. Finally, results obtained in this study were compared with previous Nc values obtained from field instrumented plates and centrifuge tests by Foray et al. (2005).

A sequence of finite element analyses was carried out using the Plaxis® code. In this part, the methodology followed is described for axi-symmetric (AXI) and plane strain (2D) analyses, as well as the results obtained.

2.1

FEM characteristics and parameters simulated

A vertical pull-out load F was applied in the anchor area in horizontal positions. Different depths D were studied. The horizontal position allowed making the comparison amongAXI and 2D results. Figure 1 shows the schematic plate anchor and the nomenclature used in this study.

735 © 2011 by Taylor & Francis Group, LLC

FEM PLATE ANCHORS STUDY

Figure 1. Schematic of plate anchor and nomenclature used.

Table 1.

2D and AXI-FEM plate anchors geometries.

Anchor

l m

L m

Dequi-axi m

t m

1 2

2 2

4 3

3.20 2.76

0.3 0.3

Table 2.

Geotechnical undrained parameters MC.

Soil

γsat kN/m3

◦

su kN/m2

Eu kN/m2

νu

Clay

17

0

1.6z

500su

0.49

φu

Figure 2. FEM model 60 m × 50 m used in 2D-MC analyses.

The shank and fluke of the anchor were idealized as simple plates. The failure load was obtained from the load-displacement curves, using a 0.1 L displacement failure criterion. Geometries used in 2D and AXI studies correspond to these ones shown in Table 1. For AXI studies an equivalent 2D plate diameter (Dequi-axi ) was used to ensure the same plate areas such as: L × 1 = π (Dequi-axi )2 /4. A high stiffness was assigned to the anchors in Plaxis®to guarantee a rigid behavior. The anchors are modeled weightless and the base plate was considered fully adhered to the soil. An adhesion factor of α = 1.0 was used for the interfaces, that corresponds to a rough surface. The soil simulated was a clay with deep sea sediment characteristics, using a constitutive MohrCoulomb (MC) model and undrained parameters. The parameters su and Eu were considered to vary linearly with depth z. Parameter su matches CPT profile in deep sea sediments in the Gulf of Guinea (Puech et al. 2005). The analysis made, did not take into account the suction developed. The geotechnical parameters used in Plaxis®are shown in Table 2. The values of Nc were calculated using:

where F = vertical pull-out load; A = plate anchor area; and su = undrained shear strength resistance at the anchor embedment depth.

2.2 Mesh used in FEM The mesh geometry in 2D and AXI conditions is sufficiently large to avoid the influence of the boundaries in the anchor plate response (Fig. 2). A prescribed displacement of 1.0 m was applied for all FEM analyses. Fifteen nodded triangle elements were used. © 2011 by Taylor & Francis Group, LLC

Figure 3. Load-displacement curves obtained from 2D-MC Plaxis® analyses for the Anchor 2 characteristics.

2.3

2D and AXI FEM results

Figure 3, shows load-displacement curves for 2D-MC analyses. Results correspond to the plate “Anchor 2” of Table 1. From load pull-out capacities, the factors Nc were calculated for each depth D analyzed. Ultimate load F was taken as described in 2.2. The same procedure was employed for theAnchor 1. Results obtained are shown in Figure 4. Two Nc − D/l normalized curves are presented; these correspond to results obtained for the Anchor 1 and Anchor 2. In the same figure are shown, comparisons with results obtained using the procedure suggested by Merifield et al. (2003) for the estimation of uplift capacity. The two curves Nco correspond to the break-out factor for the Anchor 1 and 2 respectively. Maximum and minimum values obtained of 2DMC finite element analyses are Nc = 14.9 and Nc = 12.9. Values are uniform from D = 3l depth. The

736

Figure 5. Factors Nc obtained from AXI-MC FEM analyses for circular anchors 1 and 2, and results obtained using the procedure suggested by Merifield et al. (2003) for estimation of uplift capacity.

Figure 4. Factors Nc obtained from 2D-MC FEM analyses for Anchor 1 and Anchor 2 and results obtained using the procedure suggested by Merifield et al. (2003) for estimation of uplift capacity.

difference presented among Nc values of Anchor 1 and Anchor 2 plates is not considerable and the curves followed the same trend. The value considered as representative is Nc = 12.9. High values (Nc = 14.9) obtained are not taken into account, due to the influence by proximity of the mesh surface and its influence in Nc values. Factor Nc = 12.9 obtained is higher that Nc = 11.2 reported by Merifield et al. (2003) for strip anchors with rough surfaces. This can be explained considering that the MC model gives high load capacities. However, we can consider a good agreement for both values calculated. Another factor considered to influence this higher results is, the full adherence considered in the contact soil-anchor in simulations, which increases the pull-out resistance. On the other hand, values for the break-out factor Nco depend on the embedment D/l and geometry L/l ratios. Curves obtained are lower than the values obtained in 2D-MC finite element analyses; nevertheless taking into account the effect of overburden pressure as Ncγ = Nco + γD/su , the calculated values are higher than the limiting value of the break-out factor Nc∗ = 11.2 obtained for a strip anchor. The effect of overburden pressure has an important influence in factors. Figure 5 shows results obtained in AXI conditions, considering the same parameters used in 2D analyses. Two Nc − D/Dequi−axi normalized curves are shown for the Anchors 1 and 2. A maximum value of Nc = 17 and a minimum of Nc = 15.1 were yielded. Values keep uniformity with the depth, the same as for 2D conditions. The higher Nc values corresponding to those obtained in 2D conditions are an effect of three-dimensional capacities given in AXI conditions.

© 2011 by Taylor & Francis Group, LLC

Figure 6. Failure kinematics to a pull-out load F in AXI conditions, with soil displacement vectors (left) at failure and contour displacements (right), embedment depth D = 12 m.

The difference presented among Anchor 1 and Anchor 2 curves is not considerable and the curves follow the same trend. Figure 5 shows the comparisons with the suggested procedure for estimation of uplift capacity for circular anchors included in Merifield et al. (2003). The break-out factor was calculated with Nco = S[2.56 ln(2D/Dequi−axi )] for Anchors 1 and 2. Values of the Nco are lower than this obtained in AXI conditions, however Nco increases, when is take the effect of overburden pressure, reaching Ncγ ≥ Nc∗ , Nc∗ = 12.6 value. The same figure shows the transition from shallow to deep behavior for D/Dequi−axi ≥ 3. Figure 6 shows the failure kinematics to a pull-out corresponding to a load F in AXI conditions, with

737

Figure 7. Failure kinematics considered in analytical methods formulation (Elkhatib & Randolph 2005).

Figure 9. Circular steel plate tested in “VisuCuve”.

Figure 8. Schematic plan of “VisuCuve” and coordinates of the scale plates and T-bar tests.

soil displacement vectors at failure and contours of equal displacements. A good agreement with kinematics used to formulation of analytical methods is observed (Fig. 7) (Elkhatib & Randolph, 2005). Values obtained in AXI conditions are compared with laboratory values shown in part 3 of this article, and after with values founded in the literature in section 4. 3

Figure 10. Shear strength resistance su measured using T-bar penetrometer.

LABORATORY ANCHOR TESTS

The laboratory test procedure included: a steel plate scaled down tested in two different depths, using the experimental “VisuCuve” tank developed in 3S-R laboratory by Orozco et al. (2007). Dimensions of this tank are showed in Figure 8. The circular steel plate has the following measures: L = 60 mm, t = 5 mm, L/t = 12 (Fig. 9). The embedment depths tested were: D = 310 mm (Plate 1) and D = 187 mm (Plate 2). The embedment ratios D/L are 5 and 3 approximately. The soft soil in the tank “VisuCuve” used in the plate tests presents characteristics of deep waters sediments (Orozco et al. 2007). It was formed by mixing kaolin, bentonite and natural water (liquid limit = 160%; plastic index = 132% and water content around 110%). The depth of the soil sample was 45 cm and was installed in the tank using a square trowel. To obtain factor Nc , the shear strength resistance su was measured with a T-bar penetrometer (Randolph et al. 1998). The coordinates of the test in the tank are shown in Figure 8. Figure 10 shows su profile measured. An average value of su = 4.5 kPa was employed to calculate Nc . In © 2011 by Taylor & Francis Group, LLC

the next step, the introduction of steel plate in the soil of “VisuCuve” was done (Fig. 11), after that the plate was covered by soil with equal water content. Before the application of the pull-out load, the soil was left 1824 hours in order to ensure regain of the shear strength of the remolded soil (thixotropy). Subsequently, the pull-out F load was applied with a constant speed of 1 mm/s in order to obtain undrained conditions. Figure 12 shows load-displacement curves obtained for the two different depths tested. The higher load F value obtained from the graph was taken to obtain factors Nc = 13.9 and Nc = 14.7 for D = 310 mm (Plate 1) and D = 187 mm (Plate 2) depths, respectively. Factor Nc = 14.7 found is near to axi-symmetric Nc = 15.1 value, obtained and presented in section 2.3 of this article. Applying the procedure for estimation of uplift capacity by Merifield et al. (2003), the values of Ncγ are 12.5 (Plate 1) and 9.5 (Plate 2), and the vertical load F = 160 N and 120 N, respectively. The Plate 1 is found in the transition zone from shallow to deep anchor behavior.

738

Figure 11. Introduction of the steel plate in the soil of the “VisuCuve”.

Figure 13. Soil failure surface for steel plate tested to D = 3.12 L depth.

Figure 14. Field instrumented plate. Figure 12. Load-displacement curves for D = 310 mm (Plate 1) and D = 187 mm (Plate 2).

Although the two values found in laboratory tests are smaller; there is a good agreement with FEM-AXI analysis using the MC criterion. Figure 13 shows soil failure surface for the steel plate tested to D= 3.1L depth. These results present a good agreement with AXIMC FEM analyses realized with Plaxis®. Values obtained from the literature are compared in next section.

4

COMPARISONS WITH LITERATURE RESULTS

A study of the pull-out behavior of plate anchors was carried out by Foray et al. (2005). Laboratory tests, half scale field tests (Fig. 14) and centrifuge tests were combined. © 2011 by Taylor & Francis Group, LLC

Results obtained are presented in Tables 3 and 4 for simulation of field and laboratory results. It is important to mention that the suction contribution in ultimate pull-out capacity was taken in account in the results, as well as, the anchor plate inclination. Table 3 shows the ultimate capacity obtained from field tests. There is a good agreement among the holding factor obtained by Foray et al. (2005) from field tests Nc = 15 and those obtained in results presented in this article corresponding to AXI-MC, Nc = 15.1 and Laboratory tests Nc = 14.7. Gaudin et al. (2006) obtained values of Nc among 12.3 y 13.5, for embedment ratios of 2.66 to 2.79 in case of jacked anchors tested in centrifuge tests. On the other hand, it is interesting to acknowledge Nc = 15 factors reported by Forrest et al. (1995) for long term design of plate anchors, and have a good agreement with those reported in this article and with a factor found by Foray et al. (2005).

739

Table 3.

Simulation of field conditions summary.

Anchor at 20 m, 45◦

Ultimate pull-out capacity (MN)

Holding factor Nc

4–6

15–17

– The suggested procedure for estimation of uplift capacity (Merifield et al. 2003) take into account the effect of overburden pressure as Ncg = Nco +γD/su , increasing the calculated values Nco . – Results of the Nc factor presented in this work are higher than the limiting value of the break-out factor Nc∗ = 11.2 for strip anchors and Nc∗ = 12.6 for circular anchors. Then the effect of overburden pressure has an important influence in factors.

Table 4.

Summary of laboratory results.

Tank No.

su kN/m2

Ultimate pull-out capacity N

Holding factor Nc

ACKNOWLEDGEMENTS

1 2 3

0.8–1.1 3.5–4.5 20

300–460 917–1150 7400–11600

5.4–7.8 4–4.8 6.2–9.5

The authors would like to thank the reviewers for their useful comments. REFERENCES

The values presented in this article correspond to a behavior without suction contribution, so the value corresponds to a long term behavior, where the suction does not have a participation in the final load contribution, due to the pore pressure dissipation in the time. 5

CONCLUSIONS

– Value of the bearing capacity factor Nc = 12.9 obtained in plane strain 2D-MC, close the value Nc = 11.2 calculated using the suggested procedure by Merifield et al. (2003), for strip anchors with rough surfaces. – Values for Nc = 15.1 in AXI-MC FEM analyses is higher to Nc = 12.6 obtained by Merifield et al. (2003). One factor considered to influence the higher results is, the full adherence considered for the contact in the soil-anchor simulation, which increases the pull-out resistance. – Elasto-plastic model with a Mohr-Coulomb criterion used in numerical FEM analyses is another factor that gives high load capacities, and therefore higher Nc . – Values of Nc = 15.1 factor obtained in AXI-MC FEM analyses and the laboratory tests in the tank “VisuCuve”, Nc = 14.7 have a good agreement. – A large influence of these results obtained withAXIMC FEM analyses done with Plaxis®, was the soil characteristics reproduced in laboratory conditions. – 3D model studies are necessary to carry comparisons with studies realized. – Suction contribution is important to be studied using numeric methods as FEM. – The Mohr-Coulomb criterion is a simple model used in numerical simulations; however this allowed doing comparisons with values reported in the literature. It is necessary to use constitutive models more complex, to have a better approximation of soil behaviour and also to obtain Nc values closer to those reported in the literature.

740 © 2011 by Taylor & Francis Group, LLC

Elkhabit S. & Randolph, M.F. 2005. The effect of interface friction on the performance of drag-in plate anchors, Frontiers in Offshore Geotechnics, ISFOG 2005. 171–177. Perth, Western Australia. Equihua-Anguiano, L.N. 2008. Modélisation des ancrages de structures offshore flottantes dans les grands fonds Marins. Thèse Doctorat, Laboratoire 3S-R, INPG. Grenoble, France. Foray, P.Y., Alhayari, S., Pons, E., Thorel, L., Thetiot, N., Bale, S. & Flavigny, E. 2005. Ultimate polluout capacity of SBM’s Vertically Loaded Plate Anchor (VELPA) in deep sediments, Frontiers in Offshore Geotechnics, ISFOG 2005: 185–190. Perth, Western Australia. Forrest, J., Taylor, R., & Bowman, L. 1995. Design Guide for Pile-Driven Plate Anchors. Technical Report No. TR2039-OCN, Naval Facilities Engineering Service Center. Port Hueneme, California. Gaudin, C., O’Loughlin, D., Randolph, M.F. & Lowmass, A.C. 2006. Influence of the installation process on the performance of suction embedded plate anchors. Géotechnique, 56, No. 6: 381–391. Merifield, R.S., Sloan, S.W. & Yu, H.S. 2001. Stability of plate anchors in undrained clay. Géotechnique, 51, No. 2: 141–153. Merifield, R.S., Lyamin, A.V., Sloan, S.W. & Yu, H.S. 2003. Three-dimensional lower bound solutions for stability of plate anchors in clay. Journal of Geotechnical and Geoenvironmental Engineering. ASCE, 129, No. 3: 243–253. Orozco, M., Foray, P. & Nauroy, J.F. 2007. Pipe-Soil Horizontal Dynamic Stiffness in Soft Soils. Offshore and Polar Engineering Conference & Exhibition, No. 2007JSC-267. July, Lisbon, Portugal. Puech, A., Colliat, J.L., Nauroy, J.F. &, Meunier, J. 2005. Some geotechnical specifities of Gulf of Guinea deepwater sediments. Frontiers in Offshore Geotechnics, ISFOG 2005: 1047–1053. Perth, Western Australia. Randolph, M.F., Hefer, P.A., Geise, J.M. & Watson, P.G. 1998. Improved seabed strength profiling using T-bar penetrometer. Proceedings Int. Conference Offshore Site Investigation and Foundation Behaviour – “New Frontiers”, Society for Underwater Technology: 221–235. London.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Caisson capacity in clay: VHM resistance envelope – Part 2: VHM envelope equation and design procedures S. Kay Fugro Offshore Geotechnics, Leidschendam, The Netherlands

E. Palix Fugro Offshore Geotechnics, Paris, France

ABSTRACT: A companion paper has described the development of failure envelopes for caisson capacity in clay for embedment (L/D) ratios varying from 0.5 to 4 and uniform undrained shear strength profiles. A quasi 3D finite element program (HARMONY) has been used to check the above work. This paper is mainly about using HARMONY to develop VHM ellipse/ellipsoidal equations for offshore caisson foundation design for a wide range of embedment ratios and undrained shear strength profiles. 1

INTRODUCTION

Offshore caissons are rigid circular steel cans usually vertical and installed by self-weight and suction. Originally for anchor piles, they are now commonly foundations for both fixed platforms and seabed structures. Resistance envelope equations for caissons (embedment ratio L/D > 1), even in uniform soil, are not currently available. Details of geotechnical design considerations are given in a companion paper (Palix et al. 2010).

2

FINITE ELEMENT ANALYSIS

2.1 HARMONY Proprietary program HARMONY computes the response of axisymmetric bodies subjected to nonaxisymmetric (horizontal, moment and twist) load. The soil material model is linear elastic – perfectly plastic (Mohr-Coulomb). By using Fourier expansions in the circumferential (θ) direction, the full 3D problem is solved with a quasi 2D analysis. This leads to less cost and time than for a comparable 3D analysis. Since its initial development (Griffiths 1985), HARMONY analyses have been made of various soilstructure interaction problems (e.g. Kolk et al. 2001). 2.2 Generic mesh design Eighteen cases (6 caisson embedment ratios L/D × 3 soil undrained shear strength su profiles) were analysed. The caisson diameter D was always 5 m, with embedded lengths L of 7.5, 10, 15, 20, 25 and 30 m. The lowest L/D value (1.5) is no longer a shallow foundation (L/D up to 1) and avoids possible problems with internal scoop failure. L/D ≈ 6 represents a practical © 2011 by Taylor & Francis Group, LLC

Figure 1. Caisson geometry, HM load sign convention and soil su profiles – Constant, Normally Consolidated and Stepped.

upper limit for suction installation in normally consolidated clay. To cover offshore conditions, “constant”, “normally consolidated”, and “stepped” su profiles were taken, with ez,su /L values of 1/2, 2/3 and 3/4 (Fig. 1). For consistency, each case used the same mesh, and mesh geometry was related to L and D. A 16 × 16 axisymmetric rectangular mesh of 8noded quadrilateral elements using “reduced” (2 × 2) integration was used. An inner (15 × 15) mesh of finite elements was surrounded by mapped infinite elements (Marques & Owen 1984). In the r-z plane, 8 rows × 5 columns of elements modelled the plug inside the caisson. The caisson had a top plate. In the radial direction, the caisson wall was surrounded on the outside by a “thin” steelsoil interface zone D/500 (10 mm) thick. The first column of soil outside the caisson wall was always

741

D/20 (250 mm) thick. Vertically, element rows immediately above and below the caisson tip were always D/20 (250 mm) and D/50 (100 mm) thick. The finite mesh extent was always 2L deep and 2L wide radially. Infinite mesh boundaries were at 6L depth and 6L radius. Material data included: (i) infinite tensile strength, (ii) constant soil su per element row, (iii) soil Eu /su = 500, νu = 0.49, (iv) E = 210e9 kPa for both plug and caisson to provide a rigid response, (iv) to give the correct α value, the outer interface zone had Eu and su values equal to 0.65 that of the adjacent soil, (v) soil submerged unit weight was 6.7 kN/m3 and Ko,total was unity. All analyses applied prescribed displacements to the caisson head. This facilitated “failure” definition (after 40 equal steps, the maximum displacement somewhere on the caisson was D/10) and allowed direct assessment of ez (=H load vertical eccentricity) from seafloor M and H reactions. Three harmonics were used (0, 1 and 3). The number of freedoms was 2304. Stresses were checked for yield at 30◦ intervals in the circumferential (θ) direction. 2.3 HARMONY verification To verify the performance of the “thin” interface elements, a laterally loaded disk model was analysed by extracting a single row of elements from the generic mesh: 8 finite elements modelled the soil. HARMONY Np,fixed errors were less than 0.5% of the analytical (Randolph & Houlsby 1984) solution when α = 0.65. To verify VHM loading, two laterally loaded caisson 3D FE benchmarks reported by Andersen et al. (2005) were simulated using the generic mesh. HARMONY agrees well with ABAQUS and BIFURC (Fig. 2). Extreme checks for axial capacity Vmax included a first run with “zero su ” soil below the caisson, and recovering the correct outer friction resistance Fo = α su,av,L π D L. Then, to check superposition, a second run with α = 0 gave the caisson end-bearing resistance component Qtip . 3

MH ELLIPSES AT V = 0

For each caisson L/D and soil su profile, on average 35 HARMONY analyses were made. Each applied a different displacement/rotation probe δx , θxz . “Fixed head” and “free head” conditions were included to establish four key data points Hmax , Ho , Mo and – Mmax (Fig. 3a). For caissons with L/D > 1, appropriate non-dimensional loads are M* = M/(D L2 su,av,L ) and H* = H/(D L su,av,L ). Typical M*H* load paths are given in Figure 3b. Pre-displacing the caisson to 0.9 |Hmax | minimised numerical problems associated with locating data points around -Mmax . At final displacement D/10, all “load-settlement” curves had flattened out. Even though the finite element method is a slight upper bound, it was assumed that final points lay on the resistance envelope. Mirroring was used (i.e. final point M∗ , H∗ gave –M∗ , –H∗ ). © 2011 by Taylor & Francis Group, LLC

Figure 2. HARMONY verification – laterally loaded anchor piles in NC soil su = 1.25z (Andersen et al. 2005) (a) caisson L/D = 1.5, Wsub = 300 kN (b) L/D = 5, Wsub = 1100 kN.

Figure 4 shows that M∗ H∗ final points lie on ellipses inclined at angles almost equal to ez,su /L. Rotated ellipses have the parametric form:

To find MH (rotation angle) and aMH , bMH (major and minor semi-axes’ lengths), least squares minimisation was used. Trials with differing numbers of data points and constraints showed that the best overall fits (highest correlation coefficient (Pearson’s r) R2 and lowest coefficient of variation) for all 35+ data points were obtained by simply using (M∗ , H∗ ) co-ordinates of 3 key data points Hmax , Ho and Mo , and a vertical gradient at Hmax . This fit procedure was applied to all 18 cases. Values of MH aMH , and bMH for the above 18 fits can be derived from su , Ho and Hmax using:

742

Figure 4. MH ellipses at V = 0: summary final data points, caisson D = 5 m, L/D = 1.5, 2, 3, 4, 5 and 6. Soil su profiles: (a),(b) Constant, ez,su /L = 1/2, (c),(d) NC, ez,su /L = 2/3 and (e),(f) Stepped, ez,su /L = 3/4.

4 V-HMAX ELLIPSOIDS Figure 3. MH ellipses at V = 0: (a) 4 key points Hmax , H0 , M0 and -Mmax (b) load paths and final data points, case L/D = 5, NC soil.

Correction parameter MH is due to the ellipse semi-major axis lying slightly below point Hmax . Figure 5 shows data points and fitted ellipses for 6 of the 18 cases. Overall fit quality is good. As can be seen on the insets, there is a slight deterioration for L/D = 5 near Hmax : data points are not equidistant from the ellipse major semi-axis, have a flat top surface, and a general shape resembling a “World War I tank” originally noted by Poulos & Davis (1980). Figure 6 is a key figure, plotting lateral bearing capacity factors Np,fixed and Np,free versus L/D. Trends of Np for both uniform and NC clay agree well with those presented by Randolph et al. (1998). At L/D ≈ 6, all Np values flatten out, and Np,fixed values are close to the Randolph & Houlsby (1984) value of 11.2 for L/D = ∞ and α = 0.65. Differences are due to caisson base shear/moment contributions and “shallow wedge” failure mechanisms close to seafloor. Before checking that MH resistance envelopes do not change shape when V = 0, it is first necessary to establish their shape in the V-Hmax plane. © 2011 by Taylor & Francis Group, LLC

For caissons with L/D values of 1.5, 3 and 5, Supachawarote et al. (2004) showed that V-Hmax resistance envelopes for NC clay were ellipsoidal and could be fitted with the equation

The above work was repeated and extended using HARMONY. For all 18 cases, twelve δx , δz “fixed head” (θxz = 0) displacement probes were made. Vmax and Hmax were obtained when δx = 0 and δz = 0 respectively. Envelopes remained ellipsoidal. Figure 7 shows resulting best fit aVH and bVH values. For the NC profile, good agreement was obtained between HARMONY and Supachawarote et al. (2004). Values of aVH and bVH are similar for both “NC” and “stepped” su profiles. In addition, for “constant” su profiles, the corresponding recommended simplified equations are:

743

Figure 7. V-Hmax plane: ellipsoid best fit parameters aVH and bVH versus L/D.

Figure 5. MH ellipses at V = 0: typical fits; caisson L/D = 1.5 and 5, Constant, NC and Stepped soil su profiles.

Figure 6. MH ellipse at V = 0: Np versus L/D (α = 0.65). H reference point location: caisson head.

Increased aVH and decreased bVH values are due to more competent soil near seafloor reducing interaction between V and H. 5

MH ELLIPSES AT V = 0

Applying V load decreases caisson capacity. To check that MH ellipses had similar shapes, six cases (caisson L/D = 1.5 and 5, each with 3 soil su profiles) were analysed for three additional load levels (V/Vmax = 0.6, 0.8 and 0.9). Vmax values had been established in Section 4 above. Figure 8a compares typical MH ellipses. Rotation angle MH is again essentially constant.

Figure 8. MH ellipses at V  = 0: Typical V/Vmax results; L/D = 1.5, NC soil (a) “as-is” (b) same major semi-axis lengths aMH .

744 © 2011 by Taylor & Francis Group, LLC

Figure 8b re-plots the ellipses all with the same aMH value. It is seen that the corresponding bMH decreases are modest. Therefore, it is reasonable to assume that, for non-zero V load, ellipse shape ratio aMH /bMH remain essentially unaltered: M and H reduce equally.

Figure 10. Example: effect of anchor load depth on caisson capacity in NC soil su = 1.25z, V/H = tan(30◦ ), L/D = 1.5 and 5.

8

DISCUSSION/ASSUMPTIONS

Assumptions in the above work included:

Figure 9. ellipse/ellipsoidal VHM envelope. VHM reference point location: caisson head.

Hence MH ellipses at V = 0 can be derived from MH ellipses at V = 0, plus load V and resistances Vmax and Hmax . 6 VHM ENVELOPE AND DESIGN EQUATIONS The complete VHM resistance envelope is given by rotated ellipses in the MH plane (Eq. 9) plus ellipsoids in the V – Hmax plane (Eq. 10). Factor Hmax,V /Hmax accounts for non-zero V load.

Figure 9 gives the resulting “tongue”-shaped VMH envelope. 7

DESIGN EXAMPLE

Anchor pile capacity is sensitive to lug level depth and chain/wire load angle. Supachawarote et al. (2004) and VHM equation results are compared on Figure 10. Good agreement for optimum depth (ez /L = 0.7) was obtained. For depths above optimum, the VHM equations slightly underestimate capacity decrease at L/D = 5. This is because the FE final data points are unsymmetrical (e.g. Fig. 5d). However, for depths below optimum, agreement is good, even for L/D = 5. This zone, where restoring rotation occurs, is of interest to suction pile anchor design.

745 © 2011 by Taylor & Francis Group, LLC

i ii iii iv v vi vii viii

zero caisson twist/tilt suction possible/no cracks unique soil shear strength su rigid caisson D = 5 m and 1.5 ≤ L/D ≤ 6 caisson top at seafloor clay-steel α = 0.65 MH ellipse symmetrical about semi-major axis Vmax and Hmax derived from FE analyses (not simplified analytical equations).

Ad (i) through (iii) these imply a symmetrical VHM envelope (e.g. identical V in tension and compression, no reduced reverse end-bearing). Ad (iv) use with caution for shallow foundations, but with confidence L/D > 6. Minor differences can occur for non-rigid caissons and/or other D values. Ad (v) provided that su,av,L is evaluated over the caisson embedded length, equations can also be used to assess torpedo pile capacity – pile head embedment has a small effect on Np,fixed at high L/D. Ad (vi) Randolph & Houlsby (1984) give guidance on Np,fixed variation with α for infinitely long caissons. All geotechnical foundation models are inaccurate to a certain degree. The resistance model described above is no exception, even though the basic building blocks (Randolph et al. 1998, Supachawarote et al. 2004) have been improved. Should assumptions be violated (for example reduced end-bearing or tensile gapping), it is also reasonable to consider this ellipse/ellipsoidal failure envelope for this modified situation without a significant decrease in model accuracy. 9

CONCLUSIONS

HARMONY is reliable and fast: a quasi 3D approach permitted over 1500 analyses in a rigorous and consistent manner. Based on the results in this paper, a VHM resistance envelope for caissons in clay may be reasonably approximated by three equations defining its ellipse/ellipsoidal shape. Envelope parameter values are functions of caisson geometry and soil shear strength profile. The equations obviate the need for

non-linear 3D FE analyses (except for assessing responses or soil reactions), and facilitate probabilistic and optimisation analyses.

10

NOTATION

su,z = soil undrained shear strength at any depth z su,av,L = average su,z between caisson head (seafloor) and caisson tip (L) L = 0 su,z dz/L t = ellipse parameter (0 < t < 2π) V = vertical load at caisson head (seafloor) Vmax = caisson vertical resistance (H & M = 0) Wsub = caisson submerged weight z = depth below seafloor

= clay-steel outer adhesion factor = caisson head lateral displacement = caisson head vertical displacement = caisson rotation REFERENCES = load factor = ellipse rotation angle Andersen, K.H., Murff, J.D., Randolph, M.F, Clukey, E.C., = soil undrained Poisson’s ratio Erbrich, C.T., Jostad, H.P., Jansen, B., Aubeny, C., = ellipse major semi-axis length Sharma, P. & Supachawarote, C. 2005. Suction Anchors = ellipse minor semi-axis length for Deepwater Applications, in Frontiers in Offshore = ellipsoid parameter Geotechnics ISFOG 2005: Proceedings of the First International Symposium on Frontiers in Offshore Geotech= ellipsoid parameter nics, Perth, Australia, Taylor & Francis, London, pp. 3–30. = caisson outer diameter Kolk, H.J., Kay, S., Kirstein, A. & Troestler, H. 2001. North = soil undrained Young’s Modulus Nemba Flare Bucket Foundations, Offshore Technology = H load vertical eccentricity with respect to Conference, Houston, Texas, U.S.A., OTC Paper 13057. caisson head (seafloor) Griffiths, D.V. 1985. HARMONY – A program for pre= M/H dicting the elasto-plastic response of axisymmetric bodez,su = analytical ez based on su,z ies subjected to non-axisymmetric loading, Report to L L Fugro-McClelland Engineers B.V., Simon Engineering = 0 su,z zdz/ 0 su,z dz Laboratories, University of Manchester, U.K. ez,Hmax = ez based on M/H at Hmax Marques, J.M.M.C. & Owen, D.R.J. 1984. Infinite EleFo = caisson outer skin friction resistance ments in Quasi-static Materially Nonlinear Problems, H = horizontal load at caisson head (seafloor) Computers and Structures, Vol. 18, No. 4, pp. 739–751. = non-dimensional H value H∗ Randolph, M.F. & Houlsby, G.T. 1984. The Limiting Pres= H/(D L su,av,L ) sure on a Circular Pile Loaded Laterally in Cohesive Soil, Hmax = caisson maximum “fixed head” horizontal Géotechnique, Vol. 34, No. 4, pp. 613–623. resistance (V = 0) Palix, E., Kay, S. & Willems, T. 2010. Caisson Capacity = Np,fixed L D su,av,L in Clay: VHM Resistance Envelope – Part1: 3D FEM Numerical Study, in Frontiers in Offshore Geotechnics Hmax,V = as Hmax , but with V = 0 ISFOG 2010: Proceedings of the Second International Ho = “free head” horizontal resistance (V&M = 0) Symposium on Frontiers in Offshore Geotechnics, 8-10 = Np,free L D su,av,L November, 2010, Perth, Australia. L = caisson embedded length Poulos, H.G. & Davis, E.H. 1980. Pile Foundation Analysis M = moment load at caisson head (seafloor) and Design, John Wiley and Sons, New York, Series in = non-dimensional M value M∗ Geotechnical Engineering, Chapter 7. 2 = M/(D L su,av,L ) Randolph, M.F., O’Neill, M.P., Stewart, D.P. & Erbrich, C. Mmax = caisson maximum moment resistance 1998. Performance of Suction Anchors in Fine-grained Calcareous Soils, Offshore Technology Conference, Hous(V = 0) ton, Texas, U.S.A., OTC Paper 8831. = caisson moment resistance (V & H = 0) Mo Supachawarote, C., Randolph, M.F. & Gourvenec, S. 2004. Np,fixed = “fixed head” lateral bearing capacity Inclined Pull-out Capacity of Suction Caissons, Proc. factor Fourteenth Int. Offshore and Polar Engineering Con= Hmax /(D L su,av,L ) ference ISOPE, Toulon, France, International Society of Np,free = “free head” lateral bearing capacity Offshore and Polar Engineers (ISOPE), Cupertino, pp. factor 500–506. = Ho /(D L su,av,L ) α δx δz θxz λL MH νu aMH bMH aVH bVH D Eu ez

746 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Installation and in-place assessment of drag anchors in carbonate soil M.P. O’Neill, S.R. Neubecker & C.T. Erbrich Advanced Geomechanics, Perth, Western Australia

ABSTRACT: The design of drag anchors as fixed moorings for offshore oil and gas facilities requires a detailed level of understanding of anchor behaviour in different soil types. An adequate level of safety in the system needs to be ensured, but also the anchor point must be prevented from moving during the design storm loads. In carbonate soils, factors such as sensitivity, cyclic loading and consolidation will heavily influence the level of preload required to ensure no anchor movement in service. These considerations are discussed in detail in this paper. 1

INTRODUCTION

Drag anchor installation involves preloading the anchor and chain system to a specified level in order to ensure sufficient capacity (or no movement) under the design loads. However, it is acknowledged that the in-place capacity of a drag anchor in cohesive sediments will differ from the preload applied to it. This is due to the combined effects of consolidation and cyclic loading. With fine grained carbonate soils, the difference between installation and in-place anchor capacities is amplified by the very high sensitivity they can exhibit and their generally strong susceptibility to degradation under cyclic loading. Following on from Neubecker et al. (2005), this paper describes simple anchor relationships to take account of the effect of soil sensitivity on the in-place capacity. The effect of consolidation and cyclic loading are also examined.

2

Figure 1. Anchor force system for N-R model.

where f is the anchor form factor, Ap is the total projected anchor area in the direction of travel, Nc is a bearing capacity factor (taken as 9), and su−m is the local monotonic undrained shear strength. The anchor resistance normal to the direction of travel, Tn , is defined as:

N-R ANCHOR EMBEDMENT MODEL

The drag anchor model described by Neubecker & Randolph (1996) (the ‘N-R model’) is a relatively simple method for predicting drag anchor behaviour in cohesive materials. The basis of the method is the assumption that two fundamental anchor parameters, namely the anchor form factor, f, and the resultant angle, θw , are unique to each anchor type and describe the behaviour of any sized anchor of that type in any cohesive soil strength profile. The method also assumes that a drag anchor travels along a plane parallel to its flukes inclined at an angle, β, to the horizontal (see Figure 1). The anchor resistance acting in the direction of travel, Tp , can be calculated at any depth as:

where θw is the inclination relative to the direction of travel of the total soil resistance force, Tw , which is the resultant of Tp and Tn (see Figure 1). The buoyant anchor weight, W
a , is then combined with Tw to form the resultant, Ta , which is taken as the anchor holding capacity orientated at the angle, θa , of the chain to the horizontal. The magnitude of θa may be determined using closed form solutions also developed by Neubecker & Randolph (1996) for embedded anchor chain behaviour in cohesive soils. The fundamental anchor parameters, f and θw , may be determined via calibration against field or model centrifuge test data, calibration against existing drag anchor simulation software or through analysis using the results of detailed finite element modelling. For this current discussion, the Vryhof Stevpris (Vryhof,

747 © 2011 by Taylor & Francis Group, LLC

The parameter Nc−DNV is the bearing capacity factor recommended by DNV (2000), which for this discussion has been taken as 12.5. The bearing capacity factors calculated using this simple equation are virtually identical to those derived using the equations presented in Zhou & Randolph (2009) for a T-bar, assuming no rate effect and with their strain softening parameter, ξ95 = 15, which is a reasonable value for many carbonate soils. Hence, during anchor penetration, Equation 3a and Equation 4a (describing the bearing and frictional resistances on the fluke respectively) can be re-written as: Figure 2. Anchor force system for component model.

2010) with a 50◦ fluke shank setting (typical for use in cohesive sediments) has been adopted as the ‘base case’ anchor type. The results of Stevpris model drag anchor centrifuge tests conducted in Speswhite Kaolin soil and presented in O’Neill (2000) indicate best-fit f and θw values of 1.4 and 35◦ respectively, which are adopted as the reference parameters herein. 3 ANCHOR FORCE COMPONENT MODEL An anchor force component model has been developed following the DNV (2000) general methodology. The soil resistance on the anchor projected in the direction of travel can be divided into four components, namely friction on the fluke, RFF , bearing resistance on the fluke, RFB , friction on the shank, RSF , and bearing resistance on the shank, RSB (see Figure 2). The bearing resistance components may be calculated as:

where AFB and ASB are the projected areas of the fluke and shank bearing components respectively in the direction of travel. The frictional resistance components may be calculated as:

where AFF and ASF are the friction areas of the fluke and shank respectively, α is an adhesion factor taken as 1/St and St is the soil sensitivity. Recent advances suggest that the bearing capacity factor, Nc , is strongly dependent on St . As a simple model, it can be considered that the bearing resistance on an anchor element can be divided into two components, namely one acting in virgin (i.e. intact) ground, and one acting in disturbed ground (i.e. in the ‘back-flow’ region) as the anchor penetrates through the soil. Hence, for the anchor force component model the value of Nc may be calculated as follows:

© 2011 by Taylor & Francis Group, LLC

Equation 3b and Equation 4b (describing the bearing and frictional resistances on the shank respectively) can be re-written in the same way. With regards to the Vryhof Stevpris model anchor tests reported in O’Neill (2000), it has been assumed that the Speswhite Kaolin soil used for the tests has a sensitivity of 2. It then follows that the N-R model fundamental anchor parameters calibrated for the Vryhof Stevpris anchor in Speswhite Kaolin (i.e. f = 1.4 and θw = 35◦ for St = 2) can then be used as a base case to match the anchor capacity calculated using the anchor force component model. This is achieved by varying the projected area of the shank in the direction of travel, ASB , on the basis that due to the complexity of the shank, its apparent area is not simply calculated. The other anchor area parameters AFB , ASF , and AFF are taken as fixed anchor characteristics based on the actual anchor geometry. Although the obtained shank bearing area may not strictly relate to the actual anchor geometry, it does permit the formation of an alternative anchor model which is comprised of separate anchor force components and which is calibrated against reliable model anchor centrifuge test data. 4

INCLUSION OF SOIL SENSITIVITY IN N-R MODEL

Unlike the anchor force component model described in the previous section, the ‘standard’ N-R anchor embedment model does not directly account for the sensitivity of the soil. However, it is possible to match the anchor capacity calculated by the N-R model at a given embedment depth with that calculated by the force component model for different values of St by altering the fundamental anchor parameters f and θw . Considering that θw should vary as a function of St such that the anchor bearing forces vary consistently with the bearing factor described by Equations 3a, 3b and 5, and the shear forces given by Equations 4a and 4b, an iterative approach can be employed to obtain

748

Figure 3. Anchor form factor and resultant angle versus sensitivity for 10 tonne Vryhof Stevpris anchor.

corresponding f and θw values for a given St , such that the anchor capacities predicted by the N-R model match those from the force component model. An example of the calculated f and θw values obtained using the process described above for sensitivities ranging between 1.2 and 20 and corresponding to a 10 tonne Stevpris anchor is presented on Figure 3. For increasing St , it can be seen that θw increases while f decreases. This is a simple reflection of the fact that as the sensitivity increases, both the normal and shear forces reduce, though the shear forces reduce more rapidly. Additional results from the same example are presented on Figure 4 in terms of the anchor capacity at the padeye, Ta , and the padeye embedment depth, da , during installation versus the horizontal padeye drag length, xa , normalised by the anchor fluke length, Lf . Profiles are included corresponding to the f-θw -St values contained on Figure 3. Generally, higher St implies an increased level of soil remoulding during installation, leading to lower anchor capacity and reduced anchor embedment. This trend between St , Ta and da is similar but less marked than presented by Aubeny & Chi (2010), who performed an assessment of drag anchor embedment in soft soils using a new numerical procedure. However, the Aubeny-Chi method ignored the effect of sensitivity on the bearing resistance components, and it is thought that this may explain the greater influence of sensitivity on the anchor trajectory obtained in their work. 5

Figure 4. Effect of sensitivity on installation capacity and embedment depth for 10 tonne Vryhof Stevpris anchor.

Equation 7 can be rewritten in terms of the consolidation factor, Ucons , and the cyclic loading factor, Ucy :

Tentative recommendations are made in DNV (2000) on values of Ucons and Ucy to be used for drag anchor design. However, these are not applicable to the high sensitivity carbonate soils encountered offshore Australia, since the recommended values for these factors only account for sensitivities up to 2.5, whereas carbonate soils generally exhibit much higher sensitivities. Instead, the consolidation and cyclic effects represented by Ucons and Ucy may be calculated for each individual term of the anchor force component model (i.e. RFB , RSB , RFF RSF ) and at each increment of the corresponding N-R anchor embedment analysis.

CONSOLIDATION & CYCLIC EFFECTS

As outlined in DNV (2000) and discussed in Neubecker et al. (2005), the ultimate holding capacity (or ‘characteristic anchor resistance’), Rc , of a drag anchor comprises the sum of the installation anchor resistance, Ri , and the predicted post-installation effects of consolidation (Rcons ) and cyclic loading ( Rcy ):

© 2011 by Taylor & Francis Group, LLC

5.1

Soil consolidation effects

Following installation, the effects of consolidation of the soil around the anchor will generally act to increase the anchor holding capacity. As detailed in Neubecker et al. (2005), it has been observed that in the case of vertical surfaces (such as pile walls or skirts) the loss of normal effective stress at the surface after the penetration process is only partially recovered after consolidation, due to the high soil

749

compressibility and arching mechanisms that generally occur in carbonate soils. However in the case of a horizontal surface, Neubecker et al. (2005) considered that the normal effective stress would ultimately revert to the in-situ conditions. An additional consideration is that, where the normally consolidated strength, su−NC , is greater than the remoulded strength, su−m /St the reconsolidated soil strength following full remoulding is assumed to not exceed the normally consolidated strength, su−NC , which for typical carbonate materials may be calculated as:


and where σvo is the in-situ vertical effective stress. Hence, for such cases the consolidated soil strength, su−con , acting on a frictional surface having an inclination to the horizontal, θ, may be calculated as:

The above scenario is typical of carbonate materials which generally have a relatively high sensitivity. As a consequence of the ‘dual-component’ nature of the bearing capacity factor, Nc , described by Equation 5, the fraction of the bearing resistance mobilised in disturbed ground (i.e. in the ‘backflow’ region) will be subject to consolidation effects. This may be accounted for by enhancing the bearing resistance of this component with due consideration of the appropriate re-consolidated undrained shear strength. Incorporating the soil consolidation effects described above into the component model expressions describing the bearing and frictional resistances on the fluke (RFB and RFF respectively) results in the following:

the remoulded zone. The excess pore pressures dissipate over time, and this consolidation process leads to a recovery of the effective stresses. In a simple consolidation model, it is assumed that as the excess pore pressures dissipate the effective stresses eventually recover to their original in-situ values, which as discussed previously, is also the assumption adopted by Neubecker et al. (2005) for horizontal surfaces (as incorporated in Equation 10). For the case of carbonate soils with high sensitivity, the high compressibility of the remoulded soil compared to the surrounding in-situ soil leads to the potential for arching mechanisms to develop. For these soils it cannot necessarily be assumed that complete dissipation of excess pore pressures will result in full recovery of effective stresses. In order to capture the complexity of such a consolidation process, a numerical analysis was performed using the finite difference program FLAC (Itasca, 2005) for a typical fine grained carbonate soil with a sensitivity of 20. The analysis represented a plane stain slice through an embedded anchor and surrounding soil perpendicular to the direction of anchor drag, for a particular geometry. The aim was to assess the degree of consolidation within the soil around the anchor over a 30 day period following installation. Although the anchor and soil properties adopted in the analysis were representative of a specific offshore location, the analysis results provide some useful insight into the post-installation consolidation process. The key result of the analysis is presented on Figure 5, which shows the average degree of consolidation, U, plotted against the elapsed time following installation of the anchor. The average degree of consolidation is defined as the average vertical effective stress at a given time divided by the in-situ vertical effective stress. It can be seen that after an elapsed time of 30 days, U had reached approximately 74%. However, it is also clear that the consolidation process is substantially complete after 30 days, and therefore the vertical effective stresses may only ever recover to about 75% of the in-situ stresses, unless creep processes increase this over time.

Similar expressions may be obtained for the bearing and frictional resistances on the shank (RSB and RSF respectively). The parameter U is defined as the average degree of consolidation and is discussed below.

5.2 Assessment of average degree of consolidation During installation, the process of anchor penetration results in the remoulding of soil around the anchor within the ‘back-flow’ region. This in turn results in the generation of excess pore pressures and a corresponding reduction in the effective stresses within

750 © 2011 by Taylor & Francis Group, LLC

Figure 5. Example assessment of consolidation progress within remoulded soil zone after anchor installation.

5.3 Cyclic loading effects After a period of soil consolidation, the anchor may be subjected to a period of one-way cyclic loading arising from a passing storm. The effects of this cyclic loading will generally act to decrease the anchor holding capacity. These cyclic loading effects on the in-place anchor capacity may be accounted for by calculating the cyclic undrained shear strength, su−cyc , acting on each anchor component. The magnitude of su−cyc may be calculated as:

where A, B and C are empirical coefficients determined via examination of laboratory cyclic soil
are strength test data. The parameters su−x and σv−x
set to su−m and σvo respectively for those force components acting in virgin (i.e. intact) soil, and to either su−NC or su−con and σv
(the effective vertical stress after remoulding and consolidation) respectively for those force components acting in disturbed soil (i.e. in the back-flow region). An example of the relationship between su−cyc /su−m
and su−m /σvo as described by Equation 12 for a typical carbonate sediment under one-way cyclic loading is presented on Figure 6. It can be seen that as the ‘apparent overconsolidation’ of the soil (represented
) increases, the relative undrained strength by su−m /σvo under cyclic loading (represented by su−cyc /su−m ) decreases. This aspect of carbonate soils is further discussed in Erbrich (2005).

5.4

An example set of results obtained from such an assessment is presented on Figure 7 for a 10 tonne Vryhof Stevpris with a 50◦ fluke-shank setting and a fluke length, Lf , of 3.41 m. The monotonic shear strength, su−m , was assumed to equal 5 + 1.5z kPa (where z is depth below mudline), while the effective unit weight of the soil, γ
, was taken as 5.5 kN/m3 . The average degree of consolidation, U, was assumed equal to 0.7, while the one-way cyclic undrained shear strength, su−cyc , was determined according to the relationship presented on Figure 6. For an assumed St = 5, the data contained on Figure 3 imply f = 1.01 and θw = 37.8◦ . Furthermore, it was assumed that an 80 mm bar diameter chain was connected to the anchor padeye with a friction coefficient, µ, of 0.4 and a zero uplift angle (i.e. horizontal) at the mudline. The results on Figure 7 show the anchor holding capacity at the padeye, Ta , during installation, after a period of consolidation and then after a period of cyclic loading, plotted against xa /Lf . Also included on Figure 7 are the anchor padeye and fluke tip embedment depths, da and dt respectively, plotted against xa /Lf . It can be seen that during installation and after a drag length of 40Lf , the anchor has embedded to a padeye depth of 14.6 m and has a holding capacity at the padeye, Ta (≈ Ri , see Equation 8) of 2.5 MN. If the anchor was then left at this depth for a consolidation

In-place anchor capacity – example

The N-R and force component drag anchor models described above may be used to calculate the embedment trajectory and resistance of an anchor during installation, as well as the subsequent in-place anchor holding capacity after any given amount of consolidation and after application of cyclic loading.

Figure 6. Typical one-way cyclic strength data for carbonate sediments.

© 2011 by Taylor & Francis Group, LLC

751

Figure 7. Example assessment of in-place capacity and embedment depth for 10 tonne Vryhof Stevpris anchor.

period sufficient to achieve U = 0.7, the analysis indicates that the holding capacity at the padeye would increase to 3.5 MN. Then, if the anchor were subjected to a period of one-way cyclic loading resulting from a passing storm, the holding capacity at the padeye would decrease to 3.0 MN. In terms of Equation 8, these capacities imply ‘overall’ values of Ucons = 1.40 and Ucy = 0.86. Although DNV (2000) does not provide any specific values of Ucy , it does suggest a Ucons range of 1.35–1.55 for St = 2.5.

the Bass Strait (Erbrich, 2005), where failure occurred at a cyclic preload considerably less than the static preload that had previously been applied. Drag anchors are generally not permitted to move under the design loads, but after cyclic failure there is the potential for ongoing and permanent anchor displacements (i.e. ‘ratcheting’) at loads less than the limiting cyclic failure load. Hence, avoidance of cyclic failure should be a key design objective. REFERENCES

6

SUMMARY & CONCLUSIONS

This paper has considered a simple force component model acting on different elements of a drag anchor. A method for determining variations in each of these independent force components (as a function of whether it is a shear or bearing force, element orientation, soil consolidation, cyclic load effects) has then been presented to determine the time dependent anchor capacity. Such considerations are particularly important in highly sensitive carbonate soil conditions where the anchor capacity can be shown to increase substantially after consolidation of the soil. Cyclic load effects will also influence the anchor capacity, by reducing the soil strength and force required to initiate movement of the anchor. It is only through detailed consideration of these combined effects that a final in-place capacity of a drag anchor can be established. Although it would be beneficial to obtain anchor field test data to validate the model predictions, there is a high level of confidence in the model given its sound theoretical basis. An analog to the problem of anchor behaviour in cyclically degradable soils was the spudcan preloading operations at the Trefoil field in

© 2011 by Taylor & Francis Group, LLC

Aubeny, C.P. & Chi, C. (2010), Mechanics of Drag Embedment Anchors in a Soft Seabed, Journal of Geotechnical & Geoenvironmental Engineering, ASCE, Vol.136, No. 1, pp. 57–68. DNV (2000), Design & Installation of Fluke Anchors in Clay, Recommended Practice RP-E301. Erbrich, C.T. (2005), Australian Frontiers – Spudcans on the Edge, Proc. International Symposium on Frontiers in Offshore Geotechnics – ISFOG 2005, Perth, Australia, Balkema: Rotterdam. Itasca (2005), FLAC: Fast Lagrangian Analysis of Continua, User Manual, Itasca Consulting Group. Neubecker, S.R. & Randolph, M.F. (1996), The Performance of Drag Anchor & Chains Systems in Cohesive Soil, Marine Georesources & Geotechnology, 14, pp. 77–96. Neubecker, S.R., O’Neill, M.P. & Erbrich, C.T. (2005), Preloading of Drag Anchors in Carbonate Sediments, Proc. International Symposium on Frontiers in Offshore Geotechnics – ISFOG 2005, Perth, Australia, Balkema: Rotterdam. O’Neill, M.P. (2000), The Behaviour of Drag Anchors in Layered Soils, PhD Thesis, Department of Civil & Resource Engineering, The University of Western Australia. Vryhof (2010), Anchor Manual 2010, Vryhof Anchors BV. Zhou, H. & Randolph, M.F. (2009), Resistance of Full Flow Penetrometers in Rate-Dependent and Strain-Softening Clay, Geotechnique, Vol. 59, No.2.

752

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Caisson capacity in clay: VHM resistance envelope – Part 1: 3D FEM numerical study E. Palix & T. Willems Fugro Offshore Geotechnics, Paris, France

S. Kay Fugro Offshore Geotechnics, Leidschendam, The Netherlands

ABSTRACT: This paper is about the development of HM resistance envelopes using PLAXIS 3D Finite Element Analyses for rigid circular caissons with embedment ratios L/D of 0.5, 1, 2 and 4. Clay with constant and normally consolidated undrained shear strength profiles was considered. The resistance envelopes were compared to results given by two Fugro software packages: CANCAP 2 based on limit equilibrium solutions and HARMONY, a quasi 3D finite element program. Satisfactory comparisons were obtained which improves the confidence in their use for the determination of foundation caisson dimensions. A companion paper deals with the development of VHM resistance envelope equations, applicable to offshore caisson foundation design and valid for L/D ratios between 1.5 and 6 and a wider range of undrained shear strength profiles. 1

INTRODUCTION

This section, which reviews caisson capacity assessment in clay using numerical modeling, shows that resistance envelopes for caissons under general VHM loading are not currently available, even in uniform soil. Offshore caissons are rigid circular open-ended steel cans (embedment ratio L/D > 1), usually vertical and installed by self-weight and suction. They are a technically efficient and economically effective solution to anchor floating vessels (such as FPSOs) in deep water or for jacket foundations (see Fig. 1a). Today, in the 2010’s, caisson foundations are also frequently used to support seafloor structures such as deepwater manifolds, PLEMs, PLETs, pumps, etc. (Fig. 1b). 1.1 Suction anchors Caissons were first used in the 1980’s as anchors to moor floating vessels such as FPSOs and FPUs. Design methods to assess anchor caisson installation resistance and subsequent holding capacity under inclined tension loading are now well established and documented (e.g. Andersen et al., 2005). In deep water normally consolidated (NC) clays, suction anchor caisson diameters D are typically in the range 4 m to 6 m with maximum achievable embedment ratios L/D around 6. A frequent assumption, based on limit lateral pressure distribution theory, is to take an optimum attachment point slightly deeper than 2/3L for NC clays to ensure a slight backward rotation (almost pure anchor translation and no crack behind the anchor). Uncertainty on the optimum point increases for soil profiles with a highly non-linear su profile and/or low

caisson L/D. Supachawarote et al. (2004) established ellipsoidal equations for the V-Hmax plane resistance envelope (i.e. no anchor rotation) for caissons with L/D values of 1.5, 3 and 5 in NC clay and a range of caisson-soil adhesion coefficients. 1.2

Unlike anchors, such caisson types are generally “stubbier”, with lower L/D values around 2–4. In service, they are subjected to numerous combinations of overturning H and M loads (generated by seismic activity, pressure or thermal expansion of the connected pipelines/flowlines). Vertical load components (dead weight) are generally small and compressive, and HM loads are usually overturning. Due to the high M load, associated failure mechanisms are usually rotational. Due to the number of VHM load cases, many analyses are required to optimize foundation design. When M/H is significant and L/D > 1, classical VHM bearing capacity theory, (e.g. ISO 19904-1, ISO (2003)), is inappropriate and 3D FE analysis is impractical. Offshore geotechnical engineers generally use a variety of tools: either based on upper-bound theory (Murff & Hamilton 1993) or limit equilibrium analyses (Fig. 5 and Fugro 2009). 1.3 VHM resistance envelopes Ideally, design of all the above foundation types should use VHM resistance envelopes. Poulos & Davies (1980) gave the first MH resistance envelope equations for rigid piles subjected to lateral earth pressures (pu ) either constant or varying linearly with depth.

753 © 2011 by Taylor & Francis Group, LLC

Small seafloor structure supports

Figure 2. Plaxis 3D mesh for L/D = 3 (52461 elements).

and various failure conditions. Soils other than NC clay are encountered in frontier areas. This was the reason for including a “stepped” profile in Kay & Palix (2010). 2

Figure 1. Suction caisson: recent developments are fixed platforms (Kolk et al. 2001) and manifolds (a) and (b). Sign convention (c). VHM reference point location: caisson head.

The resulting envelope is “laurel-leaf ” shaped (i.e. smooth-edged and lens shaped). Besides V = 0, the model does not account for non-uniform pu distribution (e.g. due to shallow wedge/deep flow failure mechanisms) and base shear/moment. Both effects are important for small L/D values, where rotational failure also occurs, and alter ez,su for anchors. Thanks to reliable 3D FE software running on cheap high performance digital computers, envelope methodology is re-emerging (e.g. Bransby & Randolph 1999 for rough surface foundations). More recently, papers on “shallow” strip foundations (L/D up to 1) have appeared by Bransby & Yun (2009) and Gourvenec (2008). The former assume HM loading of skirted or solid foundations and soil of constant or linearly increasing su , whereas Gourvenec researched VHM loading on solid foundations and constant su soil. Their work showed that various failure mechanisms are possible and limit resistances are related to embedment ratio. In addition, a suitable “load reference point” has to be defined: this affects the envelope shape, even though foundation capacity is unaltered. No elegant simple solutions exist for deeper caisson foundations in more complex soil conditions. In general, there is a requirement for a reliable method of assessing caisson capacity for all L/D under general (both restoring and overturning) VHM loading

© 2011 by Taylor & Francis Group, LLC

PLAXIS FINITE ELEMENT MODEL

All the finite element analyses used PLAXIS 3D version V2.2. Caissons of 5 m diameter were modeled as full rigid body avoiding any internal scoop mechanism. Only half of the problem needed to be represented because of symmetry in the lateral loading direction (M and H in the same direction) and the absence of torque load. 2.1

Mesh

PLAXIS 3D uses 15-nodes wedge elements. This type of element is a good compromise between the accuracy required for soil calculations, and memory and time consumption imposed by 3D calculations. Various mesh densities were investigated to use the maximum elements available for the longest caisson (L/D = 4) and to ensure a good accuracy. Meshes in the horizontal plane and at the caisson tip were the same for the four geometries studied. A typical three-dimensional FE mesh used for the L/D = 3 caisson is shown in Figure 2. The mesh extends 5D from the centre of the foundation and 4D beneath the foundation. Zero-displacement boundary conditions prevent out-of-plane displacements of vertical boundaries, and the base of the mesh is fixed in all three coordinate directions. 2.2

Material properties

Both constant and normally consolidated undrained shear strength profiles were considered (su = 10 kPa

754

and su = 2 + 1.5 z). The soil was modeled as an undrained, cohesive linear elastic-perfectly plastic (Tresca) material by using the PLAXIS MohrCoulomb strength model. A Poisson’s ratio ν = 0.45 was used to simulate undrained (i.e. no volume change) conditions, stiffness:strength ratio Eu /su = 500, submerged soil unit weight γ = 5 kN/m3 and initial stress ratio K0,total = 1. The caisson was modelled as a linear elastic material with very high stiffness and with the same unit weight than the surrounding soil (i.e. weightless). Isoparametric (curved) interface elements were considered with an α value (i.e. reduced interface Rint in PLAXIS) of 0.65. This value is commonly used for suction anchor design (Andersen et al. 2005) and is considered more representative of reality than a rigid interface. 2.3 Loading conditions The shape of the resistance envelope depends on the VHM load reference point, and there is a lack of consensus between researchers about the location of the reference point. For embedded shallow foundations, reference points have been defined at caisson tip or beneath the top cap (e.g. Bransby & Randolph 1999, Yun & Bransby 2007). In the present study, the reference point is at the top of the caisson (see Fig. 1c). This is consistent with offshore practice: for the vast majority of seabed support structures, external loads are provided by structural engineers at the point of load transfer, i.e. seafloor. Moment was applied by two diametrically opposite vertical loads acting on top of the caisson. FE simulations were performed considering no vertical load (V = 0). This was because Gourvenec (2007) showed that for circular shallow foundations the effect of vertical load on the resistance envelope was negligible for V < 0.3 Vmax. In addition, suction anchor pile resistance (in the V-Hmax plane) was found to be essentially independent of vertical load for V < 0.4 Vmax (Supachawarote et al. 2004) and was confirmed by Kay & Palix (2010). 2.4 Verification with suction anchor benchmark As the previous version 2.1 of PLAXIS was recognized to overestimate caisson capacity (Andersen et al. 2008), a prerequisite was to check the capability of the newly released version in which isoparametric interface elements were introduced. This was done by performing a benchmark and comparing PLAXIS 3D data with ABAQUS and BIFURC 3D data obtained by OTRC, COFS and NGI on one of the reference cases treated during the API/Deepstar industry sponsored project (Andersen et al. 2005). The anchor has been modelled with PLAXIS 3D as a rigid body with a submerged unit weight γ = 9 kN/m3 equivalent to (Wsub + Wplug )/AL. The V-H envelopes, obtained for pure caisson translation, agree well with the ones obtained by ABAQUS and BIFURC (Fig 3.). © 2011 by Taylor & Francis Group, LLC

Figure 3. Comparison of PLAXIS 3D Foundation and benchmark suction anchor computations after Andersen et al. 2005 (Case C2: D = 5 m, L/D = 1.5, su = 1.25 z, α = 0.65 and no crack at active side).

The above verification, and work done by Edgers et al. (2009), suggests that the isoparametric interface elements used in PLAXIS 3D version 2.2 accurately model interaction between curved soil-structure surfaces. 3

PLAXIS RESULTS

Resistance envelopes are presented in HM load space for V = 0. Both horizontal and moment loads are proportionally incrementally increased (i.e. M/H remain constant) until soil failure. The final loads set is defined as the (M,H) values obtained when the maximum displacement of the top of the caisson was about D/10. Due to caisson rigidity, the plastic plateau is generally obtained before this displacement. In the following, it was assumed that final points lay on the resistance envelope. Only the first quadrant (H and M positive) and fourth quadrant (H positive and M negative) are presented. The remainder of the envelope can be obtained by symmetry. The first quadrant corresponds to a load case where the moment is induced by the horizontal external load (overturning). This is interest for the design of seabed support structures where moments are induced by horizontal forces applied above the top of the foundation. The fourth quadrant is mainly of interest for design of jacket foundations and suction anchors when the moment is restoring (M counteracts with the horizontal load H) or when the load application point is below the seabed (lug level selected to minimize anchor rotation). Resistance envelopes under combined moment and horizontal loading are found to be close to rotated ellipses. For L/D > 1 the envelope is almost linear in the first quadrant. Figures 4 and 8 show that, close to Hmax , data points on the resistance envelope are not equidistant from the

755

Figure 4. non-dimensional MH envelope curves for uniform clay.

ellipse major axis. The shape of the envelope tends to flatten for slight backward caisson rotation. This phenomenon is accentuated when the L/D ratio increases. For a L/D ratio of 4, the shape of the HM envelope is close to those proposed by Poulos & Davis (1980) for unrestrained rigid piles under HM loading applied at top of the pile. On Figures 4 and 6, resistance envelopes are drawn in non-dimensional space using M* = M/(DL2 su,av,L ) and H∗ = H/(DLsu,av,L ) for L/D ≥ 1 and M∗∗ = M/(D2 Lsu,av,L ) and H∗∗ = H/(D2 su,av,L ) for L/D < 1. The variation of lateral bearing capacity factors Np,fixed and Np,free with L/D for uniform clay profile can be estimated by comparing respectively Hmax and Ho values given on Figure 4. The trends of Np,fixed and Np,free factors are given in a companion paper (Kay & Palix 2010) for 1.5 < L/D < 6 and three different soil profiles, and these confirm and extend data obtained by Randolph et al. (1998) with AGSPANC analytical software. It can be observed that for L/D ratios of 2 and above, the envelopes are similar. The grouping of the curves is mainly due to a common global shape and that Np,fixed and Np,free are almost constant for L/D ≥ 2.

4

COMPARISON WITH LIMIT EQUILIBRIUM MODEL

For preliminary design, the use of 3D FE software is not appropriate: engineers need less sophisticated and time consuming tools for capacity analyses. Hence, limit equilibrium analyses are often used at preliminary stage to identify critical load cases and optimize caisson geometry. For caissons (cans) with embedment ratios (L/D) up to around 2, CANCAP2 (Fugro 2009) can be used to compute Factors of Safety against rotational failure under combined VHM loading in layered clay soils. Three potential rotational failure modes are considered (Fig. 5): 1 rotational failure of the whole can in the direction of the overturning moment (shallow rotational failure, SR), © 2011 by Taylor & Francis Group, LLC

Figure 5. CANCAP2 rotational failure modes: (a) Shallow rotational failure, SR, (b) intermediate rotational failure IR (c) deep rotational failure DR. Typical CANCAP2 results (L/D ≈ 0.5, NC clay) DR mode (d) failure inside can, minimum FOS = 1.1, (e) failure outside the can, minimum FOS = 1.9.

2 rotational failure of the whole can, combined with active and passive failure wedges at part of the back and front of the can wall (intermediate rotational failure, IR) 3 bearing capacity failure – either tangential to can toe level or penetrating soil plug within can – combined with active and passive failure wedges at the back and front of the can wall (deep rotational failure, DR). Circular cans are usually modeled as a square can with the same cross sectional plan area. Failure is assumed to occur in a vertical “slice” of soil enclosing the square can. Resisting forces (including frictional forces on the two sides of the failing soil slice) in the plane of the applied VHM loads are computed using limit equilibrium solutions for various soil segments. These resisting forces use the maximum soil shear stress values divided by an unknown material factor (γm ) which is found from moment equilibrium of the resisting forces and the external (VHM) loads. Analyses are done for various centres of rotation (on and off the centre-line) until the minimum γm value is found. In the present paper, deep rotational failure was forced to be tangential to the can toe level (Fig. 5e) in order to simulate a solid foundation. For V = 0, HM envelopes determined using PLAXIS 3D and CANCAP2 are comparable (Fig. 6). Insets show that PLAXIS soil displacements around the caisson emulate those in CANCAP2. An exception is for H values close to Hmax . This corresponds to a shallow wedge / deep flow failure type model for

756

Figure 8. Comparison between PLAXIS 3D and HARMONY HM envelopes (V = 0 and su = 10 kPa).

5

Figure 6. Comparison between PLAXIS 3D and CANCAP2 HM envelopes (V = 0): (a) for Su = 10 kPa & (b) for normally consolidated clay and L/D = 1.

An alternative time saving approach for caisson modelling can be to use a quasi 3D finite element program. HARMONY (Fugro 1994) computes the response of axisymmetric bodies subjected to non-axisymmetric load. Details of HARMONY model, verification and results are given in a companion paper (Kay & Palix 2010). Basically, HARMONY was extensively used to analyze six caisson L/D ratios between 1.5 and 6 and three su profiles. Based on these data, ellipse/ellipsoidal equations were developed for defining the complete VHM envelope. Figure 8 compares results for L/D values of 2, 3 and 4. HARMONY results are seen to superimpose on every part of the corresponding PLAXIS 3D envelope. 6

Figure 7. (a) Soil displacement around the caisson for M/H =− 11, (b) incremental displacement at 15 m depth (L/D = 4 and uniform soil profile).

(almost) pure lateral translation of the caisson. Nonrotational mechanisms are (cautiously) not considered in CANCAP2. On all other parts of the HM envelopes, good matches were found between the lines of failure and centers of rotation obtained by both methods (Fig. 6b). The use of CANCAP2 is limited to caisson with L/D up to 2. For higher L/D values, CANCAP2 failure envelopes become unconservative, particularly in the fourth quadrant (H positive and M negative). Flow mechanism starts to occur at caisson base (Fig. 7). This type of mechanism is not considered by CANCAP2. © 2011 by Taylor & Francis Group, LLC

COMPARISON WITH HARMONY

CONCLUSIONS

The aim of the work presented here was to investigate numerically the shape of HM resistance envelope for caisson foundations in clay with embedment ratios between shallow foundations (L/D < 1) and short rigid piles. Based on the above results, HM resistance envelopes (at V = 0) have been found to be dependent on the embedment ratio L/D and the soil shear strength profile. The global envelope shape is almost elliptical. A companion paper (Kay & Palix 2010) gives VHM ellipse/ellipsoidal equations for the design of offshore caisson foundations with L/D values of 1.5 and above. Simpler tools than 3D FE software can be used for preliminary design: limit equilibrium analysis using CANCAP2 has proved to be reliable and efficient for caissons with L/D ≤ 2, with or without internal crossplates, and under significant V load. For final design, 3D FE analyses will still be needed to assess foundation responses and soil reactions. 7

NOTATION

A = caisson area = πD2 /4 α = clay-steel outer adhesion factor

757

D ez ez,su γm H H∗ H∗∗ Hmax Ho

L M M∗ M∗∗ M

Mmax Mo Np,fixed Np,free pu su,z su,av,L

V Vmax Wsub Wplug z

= caisson outer diameter = H load eccentricity with respect to caisson head (seafloor) = analytical ez based on su L L = 0 su,z z dz / 0 su,z dz = material factor = horizontal load at caisson head (seafloor) = non-dimensional H value [−] = H/(D L su,av,L ) = H/(D2 su,av,L ) = caisson maximum (“fixed head”) horizontal resistance (with V = 0) = Np,fixed L D su,av,L = caisson “free head” horizontal resistance (with V = M = 0) = Np,free L D su,av,L = caisson embedded length = moment load at caisson head (seafloor) = non-dimensional M value = M/(D L2 su,av,L ) = M/(D2 L su,av,L ) = transformed moment = maximum moment resistance (V = 0) = caisson moment resistance (with V = H = 0) = “fixed head” lateral bearing capacity factor = Hmax /(D L su,av,L ) = “free head” lateral bearing capacity factor = Ho /(D L su,av,L ) = ultimate lateral resistance = soil undrained shear strength at any depth z = average su,z between caisson head (seafloor) and caisson tip (L) L = 0 Su,z z dz / L = caisson vertical load at caisson head (seafloor) = caisson vertical resistance (with H = M = 0) = caisson submerged weight = soil plug submerged weight = depth below seafloor

Deepwater Applications”, in Gourvenec, S. and Cassidy, M. (Eds.), Frontiers in Offshore Geotechnics ISFOG, Perth, 2005, Taylor & Francis, London pp. 3–30. Andresen, L., Edgers, L. and Jostad, H. P. (2008), “Capacity analysis of suction anchors in clay by PLAXIS 3D Foundation”. PLAXIS Bulletin, issue 24, October, 5–9. Bransby, M.F. & Randolph, M.F. (1999) “The effect of embedment depth on the undrained response of skirted foundations to combined loading”, Soil Found. 39, N◦ . 4, 19–33. Bransby, M.F. & Yun, G.J. 2009. “The undrained capacity of skirted strip foundations under combined loading” Géotechnique 59, N◦ . 2, 115-125. Edgers, L., Andersen, L. & Jostad, H.P. 2009. “Evaluation of loading-carrying capacity of suction anchors in clay by 3D finite element analysis”, 1st Int. Symp. on Computational Geomechanics (ComGeo I), Juan-les-Pins, France. Fugro. 1994. HARMONY user’s manual HARMONY 00.09. Fugro. 2009. CANCAP2 user’s manual CANCAP2 00.17. Gourvenec, S. 2007. “Failure envelopes for offshore shallow foundations under general loading”, Géotechnique 57, N◦ . 9, 715–728. Gourvenec, S. 2008. “Effect of embedment on the undrained capacity of shallow foundations under general loading”, Géotechnique 58, N◦ . 3, 177–185. ISO International Organization for Standardization. 2003. “Petroleum and Natural Gas Industries – Specific Requirements for Offshore Structures – Part 4: Geotechnical and Foundation Design Considerations”, International Standard ISO 19901-4:2003. Kay,S. & Palix, E. 2010. “Caisson Capacity in Clay: VHM Resistance Envelope – Part 2: Envelope Equation and Design Procedures”, in Frontiers in Offshore Geotechnics ISFOG 2010: 8-10 November, 2010, Perth, Australia. Kolk, H.J., Kay, S., Kirstein, A. & Troestler, H. 2001. “North Nemba Flare Bucket Foundations”, Offshore Technology Conference, 30 April-3 May 2001, Houston, Texas, U.S.A., OTC Paper 13057. Plaxis BV. 2008. PLAXIS 3D Foundation Version 2.2. Poulos, H.G. & Davis, E.H. 1980. “Pile Foundation Analysis and Design”, John Wiley and Sons, New York, Series in Geotechnical Engineering, Chapter 7. Randolph, M.F., O’Neill, M.P., Stewart, D.P. & Erbrich, C. 1998. “Performance of Suction Anchors in Fine-grained Calcareous Soils”, in 30th Annual Offshore Technology Conference, 4–7 May 1998, Houston, Texas, U.S.A.: Proceedings, Vol. 1, OTC Paper 8831, pp. 521–529. Supachawarote, C., Randolph, M.F. & Gourvenec, S. 2004. “Inclined Pull-out Capacity of Suction Caissons”, ISOPE, Toulon, France, May 23–28, 2004, International Society of Offshore and Polar Engineers (ISOPE), Cupertino, pp. 500–506.

REFERENCES Andersen, K.H., Murff, J.D., Randolph, M.F, Clukey, E.C., Erbrich, C.T., Jostad, H.P., Jansen, B.,Aubeny, C., Sharma, P. & Supachawarote, C., (2005) “Suction Anchors for

© 2011 by Taylor & Francis Group, LLC

758

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Numerical investigation of the behaviour of suction caissons in structured clays S. Panayides & M. Rouainia Newcastle University, Newcastle Upon Tyne, UK

A. Osman School of Engineering, Durham University, Science Laboratory, Durham, UK

ABSTRACT: This paper investigates the ultimate capacity of suction anchor foundations, using an advanced constitutive model. The paper explores the effect of changing the aspect ratio of the caisson on the undrained load-carrying capacity of the bucket foundations in clay. Illustrative numerical results for an inorganic clay of low sensitivity from the Norrköping region in southern Sweden demonstrate the potential of the constitutive model.

1

INTRODUCTION

The relative inefficiency of piles in resisting lateral forces has led the offshore industry to consider alternative anchorage systems such as suction caissons. Suction caissons can be installed very quickly and precisely at the desired location with less heavy installation equipment and at lower cost. Therefore they are considered as a viable anchorage system in a wide variety of soils ranging from soft clay to dense sands and overconsolidated clays and for a wide variety of structures ranging from floating exploration platforms to permanent production facilities. The development of suction caissons in recent years has seen them used around the world in more than 36 fields in the last decade alone (Andersen et al., 2002). Suction caissons are large cylindrical shells, with an open bottom and a closed top fitted with valves. The aspect ratio of these piles, defined as the length to diameter ratio, is relatively small when compared with the aspect ratio of conventional piles, typically six or less (Andersen et al., 2005). Internal stiffeners are usually added, to resist buckling during the installation process, since the caisson walls are relatively thin. They are installed partly by self weight and partly by differential pressure between the surrounding environment and the inside of the skirted foundation. In some cases, dead weights can be applied on the top of the cap to ensure that compressive loads are acting on the suction anchors (Zdravkovic et al., 2001). Once full penetration has been achieved, the valve is closed. Any vertical movement during service will result in the generation of suction pressure inside the anchor which will mobilize the reverse end-bearing mechanism, as it is described by Byrne and Finn (1972). Foundations for offshore structures, however, experience significant environmental loads from waves, currents and wind giving

© 2011 by Taylor & Francis Group, LLC

rise to lateral loads. The direct consequence of that is that resultant loads can be inclined to the vertical. It is well documented that suction anchors are capable of resisting both lateral and axial loads as well as inclined loads. The ultimate capacity of suction caissons has been the focus of many investigations in recent years. Following the work of Hogervorst (1980), Keaveny et al. (1994) showed that lowering the load attachment point at mid depth increased the capacity significantly. Suction caisson capacity studies based on upper bound limit analyses from Randolph et al. (1998) and a finite element study from Sukumaran et al. (1999) indicate that the anchor capacity can be maximised when the load is located at a point which forces the anchor to fail in a translational mode of failure rather than rotational. Murff and Hamilton (1993) presented a three dimensional quasi upper bound formulation for predicting the ultimate capacity of laterally loaded piles. The three dimensional mechanism which they proposed comprised of a conical wedge near the free surface and a flow around zone below the wedge (Randolph and Houlsby, 1984). This paper presents a study of the shot-term pullout capacity of suction caissons in soft clay using an advanced constitutive model. The failure envelopes were produced for two reference caissons with length to diameter ratios of 1.5 and 3 respectively. The soil was modelled using the Kinematic Hardening Soil Model (KHSM) as proposed by Rouainia & Muir Wood (2000). 2

GEOMETRY

The geometry for the first reference suction anchor foundation adopted for this study is provided in

759

Figure 2. Finite element model for the reference suction caisson with an aspect ratio of L/D = 1.5. Figure 1. Schematic representation of the suction caisson model (Supachawarote, 2004).

estimated using the strength reduction factor and the soil properties as follows:

Figure 1. It comprises of a cylindrical suction anchor with closed top with a diameter D of the cylinder of 5 m, while the skirt length L is varied between 7.5 m and 15 m, giving rise to an aspect ratio of L/D = 1.5 and 3 respectively. A wall thickness of 50 mm (or D/100) was used in all cases. The caisson is embedded with the top cap flush with the surrounding ground level and the load attachment point, or padeye location, at a depth zp along the caisson shaft. Loads are applied at an angle θ from the horizontal, and the depth to the point of intersection of the line of action of the load with the centre-line of the caisson is denoted by zcl . The caisson is considered to be very stiff compared to the soil. The pullout loads are applied on different points on the side of the caisson with at an inclination θ to the horizontal to produce the failure envelopes. The geometry is extended 3 times the length L around to avoid influence of the geometry boundaries.

3

where ϕi
and ci
are the interface effective friction angle the interface effective cohesion, respectively and Eoed is the constrained modulus of the soil. ϕ
and c
are the friction angle and effective cohesion of the soil. Since pore water pressures or the installation of the suction anchor, is not considered the phreatic level was placed at the bottom of the geometry. In this study, the Kinematic Hardening Soil Model, developed by Rouainia & Muir Wood (2000) is use. It should be noted that this constitutive model admits the possibility, with certain combinations of soil parameters, of creating rapidly strain-softening materials. For this reason, displacement controlled analyses were carried out in Plaxis 2D software, in order to capture the strain softening behaviour of the clay as it can be seen in Figure (4).

FINITE ELEMENT ANALYSES

The cylindrical suction anchor was modelled using the Plaxis 3D with 15-noded wedge elements. The anchor was modelled by linear elastic wall elements with a high stiffness making them virtually rigid. Since the governing failure mechanisms do not involve the soil plug inside the anchor, this soil was modelled as a stiff, elastic material. The mesh used for the short caisson can be seen in Figure 2. Approximately ∼13000 elements and ∼26000 nodes was found to be sufficiently refined in order to minimize the discetization error for the first case, whereas for the L/D = 3 a mesh of ∼16500 elements and ∼45000 nodes was used. For all the FE-models in this study, interface elements along the outside skirt walls have been used. with the strength reduction factor of the interface (Rinter) set to 0.65. The interface properties are

© 2011 by Taylor & Francis Group, LLC

4

MATERIAL MODEL

The model used in this study was formulated for natural clays within the framework of kinematic hardening with some elements of bounding surface plasticity. It is a rate independent model and it takes into account the effects of damage to structure caused by irrecoverable plastic strains, resulting from sampling or geotechnical loading. KHSM is an extension of the well known Cam-Clay model. The steady fall of stiffness with strain is controlled by an interpolation function which ensures a smooth advancement of the elastic domain (which is enclosed in a small bubble) towards the bounding surface during loading. A scalar variable r, which is a monotonically decreasing function of the plastic strain, represents the progressive degradation of the material. Accordingly,

760

Table 1.

Soil parameters for Norrköping Clay.

Material constants

Value

Slope of swelling line κ∗ Slope of normal compression line λ∗ Poisson’s ratio v Critical state stress ratio M Ratio of size of bubble and reference surface R Stiffness interpolation parameter B Stiffness interpolation parameter ψ Destructuration parameter k Destructuration strain parameter A Initial degree of structure r0 Anisotropy of initial structure η0

0.0297 0.252 0.22 1.35 0.145 1.98 1.547 4.16 0.494 1.75 0.5

the following exponential destructuration law, is proposed

where ro denotes the initial structure and kis a parameter which describes the rate of destructuration process with strain. The rate of the destructuration strain εd will be assumed to have the following form

where A is a non-dimensional scaling parameter and p p εq and ε˙ v are the plastic shear strain and the plastic volumetric strain, respectively. The governing constitutive relations of the KHSM are summarized in the Appendix. 5

CLAY PROPERTIES

The parameters required for the analysis correspond to inorganic clay of low sensitivity from the Norrköping region in southern Sweden. The soil has undrained shear strength of 10 kPa down to 3 m, increasing linearly with depth below that level at a rate of 2 kPa/m. An effective unit weight γ
= 10 kN/m3 was used. The model parameters for the soil and the interface were taken from Westerberg (1995). An over consolidation ratio (OCR) of 1 was adopted for the analyses. The coefficient of lateral earth pressure (K0NC = 1-sin ϕ
) was taken as 0.5 which corresponds to an effective friction angle φ’ of 300 . For the soil-structure interface, an oedometric Young’s modulus (Eoed ) of 1800 kPa and cohesion c
of 2.1 kPa were used. These optimized parameters are described as reference parameters and correspond to the KHSM model in all the analyses below. 6 6.1

RESULTS AND DISCUSSION Failure mechanisms

Figures 3(a) and 3(b) show the 3D and 2D deformed meshes from the analyses of horizontally loaded

Figure 3. Deformed mesh from analyses of pure horizontal loading, for caisson L/D = 1.5 (a) 3D model and (b) 2D model.

suction caissons with an aspect ratio of L/D = 1.5. It can be seen that clear failure surfaces form on the active side of the caisson. On the passive side, the well defined failure surfaces form very close to the caisson, while the caisson translates horizontally. The failure mechanisms become more distinctive with the plot of the displacement vectors as it will be discussed in the following section (Figure 6). As it is evident in Figure 4, the KHSM model can replicate the strain softening behaviour of the clay. The KHSM predicts a peak load of 583 KN, which is ∼9% higher than the ultimate load predicted by the bubble model for which the structure behaviour is switched off. This behaviour would correspond to reconstituted soils.The increased ultimate capacity can prove critical when designing caissons for offshore structures, as it will enable engineers to utilize more economic designs in the future. As a direct consequence, the cost of offshore founding systems may be reduced significantly. It should be noted that this difference in the behaviour is also observed when the caissons are analysed using the three-dimensional finite element model. Figure 5 depicts the displacement shadings in 3D for the case where purely vertical load was applied on the suction caisson with an aspect ratio of L/D = 1.5. It can be seen that the model predicts the resistance to uplift which corresponds to full mobilisation of a reverse bearing mechanism. Figures 6 depict the displacement vectors and failure mechanism obtained from the analysis corresponding to pure horizontal loading where the load was attached at the optimal point. It can be seen that a

761 © 2011 by Taylor & Francis Group, LLC

Figure 6. Displacement vector for horizontal ly loaded suction caissons L/D = 3.

Figure 4. Load-displacement curve in 2D for horizontally loaded suction caissons L/D = 1.5.

Figure 5. Failure kinematics for vertically loaded 3D suction caissons L/D = 1.5.

well defined failure surface develops on both sides of the caisson. This is in very good agreement with the failure kinematics that have been proposed in the literature for lateral loading of caissons. It can also be seen that the failure extends to the bottom of the caisson, with no flow around zone occurring for the caisson. It should be noted however that the gradient of the wedge varies with depth. As the wedge approaches the tip of the caissons it tends to curve passing tangentially at the bottom of the caisson. 6.2

Failure envelope comparison

Figure 7(a) shows the comparison of the failure envelope for non-horizontal loadings for the structured model (KHSM), the Bubble model and two analytical methods suggested by Supachawarote (2005) and Senders & Kay (2002). It is evident that the structured model predicts the ultimate load for all loading angles very well. In contrast, the bubble model, consistently, underestimates the ultimate load by an average of ∼12%. This behaviour is as expected, since the bubble model was formulated to represent the behaviour of reconstituted material and cannot account for added strength the natural clay deposits exhibits.

Figure 7. Failure envelopes for suctions caissons of different aspect ratios: (a) L/D = 1.5 and (b) L/D = 3.

The analytical equation proposed by Supachawarote et al (2004) for the failure envelope has the following form:

762 © 2011 by Taylor & Francis Group, LLC

where: a = 0.5 + L/D; b = 4.5 − L/3D

into account, this results in an increase in ultimate capacity of suction caissons, which in turn can lead to a more economical design.

The expression is similar to the one proposed by Senders and Kay (2002), who suggested that both coefficients a and b are equal to 3 for all aspect ratios. Figure 7(b) provides an insight on the effects an increasing of the embedment length of a caisson, has on the capacity. It is evident from the two failure envelopes that the capacity increases significantly with increasing aspect ratio. Moreover, the longer caisson exhibit maximum capacity for loads that have higher lateral components, which is in contrast with the shorter caisson, where the maximum capacity is observed under loads with high vertical components. The capacity for pure horizontal loading is four times larger when the aspect ratio is increased from 1.5 to 3, whereas the maximum load corresponding to pure vertical loading, increases by a factor of ∼2.5. As it can be seen the finite element results agree very well with the curve fitting equation proposed by Supachawarote (2004). The equation recommended by Senders and Kay (2002), predicts lower capacities for inclined loading. For load angles that are close to the pure vertical or pure horizontal loadings, the results from the analytical solution are very similar to the finite element capacities. As the ellipsoidal equation from Supachawarote is made to depend on the caisson geometry, the predicted results are deemed to be more reliable for all caisson geometries.

7

8 APPENDIX I. Non-linear elastic constitutive law

where κ∗ is the swelling line in a volumetric strainlogarithmic mean compression plane and K and G are the bulk and shear modulus, respectively. The stress and strain tensors can be expressed in terms of the volumetric and the deviatoric components and the basic elasto-plastic assumption is the additive decomposition of the strain rate, ε, into an elastic and a plastic part. II. Yield functions

SUMMARY AND CONCLUSIONS

This paper presented numerical analyses of the undrained ultimate capacity of suction caissons installed in a soft structured clay deposit. Two finite element models were simulated for caissons with aspect ratios L/D set to 1.5 and 3 in order to compute the ultimate capacity, for a variation of load inclinations in order to provide the full failure envelopes. The clay deposit was modelled with an advanced constitutive model for natural clays, namely the Kinematic Hardening Structure Model (KHSM), which can simulate the destructuration process. The 2D study included displacement controlled analysis in order to capture the strain softening nature of the soil. Plaxis 3D was used to produce the full vertical and horizontal interaction diagram for the two cases. The two failure envelopes were compared with two analytical equation as proposed by Senders and Kay (2002) and Supachawarote et al (2004). The shape of the yield envelopes obtained with 3D calculations is in good agreement with the analytical solutions and may be used in practice to obtain the optimal capacities on suction caissons. Furthermore, the kinematics obtained from the FEM in both 2D and 3D simulations for various loading inclinations agree very well with the failure mechanisms proposed in analytical methods. The above conclusions imply that the KHSM model can predict the undrained ultimate capacity of suction caissons accurately. When the soil structure is taken

© 2011 by Taylor & Francis Group, LLC

where M is the critical state stress ratio and pc is a scalar variable which defines the size of the outer surface. a¯ denotes the location of the centre of the bubble and R represents the ratio of the sizes of the elastic bubble and the outer surface. η0 is a dimensionless deviatoric tensor controlling the structure and r is the ratio of the sizes of the structure surface and the reference surface. Note that the position of the centre of the structure surface is given by aˆ = {rpc , (r − 1)η0 pc }. III. Isotropic hardening law

λ∗ is the slope of the normal compression line expressed in the same plane as κ∗ . IV. Kinematic hardening law

763

where µ ˙ is a positive scalar of proportionality and σ c is there the conjugate stress tensor. V. Plastic modulus at current stress

The additional material parameters ψ and B control the rate of decay of stiffness with strain and the magnitude of the contribution of the interpolation term, respectively. For more information regarding the formulation of the constitutive model, the reader is referred to Rouainia &Muir Wood (2000).

REFERENCES Andersen, K.H. and Jostad, H.P, (2002). “Shear strength along outside wall of suction anchors in clay after installation,” Proc. 12th International Offshore and Polar Engineering Conference, pp 785–794. Andersen,K.H., Murff J.D., Randolph M.F.,Clukey E.C., Erbrich C., Jostad H.P., Hansen B., Aubeny C., Sharma P., and Supachawarote C. 2005. Suction anchors for deepwater applications. Int. Symp. on Frontiers in Offshore Geotechnics, ISFOG. Sept. 2005. Perth, Western Australia. Proc. A.A. Balkema Publishers. Andersen L.,Edgers L.and Jostal H.P.(2008) Capacity Analysis of Suction Anchors in Clay by Plaxis 3D Foundation. Plaxis Bulletin, Issue 24. Oct. 2008. Aubeny, C., S. Moon, and J. Murff (2001). Lateral undrained resistance of suction caison anchors. International Journal of Offshore and Polar Engineering 11(2), 95–103. Aubeny, C., J. Murff, and J. Roesset (2001).Geotechnical issues in deep and ultra deep waters. International Journal of Geomechanics 1(2), 225–247. Aubeny, C.P., Moon, S.K., and Murff, J.D. (2001b). “Lateral undrained resistance of suction caisson anchors.” International Journal of Offshore and Polar Engineering., 11(2), 95–103. Aubeny, C.P., Han, S.W., and Murff, J.D. (2003a). “Inclined load capacity of suction caissons.” International Journal for Numerical and Analytical Methods in Geomechanics., 27(14), 1235–1254. Aubeny, C.P., Han, S.W., and Murff, J.D. (2003b). “Refined model for inclined load capacity of suction caissons.” 22nd International Conference on Offshore Mechanics and Arctic Engineering, Cancun. Bransby, M.F., and Randolph, M.F. (1998). “Combined loading of skirted foundations.” Geotechnique., 48(5), 637–655.

© 2011 by Taylor & Francis Group, LLC

764

Hogervorst, J.R. (1980). Field trials with large diameter suction piles. Offshore Technology Conference,Houston, Texas. Paper 3817. Keaveny, J.M., Hansen, S.B., Madshus, C. and Dyvik, R. (1994). Horizontal capacity of large-scale model anchors. Proceedings of the 8th International Conference on Soil Mechanics and Foundation Engineering, New Delhi, India. Murff, J.D., and Hamilton, J.M. (1993). “P-Ultimate for undrained analysis of laterally loaded piles.” ASCE, Journal of Geotechnical Engineering., 119(1), 91–10. Randolph, M.F., and Houlsby, G.T. (1984). “The limiting pressure on a circular pile loaded laterally in cohesive soil.” Geotechnique., 34(4) 613–623. Randolph, M.F., O’Neill, M.P., Stewart, D.P., and Erbrich, E. (1998). “Performance of suction anchors in fine-grained calcareous soils.” Proc. Offshore Tech. Conf., Houston, OTC 14236. Randolph, M.F., & House A.R. (2001). “Analysis of Suction Caisson Capacity in Clay” Proc. Offshore Tech. Conf., Paper No.8831, 521–52. Rouainia, M. and D. M.Wood (2000). A kinematic hardening constitutive model for natural clays with loss of structure. Geotechnique 50(2), 153–164. Senders, M. and Kay, S. (2002) “Geotechnical Suction Pile Anchor Design in Deep Water Soft Clays”, Conference Deepwater Risers, Moorings and Anchorings, London, UK. Sukumaran, B., and McCarron, W.O. (1999). “Total and effective stress analysis of suction caisson for Gulf of Mexico conditions.” OTRC 99 International Conference on Analysis, Design, Construction, and Testing of Deep Foundations, Austin, TX, 247–260. Supachawarote, C., Randoplh, M.F and Gourvenec, S. (2004). “Inclined Pull-Out Capacity of Suction Caissons” Proceedings of the 14th International Society of Offshore and Polar Engineering Conference, 500–506. Zdravkovic, L., Potts, D.M. and Jardine, R.J. (1998). “PullOut Capacity of Bucket Foundations in Soft Clay”. Proceedings of the International Conference on Offshore Site Investigation and Foundation Behaviour, 301–324. Zdravkovic, L., Potts, D.M. and Jardine, R.J. (2001). “Parametric Study of the Pull-Out Capacity of Bucket Foundations in Soft Clay”, Geotechnique, 51 (1) 55–67.

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Cyclic moment loading of suction caissons in sand B. Zhu Department of Civil Engineering, Zhejiang University, China

B.W. Byrne & G.T. Houlsby Department of Engineering Science, University of Oxford, UK

ABSTRACT: Suction caissons are being considered as a foundation option for offshore wind turbines. One configuration involves use of the caisson as a mono-foundation, in which case the moment-rotation response of the caisson must be well understood by the designers. The monotonic response has been the focus of recent research work and there is some guidance now available for engineers to use. However, less focus has been applied to the cyclic response and particularly to the long term response under cyclic loading. Over the lifetime of the turbine the foundation is likely to be exposed to many millions of cycles, of varying amplitude and period. One of the key issues that must be addressed by a designer is whether the accumulated rotation, over the life of the turbine, is acceptable. This paper presents experimental data from long term cyclic loading tests on caisson foundations. These tests can be used to develop a framework of response for long term cyclic loading. This paper describes the small scale cyclic loading rig used at Oxford and presents some typical results that give insight into the caisson response to a range of cyclic moment loading patterns. 1

INTRODUCTION

Offshore wind turbines can be founded on either a multi-footing structure, usually when the water depth is greater than 20 m, or a single footing structure, typically for water depths less than 20 m. Only a few multi-footing structures have been installed (for example the two Beatrice wind turbines north of Scotland) so most installed wind turbines are located on monostructures. The monopile is the usual design, although gravity base foundations have also been adopted by some developers. A minor variation on the gravity base concept is the use of a skirted foundation such as a suction installed caisson, as shown schematically in Figure 1(a). Figure 1(b) shows a suction installed foundation at Frederikshaven in Denmark. This was installed in 2002 and supports a V90 wind turbine, though the site is onshore (in a shallow lagoon) rather than truly offshore. More recently a suction caisson foundation was installed offshore at Horns Rev 2 Offshore Wind Farm (also in Denmark) as a foundation for a meteorology mast (LeBlanc, 2009). The suction installed foundation might be a preferred option to a monopile, not because they may use a lower amount of steel, but because the concept allows the installation contractor to use lighter duty installation processes. In fact it is possible, as appears to be the case for the meteorology mast foundation, to develop a floating, self-installing structure that does not require a heavy lift vessel. This option therefore has potential not only around the UK coast but also at other locations such as China, where there is an increasing demand for electricity from renewable energy sources.

© 2011 by Taylor & Francis Group, LLC

765

Figure 1. Suction caisson foundation for wind turbine.

For the monopod configuration the designer of the suction caisson must understand accurately the moment-rotation response. The monotonic response is important for developing ultimate capacity calculations and for developing appropriate stiffness parameters for structural dynamics calculations. The cyclic loading response is important for understanding accumulated deformations and changes in stiffness with

time. For design the accumulated angular rotation must be limited in the lifetime of the wind turbine. For the gravity base foundation at Thornton Bank the design rotation under cyclic loading was limited to 0.25◦ (Peire et al. 2009). In China the design codes require the cyclic accumulated angular rotation of the foundation to be less than 0.17◦ for onshore wind turbines, and there is no corresponding criterion for offshore turbines yet. Any changes in stiffness arising from the cyclic loading must also be well understood as there could be significant effects on the dynamic response of the turbine. Wind turbines are normally designed to have natural frequencies in the range between the frequency bands of the rotor rotation and the blade passing frequency, usually denoted by 1P and 3P. This avoids resonances, but changes in stiffness might result in interference between the first natural frequency and the excitation frequencies, 1P or 3P of the wind turbine. Research on suction caisson foundations has recently been carried out at the University of Oxford, Aalborg University and the University of Western Australia (Senders, 2008). This has concentrated on the monotonic response (Feld 2001, Houlsby et al. 2005, Villalobos 2006) and with a small amount of work on the cyclic response (though for only small numbers of cycles). There has been very little work looking at the long term response of caissons to cyclic loading. This paper addresses this gap by presenting results from cyclic moment loading experiments on suction caisson foundations. A novel cyclic loading rig, located at the University of Oxford, is described and some typical results from the tests are given. These results can be used to develop a framework to predict the longterm cyclic accumulated angular rotation of the suction caisson foundation.

2 TEST EQUIPMENT AND TEST PROGRAM 2.1

Cyclic loading rig and testing procedure

The loading rig is shown in Figure 2. A simple, motor driven, loading system is used to apply cyclic loading to the caisson. The rig consists of a soil container (0.55 m by 0.6 m by 0.6 m), a steel frame with pulleys, three weight-hangers and a lever with a driving motor. The lever is attached to the steel frame through a pivot and carries a motor, which rotates a mass m1 to provide cyclic loading on the foundation. The weights can be adjusted to impose different loading regimes. The motor is a geared single-phase AC motor rotating at a frequency of 0.106 Hz. The rig is very stable and can accurately provide a sinusoidal loading for a large number of cycles. The rig was originally developed by Rovere (2005) but more recently described and used by LeBlanc (2009) and LeBlanc et al. (2010) to carry out a series of cyclic loading tests on stiff piles (representing monopiles). The maximum number of cycles applied during LeBlanc’s work was 65,370.

© 2011 by Taylor & Francis Group, LLC

Figure 2. Long-term cyclic loading rig. Table 1.

Redhill 110 properties (Villalobos, 2006).

Parameters

Values

D10 , D30 , D50 , D60 (mm)

0.08, 0.10, 0.12, 0.13 1.63, 0.96

Coefficients of uniformity, Cu and curvature Cc Specific gravity, Gs Minimum dry density, γmin (kN/m3 ) Maximum dry density, γmax (kN/m3 ) Critical state friction angle, φcs

2.65 12.76 16.80 36◦

For the tests described in this paper a load cell was used to monitor the overturning load on the caisson. Three LVDTs were also installed on the caisson to measure the rotation and deflection of the caisson. The diameter (D), the length of skirt (L) and the thickness (t) of the skirt wall of the caisson used was 0.2 m, 0.1 m and 0.001 m, respectively. The dimensions of the caisson were chosen to represent 1:75 model of a proposed suction caisson mono-foundation for an offshore wind turbine. A fine, silty sand was used for the experiments, as this is one of the more usual soil types at the sites of potential offshore wind farms in China. The sand was Redhill 110, a commercially produced sand, with properties given in Table 1. The sand bed and equipment were set up in a repeatable way to ensure similar conditions for each test. The sample-box was carefully filled with sand by pouring the dry sand from a very low drop height to achieve a loose state. The relative density of the fine sand was controlled to approximately 11%. The caisson was installed by a pushed installation rather than by suction, as the soil used was dry and not saturated. Previous work by Villalobos (2006) has provided guidance on the effects of suction installation, as compared to pushed installation, on the caisson response under loading. To ensure a vertical penetration of the caisson, in the centre of the soil container, a vertical rod within a guide was used. The load for the pushing installation was applied by

766

Figure 3. Determination of ultimate moment capacity.

Figure 4. Cyclic loading ranges in relation to monotonic moment capacity of caisson.

adding weights to a loading hanger located centrally on the caisson by a point contact. The maximum vertical load applied to the caisson was 113 N which corresponds to a non-dimensional vertical load of V˜ pre = Vpre /(γ
D3 ) = 1.07. Once installed the loading wires were connected to the vertical rod on the caisson. The vertical installation rig was carefully dismantled and the appropriate weight, m1 , was added to the hanger connected to the motor in order to apply the right amount of cyclic loading. The load cell and LVDTs enabled load and displacement data to be electronically recorded. The instruments were powered by a RDP M600 system which also allowed the resulting signals to be amplified. The electronic data was collected by a National Instruments USB-6009 digital card. The data collecting frequency was set at 2 Hz and the data were collected continuously during each cyclic test. Processing of the signals, by means of a digital filter, was also carried out to remove unwanted electrical noise. 2.2 Test program The test program consisted of both cyclic and monotonic testing of the caisson. The monotonic tests were carried out at different ratios of M /(HD) to determine the ultimate moment capacity of the caisson, MR . A typical moment-rotation curve is shown in Figure 3. To define the point of yield (or capacity) the method described by Villalobos (2006) was used. Straight lines are fitted to the initial elastic section and the later plastic section and the intersection is used to define the yield load. This exercise is repeated for different values of M /(HD) to define a yield envelope. Figure 4 shows the results of five tests as well as the line of best fit plotted through the experimental data points to represent the yield envelope, which is approximately straight in this region. Also shown on Figure 4 is a line of constant M /(HD) extending from the origin. This line is chosen to reflect the field case where typically the diameter may range from 15 m to 25 m, and the height of the resultant lateral load (both wind and wave) might be about 30 m (Byrne and Houlsby, 2003) which gives

© 2011 by Taylor & Francis Group, LLC

Figure 5. Definition of cyclic parameters (from LeBlanc et al., 2010).

an eccentricity e = M /(HD) between 1.2 and 2.0. Along this line are plotted four important design loads for the wind turbine. These are: (i) the ultimate loadcarrying capacity related to the limit state ULS; (ii) the worst expected transient load related to ULS/1.35; (iii) the serviceability state SLS which occurs 102 times during the lifetime of the wind turbine; and (iv) the fatigue limit state FLS which occurs 107 times during the lifetime of the wind turbine (DNV, 2007). Devising the cyclic test program requires considerable care to limit the number of tests carried out. There is also a need for a consistent approach to defining the applied loading. LeBlanc et al. (2010) defined two parameters to characterize the applied cyclic load:

767

Table 2. tests.

Normalized parameters of caisson in 1 g model

Parameters

Normalized expressions

Moment M Vertical load V Angular rotation θ Unloading stiffness in each cycle k Vertical displacement w

˜ = M /(γ
D4 ) M ˜ V = V /(γ
D3 ) θ˜ = θ(pa /γ
D)0.5 k˜ = k/[(γ
pa )0.5 D3.5 ] w˜ = (w/D)(pa /γ
D)0.5

Figure 7. Angular rotation calculated by displacements.

Figure 6. Displacements measured by LVDTs.

Here Mmax and Mmin are the maximum and minimum moment in the load cycle as shown in Figure 5, and MR refers to the static capacity at the selected ratio of M /(HD). As illustrated in Figure 5 the dimensionless parameter ζc denotes the type of loading ranging from 0 for one-way loading to −1 for two-way loading. A test program was therefore devised that considered a range of these parameters to explore the accumulated deformation behavior of the caisson. By carrying out a dimensional analysis it is possible to develop a number of normalized groups with which to present the results and these are shown in Table 2. 3 3.1

CYCLIC MOMENT LOADING TEST RESULTS Cyclic accumulated deformation

Some typical test results are shown in Figures 6 through to 9. Figure 6 shows the raw displacement data as collected with time. By accounting for the geometry of the set-up it is possible to deduce the angular rotation and the settlement of the caisson and this is shown in Figures 7 and 8 respectively. Together with the data of continuous cyclic loads, the plot of the normalized angular rotation in response to the applied moment is given in Figure 9. It can be observed that the first loading cycle generates a larger displacement than the following ones, and the accumulated cyclic angular

© 2011 by Taylor & Francis Group, LLC

Figure 8. Angular rotation calculated by displacements.

rotation increases with the number of cycles but that the angular rotation specific to one cycle decreases with the number of cycles. In a similar way to a piled foundation (Lin & Liao 1990, LeBlanc 2009) the cyclic accumulated angular rotation of the caisson can be evaluated in terms of the dimensionless ratio (θ N − θ0 )/θ s . Here θ N and θ 0 are illustrated in Figure 5 and θ s is the angular rotation that would be experienced in a monotonic test if the maximum load under cycling were applied. The relationship between (θ N − θ 0 )/θ s and the number of cycles for different characteristics of the cyclic load is shown in Figure 10. It would appear that the dimensional ratio (θ N − θ 0 )/θ s is almost linear with the number of cycles in logarithmic coordinates within the 10,000 cycles applied. The normalized accumulated settlements of the caisson for different ζ b are shown in Figure 11. The values increase with the increasing cycles and amplitude of the cyclic loads. Settlement of the caisson foundation is of course also an important consideration for this application.

768

Figure 11. Cyclic accumulated settlement against cycle number.

Figure 9. Relationship between moment and angular rotation.

Figure 12. Instantaneous centre for monotonic and cyclic tests.

position of the instantaneous centre of rotation. For a common eccentricity e = M /(HD) = 1.875, the movements of the instantaneous centre for monotonic and cyclic tests are shown in Figure 12. For the monotonic test, the centre moves from beneath the skirt to the horizontal plane at the skirt tip at the end of the test. For the cyclic loading test the instantaneous centre starts at a much greater depth below the caisson and does not change much during the test.

3.3

Figure 10. Normalized accumulated angular rotation against cycle number.

3.2 Rotation of the caisson During both monotonic and cyclic moment tests it is possible, given the displacements x1 , y1 and y2 measured by three LVDTs (see Figure 1), to deduce the © 2011 by Taylor & Francis Group, LLC

769

Unloading stiffness

If the foundation stiffness changes during the lifetime of the wind turbine then there might be a concern that dynamic characteristics of the structure will be altered. In particular the 1st natural frequency of the structure might start to move towards one of the excitation frequencies. Figure 13 presents curves of normalized unloading stiffness for three typical cyclic tests. It can be observed that the unloading stiffness of the caisson is constant irrespective of the accumulated angular rotation, for the number of cycles explored here. For these tests the stiffness appears to decrease with increasing maximum cyclic load.

ACKNOWLEDGEMENT This work was supported by National High-Tech R&D Program of China (863 Program) (research grant: 2007AA05Z427) and National Natural Science Foundation of China (research grant: 50979097). The first Author acknowledges the support from the Department of Engineering Science at Oxford. REFERENCES

Figure 13. Unloading stiffness in relation to angular rotation.

4

CONCLUSIONS

Suction caisson foundations might be used as monopod foundations for offshore wind turbines. One of the key issues will be the accumulated rotation over the lifetime of the turbine and whether this can be predicted. This paper describes a series of tests aimed at understanding the long term cyclic loading response of suction caissons. A unique loading rig for applying cyclic loading to foundations was described. A set of experimental results were presented and the test results were normalized. A framework for understanding the cyclic accumulated angular rotation and unloading stiffness of the caisson was introduced. As would be expected the accumulated angular rotation of the caisson is significantly affected by the characteristics of the applied cyclic load. The dimensional ratio of accumulated rotation, (θ N − θ0 )/θ s , was found to be almost linear with the logarithm of the number of cycles within the limits of 10,000 cycles applied. Further work is required to explore whether this relationship can be used to predict the cyclic accumulated angular rotation of the caisson during an ocean storm or the lifetime of the wind turbine. This will require tests to larger numbers of cycles. In addition, centrifuge and full scale tests as well as tests on a higher relative density sand would be required to provide added confidence in the scalability of these results.

Byrne, B. W. & Houlsby, G. T. 2003. Foundations for offshore wind turbines. Phil. Trans. R. Soc. Lond. A, 361: 2909– 2930. LeBlanc, C. 2009. Design of Offshore Wind Turbine Support Structures. PhD Thesis, Aalborg University, Denmark. LeBlanc, C., Houlsby, G.T. and Byrne, B.W. 2010. Response of stiff piles in sand to long term cyclic loading. Geotechnique, 60(2): 79–90. DNV. 2007. Offshore Standard (DNV-OS-J101): Design of Offshore Wind Turbine Structures. Det Norske Veritas, Hovek, Norway. Feld, T. 2001. Suction Buckets, a New Innovative Foundation Concept, applied to Offshore Wind Turbines. PhD Thesis, Aalborg University, Denmark. Houlsby, G. T. Ibsen L. B. & Byrne B. W. 2005. Suction caissons for wind turbines. In Gourvenec & Cassidy (eds.), Frontiers in Offshore Geotechnics: ISFOG, Perth, Australia, 19–21 September 2005. London: Taylor & Francis. Lin, S. S. & Liao, J. C. 1990. Permanent strains of piles in sand due to cyclic lateral loads. Journal of Geotechnical and Geoenvironmental Engineering, ASCE. 125(9): 798–802. Kelly, R. B., Houlsby, G. T. & Byrne, B. W. 2006. A comparison of field and laboratory tests of caisson foundations in sand and clay. Geotechnique 56(9): 617–626. Peire, K. Nonneman, H. & Bosschem, E. 2009. Gravity based foundations for the Thornton Bank Offshore Wind Farm. Terra et Aqua, 115: 19–29. Rovere, M. 2005. Cyclic loading test machine for suction caisson foundations. Project Report. Politecnico di Milano. Senders, M. 2008. Suction caissons in sand as tripod foundations for offshore wind turbines. PhD Thesis, The University of Western Australia, Australia. Villalobos, F. A. 2006. Model Testing of Foundations for Offshore Wind Turbines. DPhil Thesis, University of Oxford, UK.

770 © 2011 by Taylor & Francis Group, LLC

10 Pipelines and risers

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Multidirectional analysis of pipeline-soil interaction in clay R.G. Borges PETROBRAS Research & Development Centre, Rio de Janeiro, Brazil

J.R.M.S. Oliveira Military Institute of Engineering, Rio de Janeiro, Brazil

ABSTRACT: This paper discusses the interaction, in all directions, between buried pipes and the surrounding soil. In numerical terms, this interaction is normally modelled via separate elasto-plastic springs positioned in the main directions (vertical and horizontal). Many experimental results are available in the literature dealing with soil-pipe interaction in the main directions on a plain submarine slope. Equations are accessible to calculate the maximum soil resistance for design purposes. However, the direction of movement of the pipeline is not taken into account and, depending on the soil characteristics, it may lead to a sub-estimation of the confinement forces. A comprehensive set of numerical analyses was carried out to study the soil response in all directions. The software SIGMA/AEEPECD was used in these Finite Element analyses. A reference curve for soil response in any directions was achieved. The numerical results were compared with a standard for pipe-soil interaction, showing good agreement. 1

INTRODUCTION

seabed, which can induce overloads that can cause the pipeline to rupture.

Offshore pipelines in shallow waters are generally buried to provide mechanical protection and constraint. A safe buried pipeline design must take into account a reliable evaluation of the pipe-soil interaction forces and the associated displacements, in order to assure structural integrity during operation. Simplified usual design practice (ALA, 2005) based on Hansen (1961), assumes the pipe as a beam and the soil restraint as axial and transversal Winkler springs placed along its external body. The properties of these springs vary with direction and position along the pipeline, showing no relationship with each other. Many authors have proposed theoretical, numerical and experimental approaches to investigate the soil-pipe interaction. Among those who have chosen the numerical simulation, Fernando & Carter (1998), Popescu & Konut (2001), Popescu et al. (2002), Guo & Popescu (2002), Oliveira et al. (2005) and Borges (2008) focused on horizontal or vertical pipe-soil interaction. Furthermore, Zhang et al. (2002) and Calvetti et al. (2004) studied failure envelopes for pipelines subjected to combined force and displacement loadings both experimentally and numerically. In that way, the main purpose of this work is to investigate the soil response envelope for any direction the pipe may take between fully downwards and upwards. Such loading happens when pipelines are installed in harsh environments, such as along an unstable slope, crossing an active fault plane, or buried in soil layers likely to suffer liquefaction upon the occurrence of an earthquake. When one of these conditions occurs, the pipeline can be affected by relative movements of the

2

Numerical analyses using Finite Element Method (FEM) were used in this research to investigate the soil-structure interaction behaviour of an offshore steel pipe buried in marine clay, considering several burial depths and a comprehensive set of displacement directions, all in monotonic conditions. 2.1 The SIGMA/AEEPECD software For the numerical simulation of the non-linear physical behaviour of the soil, a continuous medium was considered, which demanded the application of interactive incremental integration algorithms. The software SIGMA v. 5.32 (Amaral et al., 1997, LIRA, 1998), was used for the bi-dimensional Finite Element discretization and model post-processing. This system is based on the integration of numerical analysis modules developed by the Scientific Methods Team of Petrobras Research and Development Centre, with graphical tools conceived by an agreement signed between Petrobras and Tecgraf/PUC-Rio. These tools consist of programs used for geometric modelling, specification of attributes, finite element mesh generation and results visualization provided by the numerical simulators. Such tools were used as pre- and postprocessors of the models, assisting in the modelling of the Engineering problem. The version 3.02 of the numerical simulator AEEPECD (Cardoso, 2005), was used to solve the

773 © 2011 by Taylor & Francis Group, LLC

NUMERICAL MODELLING

Figure 1. Espadarte Field. Offshore Brazil. Fugro (2005).

equilibrium of continuum differential equations. This software allows linear and non-linear analyses of plane strain, plane stress and axi-symmetric models. The program uses constitutive laws for plastic and elastoplastic rheologies. The adopted plastic criteria are Von Mises and Mohr-Coulomb, incorporating geometrical and physical non-linearities. 2.2

Soil material properties

The soil strength profile was based on the undrained shear strength (su ) results from in situ and laboratory tests in a Southeastern Brazilian coast marine site. Piezocone penetration tests (PCPT) were undertaken in Espadarte marine field (Figure 1), using the Deepwater Seacalf System as well as laboratory tests in Jumbo Piston Core (JPC) samples (Costa et al., 2002). The JPC samples were obtained using a 20 m long piston launched in free fall 2.5 m above the seafloor. Afterwards, the soil undisturbed samples were conditioned in special cases and taken to the laboratory. Data from the PCPT tests were interpreted using the methodology suggested by Amaral & Costa (1998). Based on the results obtained from the set of tests in Espadarte oil field, the geotechnical borehole GT212 was chosen as a minimum strength profile, with a water depth of 963.39 m, as shown in Figure 2: Equations (1) and (2) show total and effective stress regression lines, respectively:

Figure 2. Geotechnical strength profile for borehole GT-212.

The total strength regression line was chosen to simulate the soil resistance with depth. The mean value of the submerged unit weight was adopted from the test JPC-212 (γ
= 5.45 kN/m3 ). The earth pressure coefficient k0 was assumed as 1, the Poisson ratio ν = 0.49 and the soil Young modulus (E) was obtained from Equation (3), proposed by Amaral et al. (2002):

The mobilized shear strength in the soil-pipe interface (τ u ), was calculated using Equation (4):

Where r is an interface parameter which varies from 0 (perfectly smooth surface) to 1 (perfectly rough surface). In this work, a mean value of r = 0.5 was used to simulate an intermediate situation. The non-linear behaviour of the foundation soil was represented by the Mohr-Coulomb plastic yielding model, considering undrained conditions, which means that the internal friction angle is zero. This implies that the Mohr-Coulomb plastic yielding model turns into the Tresca criterion.

774 © 2011 by Taylor & Francis Group, LLC

Table 1.

Geometrical and mechanical properties of the pipe.

Property

Value

External Diameter (D) Thickness (e) Length (L) Young Modulus (E) Poisson’s Ratio (ν) Yielding Stress (σy ) Ultimate Stress (σu )

0.60 m 0.0254 m Infinite 2.06 108 kPa 0.30 3.58 105 kPa 4.55 105 kPa Figure 4. FE spatial discretization for H/D = 50%.

Figure 3. Definition of angle α and burial depth H .

2.3 Pipe material properties

Figure 5. Displacement vector distribution for H /D = 100% and α = 45◦ .

Table 1 resumes the main geometrical and mechanical properties adopted for the weightless empty pipe:

depth. A typical mesh for this problem, along with the applied displacement boundary conditions, is show in Figure 4 (Borges, 2009): The total lateral and vertical dimensions of the models were adopted as 8 m and 4 m, respectively. These values were considered large enough to avoid any boundary influence.

2.4 Description of the simulation models Small strain analyses were performed on the preembedded pipe, considering six different burial depths (H /D = 25, 50, 75, 100, 150 and 200%), in order to obtain the soil response for a pipe subjected to constant incremental displacements in various directions, which are defined by the angle α with the vertical (Figure 3). The chosen angles were α = 0, 15, 30, 45, 60, 75, 90, 105, 120, 135, 150, 165 and 180◦ , leading to a total number of 78 processed analyses. The finite element mesh was composed by isoparametric quadrilateral quadratic elements with 8 nodes, and the pipe-soil contact elements were composed by special interface elements with 6 nodes and quadratic variation of the relative displacements. The soil-pipe contact was represented by special interface elements, assuming a non-linear constitutive law for the shear stress-strain relationship, and a Mohr-Coulomb criterion for the contact material. In these interface elements, it was considered the nonlinear behaviour of the soil in the tangential direction, assuming a maximum shear stress equal to the soil undrained shear strength. In the normal direction, it was considered a linear behaviour. The actual distribution and concentration of elements varied as a function of the pipe embedment

© 2011 by Taylor & Francis Group, LLC

3 3.1

SIMULATION RESULTS Displacement vectors

For better visualization of the deformation mechanisms occurring within the soil mass, Figures 5 and 6 show the displacement vectors for H /D = 100% and direction angles α = 45◦ and 135◦ . It was noticed the formation of a gap behind the pipe (because separation is permitted), but the analyses performed didn’t consider large displacements/large deformations. It could only be captured the initial pipe-soil interaction behaviour when the maximum resultant force was mobilized. The associated maximum displacements were small (4.5% of the pipe diameter).

775

3.2

Resultant force versus displacement curves

Figures 7 and 8 show the resultant force FR (FR = (F2H + F2V )1/2) versus displacement δ relationship for

Figure 9. Resultant forces with α angle and H /D ratio.

Figure 6. Displacement vector distribution for H /D = 100% and α = 135◦ .

Figure 10. Comparison between SIGMA/AEEPECD lateral forces and ALA (2005).

Figure 7. Resultant force versus displacement for α = 45◦ .

For a better evaluation of the resultant force variation with α and H /D ratio, the Figure 9 is presented below. It can be seen that the resultant forces increase with the α angle, and also with the pipe embedment ratio H /D. The smallest resistance offered by the soil against the pipe motion occurred for purely vertical upwards pipe movement, because of the lesser weight of soil to be displaced by the pipe, and the highest one was the vertical downwards. 3.3 ◦

Figure 8. Resultant force versus displacement for α = 135 .

six distinct H /D ratios and direction angles α = 45◦ and 135◦ , respectively: The results show that for α = 45◦ the full mobilization of soil resistance happens in earlier displacements than for α = 135◦ . This fact is probably associated with the relevant increase in soil mass involved in failure process from α = 45◦ to 135◦ , as can be seen in Figures 5 and 6.

ALA (2005) presents a series of equations to represent the soil springs in axial, lateral and vertical (upwards and downwards) directions. However, those equations only consider α = 0, 90 and 180◦ . Figure 10 shows a comparison between the lateral force values obtained in this work with SIGMA/ AEEPECD software and ALA (2005): The results show that ALA (2005) data for lateral response (α = 90◦ ) agree well with SIGMA/ AEEPECD. However, according to Borges (2009), the comparisons for vertical directions (α = 0 and 180◦ ), ALA (2005) values are lower than those obtained

776 © 2011 by Taylor & Francis Group, LLC

Comparison with ALA (2005)

Figure 11. Failure envelopes for a pipeline buried in clay.

Figure 12. Failure envelopes normalized by Equation (5).

with SIGMA/AEEPECD, suggesting that the first methodology is more conservative. 3.4 Failure envelopes The resultant loading paths from the numerical analyses of combined displacements in the H-V plane are shown in Figure 11. They follow the same trend of those previously reported for shallow foundations (Bransby & Randolph, 1998), and for “wished-inplace” pipelines (Merifield et al., 2008). In addition, arrows in this chart indicate the directions of pipe motion, indicating the resultant vector of incremental displacements. By plotting the peak values for each loading component in the H-V plane, it is possible to recognize the influence of the relative depth H /D and the direction of movement of the pipeline. Besides that, through Figure 11 it is easy to identify a remarkable coupling effect between the horizontal and vertical forces, and one can define the soil-pipe interaction coupled domain through these failure envelopes. These results show how the soil-pipe interaction in different monotonic loading directions can be conveniently approximated by numerical models using the Finite Element Method, considering the undrained behavior and an elasto-plastic constitutive model for the clayey soil. Horizontal and vertical components of the resultant forces in buried pipes in clay are usually normalized by the soil undrained strength at the pipe invert and pipe diameter, in terms of the Equation (5):

Figure 12 shows the normalized failure envelopes for H /D from 25% to 200%, where a clear increase in horizontal and vertical forces can be observed: It was observed that the failure envelopes are not symmetrical around the horizontal plane, because the undrained shear strength of the soil increases linearly with depth in the models. The maximum horizontal © 2011 by Taylor & Francis Group, LLC

Figure 13. Failure envelopes normalized by the maximum values.

force occurs at α = 90◦ , and the maximum vertical force at α = 180◦ , as expected. Figure 13 presents the same horizontal and vertical force values presented in Figures 11 and 12 but, this time, they were normalized by their respective maximum values. The maximum horizontal force assumed for each H /D ratio was the value associated with α = 90◦ . For the maximum vertical force, the values associated with α = 0◦ and α = 180◦ were adopted for the upward and downward movements, respectively. The result is a set of superimposed curves that can be considered roughly as a single curve. Based on these data, Equation (6) allows to calculate the horizontal and vertical force components acting on a buried pipe moving towards any direction:

4

CONCLUSIONS

This work studied the soil-structure interaction for the case of pipelines subjected to combined vertical and horizontal loadings when buried in a clayey soil in

777

the submerged condition. In other words, results of a numerical investigation on the resistance of the soil against the oblique motion of a pipe in marine clay were presented. The failure envelopes, drawn with the purpose to establish a basis for estimating the maximum resistance offered by the soil, tried to reproduce the force-displacement response for a steel buried pipe in marine clay, obtained from finite element analyses of a pipe with intermediate roughness considering several embedment/diameter ratios. The software SIGMA/AEEPECD, developed by PETROBRAS, was used in these finite element twodimensional analyses to achieve a set of failure envelopes that is able to describe the mobilized soil resistance behaviour during pipe movement. It was found that there is a relationship between the vertical and horizontal loading directions. The curves obtained in this work are similar to those achieved by Bransby and Randolph (1998) for shallow foundations, but capture the additional effects due to the curved shape of the pipe surface. Also, the results were comparable to those obtained by Merifield et al. (2008) for buried pipes. Based on Equation (6), it was possible to determine the horizontal and vertical force components acting on a buried pipe moving towards any direction. The numerical simulations proved to be useful as a pseudo-experimental tests in order to gauge the shape and size of the soil-pipe interaction domain, when displacement directions between the vertical and horizontal are imposed to the pipe. Even though these relationships must be used with criteria, they proved to be useful in design practice, allowing quick access to complex simulations. ACKNOWLEDGEMENTS The authors would like to gratefully acknowledge to TRANSPETRO and PETROBRAS/CENPES for the sponsorship and support, respectively, as well as all people involved in this research program. REFERENCES Amaral, C.S., Costa, A.M., Carvalho, M.T.M. et al. 1997. Description of SIGMA System – Geotechnical Integrated System for Multiple Analyses (in Portuguese), RT SEDEM n◦ 007/1997, PETROBRAS/SUPEN/DIPREX/SEDEM, Rio de Janeiro, Brazil. ALA. 2005. Guidelines for the Design of Buried Steel Pipe. American Lifelines Alliance. http://www.americanlifelinesalliance.org/pdf/ Update061305.pdf. Amaral, C.S. & Costa, A.M. 1998. Methodology for Interpretation of Wilson Piezocone Tests using the Seacalf Penetration System (in Portuguese). Partial Report 600.234 PETROBRAS/CENPES/PDP/MC, Rio de Janeiro, Brazil.

© 2011 by Taylor & Francis Group, LLC

Amaral, C.S., Costa, A.M., Cardoso, C.O. 2002. Soil Reaction due to Riser Displacement in Roncador Field, Partial Report 600.728 PETROBRAS/CENPES/PDP/MC, Rio de Janeiro, Brazil. Borges, R.G. 2008. Guidelines for Soil-Pipe Interaction Design–Phase 1 (in Portuguese), PETROBRAS/CENPES/PDP/MC, RT MC n◦ 062/2008, Rio de Janeiro, Brazil. Borges, R.G. 2009. Numerical Modelling of the Pipe-Soil Relative Movement in the Oblique Direction in Clay (in Portuguese), PETROBRAS/CENPES/PDP/MC, RT MC n◦ 086/2009, Rio de Janeiro, Brazil. Bransby, M.F. & Randolph, M.F. 1998. “Combined Loading of Skirted Foundations”, Géotechnique, v. 48, n. 5, pp. 637–655. Calvetti, F., Di Prisco, C., Nova, R. 2004. “Experimental and Numerical Analysis of Soil-Pipe Interaction”, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, v. 130, n. 12, pp. 1292–1299. Cardoso, C. O. 2005. Methodology for the Analyses and Design of Submarine Pipelines Subjected to High Pressures and Temperatures via Finite Element Method (in Portuguese), D.Sc. Thesis, COPPE/UFRJ, Rio de Janeiro, Brazil. Fernando, N.S.M. & Carter, J.P. 1998. “Elastic Analysis of Buried Pipes under Surface Patch Loadings”. Journal of Geotechnical and Geoenvironmental Engineering, v. 124, n. 8, pp. 720–728. Fugro. 2005. Geotechnical Investigation – Espadarte and Espadarte Sul Fields, Campos, Offshore Brazil. FugroMcClelland Marine Geosciences, Houston, Texas, USA. Guo, P. & Popescu, R. 2002. “Trench Effects on Pipe/Soil Interaction”, In: Proc. 2nd Canadian Specialty Conference on Computer Applications in Geotechnique, Winnipeg, Canada, pp. 261–269. Hansen, J.B. 1961. The Ultimate Resistance of Rigid Piles Against Transversal Forces, The Danish Geotechnical Institute, Bulletin n. 12, pp. 5–9, Copenhagen, Denmark. Lira, W.W.M. 1998. An Integrated Configurable System for Simulations in Computational Mechanics (in Portuguese). M.Sc. Thesis, PUC-Rio, Rio de Janeiro, Brazil. Merifield, R., White, D.J., Randolph, M.F. 2008. “The Ultimate Undrained Resistance of Partially Embedded Pipelines”, Géotechnique, v. 58, n. 6, pp. 461–470. Oliveira, J.R.M.S., Almeida, M.S.S., Almeida, M.C.F. et al. 2005. “Physical and Numerical Modelling of Lateral Buckling of a Pipeline in Very Soft Clay”, In: Proc. International Symposium on Frontiers in Offshore Geotechnics – ISFOG 2005, Perth,Australia, pp. 607–613. Popescu, R. & Konuk, I. 2001. “3D Finite Element Analysis of Rigid Pipe Interaction with Clay”, In: Proc. 10th International Conference on Computer Mechanics Advances in Geomechanics, Tucson, Arizona, v. 2, pp. 1203–1208. Popescu, R., Phillips, R., Konuk, I. et al. 2002. ”Pipe-Soil Interaction: Large-Scale Tests and Numerical Modeling”, In: Proc. International Conference on Physical Modelling in Geotechnics – ICPMG’02, St. John’s, Newfoundland, Canada, pp. 917–922. Zhang, J., Douglas, P.S. & Randolph, M.F. 2002. “Modelling of Shallowly Embedded Offshore Pipelines in Calcareous Sand”, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, v. 128, n. 5, pp. 363–371.

778

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Geotechnical challenges for deepwater pipeline design – SAFEBUCK JIP D.A.S. Bruton, M. Carr & F. Sinclair Atkins Boreas, UK

ABSTRACT: When the SAFEBUCK JIP started research into the interaction between pipelines and marine clays in 2002, no one could have predicted how that research would impact deepwater pipeline design, or how far the research test methods would develop and improve. The complexity that lies behind the lateral and axial pipe-soil response in soft clays has led to a radical overhaul of established geotechnical practice, based on a significant research effort to evaluate, quantify and understand these pipe-soil interaction mechanisms. This work has led to a number of key insights into the fundamental pipe-soil behaviour and the development of models to simulate this behaviour. This paper demonstrates why pipe-soil interaction is so important to pipeline design, and why it is the largest uncertainty that pipeline designers face in addressing the challenges associated with lateral buckling and pipeline walking; with serious implications for field layouts, pipeline configurations and mitigation solutions.

1

INTRODUCTION

When the SAFEBUCK JIP started research into the large displacement cyclic response of pipelines on very soft deep-water marine clays in 2002, it would not have been possible to predict how that research could now impact pipeline design, or how far the research test methods would develop and improve. From the beginning, this JIP pioneered the idea of small-scale lateral pipe-soil tests in a centrifuge, which is now established as a reliable evaluation method for new projects. Neither could the JIP have envisaged the development of a large-scale in-situ test rig called SMARTPIPE® (Hill & Jacob 2008), using the lessons learned from large-scale laboratory tests pioneered by SAFEBUCK. The complexity that lies behind the lateral and axial pipe-soil response in soft clays has led to a radical over-haul of established geotechnical practice in pipesoil interaction. The JIP and many recent projects have invested significant research effort to evaluate, quantify and understand these pipe-soil interaction mechanisms. This work has led to a number of key insights into the fundamental pipe-soil behaviour and the development of models to simulate this behaviour. There have also been some surprises along the way, for example, the observed link between pipe-soil friction response and the flow regime inside the pipeline. This paper demonstrates why pipe-soil interaction is so important to design, and why it is the largest uncertainty that deepwater pipeline designers face. The pipe-soil response affects the design limit states associated with lateral buckling, pipeline walking, route-curve pullout and flowline anchoring. It can also influence the choice of flowline configuration, for example by choosing pipe-in-pipe, increasing the wall thickness or modifying the coatings. These design

© 2011 by Taylor & Francis Group, LLC

issues can also affect the layout and architecture of new developments, requiring careful definition of pipeline routes, bathymetric profiles and the location of production facilities, as well as the design of pipeline crossings and tie-ins to risers and subsea structures. Lateral buckling and pipe walking behaviour is extremely sensitive to pipe-soil interaction forces and there is significant uncertainty associated with the characterisation of these forces in design. A rigorous theoretical understanding of the basic phenomena involved has only recently emerged, principally through research activity associated with the SAFEBUCK JIP.

2

DESIGN CHALLENGES

Any pipeline which is subjected to above ambient temperatures and pressures has a tendency to relieve the resulting high axial stress in the pipe wall by expanding longitudinally. This expansion is resisted by the axial soil resistance between the pipe and the seabed. This restraint causes an axial compressive force to develop in the pipeline, which can cause buckling. Where pipelines are laid on the seabed without trenching, there is no lateral or uplift restraint acting on the pipeline to prevent buckling, apart from the lateral resistance between the pipe and the soil. A key challenge for the design of such pipelines is the control of lateral buckling, pipe walking, or route curve instability. If left uncontrolled, these behaviours can have serious consequences for the integrity of a pipeline. Uncontrolled lateral buckling can lead to high strains and cyclic loads at the buckle crown, which can compromise the integrity of a pipeline, due to

779

levels of 1 to 10 kPa, associated with pipeline submerged weights; (ii) the near-surface soils to a depth of one or perhaps two metres of the seabed; (iii) more significant disturbance of the seabed, from the pipe embedment process and large-amplitude lateral and axial movements. It is important to bound the soil properties as tightly as possible (Bruton et al., 2007), to minimise the range of lateral and axial friction response.

3.1

Figure 1. Engineered buckle initiators.

local buckling of the pipe section, weld fracture or fatigue failure. Careful consideration is also required to prevent unplanned ‘rogue’ buckles forming at inline structures or pipeline crossings. Lateral buckling is therefore controlled using engineered buckle initiators (Figure 1) placed at regular intervals along each flowline, thereby triggering regular lateral buckles and minimizing the lateral buckle loads (Sinclair et al., 2009). Pipe-walking can occur when a pipeline is subjected to thermal cycling while tension is applied at one end by a riser, or the pipeline is laid down a slope or is subjected to steep thermal transients during shutdown and restart cycles (Carr et al., 2006 & Bruton et al., 2010). Over a number of cycles, this movement can lead to very large global axial displacements, with associated overload of the connections. Where the predicted pipe walking displacements threaten system integrity, walking will be controlled by the use of pipeline anchors, typically installed at the end of the pipeline from which it is walking. However, pipeline anchors result in very high levels of tension at shutdown, which can lead to route-curves becoming unstable. This is usually overcome by increasing the routecurves radius, or removing route curves altogether by changing the field layout.

3

GEOTECHNICAL UNDERSTANDING

Conventional methods of geotechnical investigation and analysis that have evolved for foundation design are not suited to the analysis of pipe-soil behaviour. Compared with conventional foundation design, pipesoil interaction is concerned with (i) far lower stress

© 2011 by Taylor & Francis Group, LLC

Geotechnical survey data – from the field

To narrow the range of soil properties, it is essential that a detailed soil investigation provides undrained shear strength profiles, sensitivity, and unit weight in as much detail as possible. Tests carried out in-situ on softer soils (or box core samples of those soils) should include T-bar tests, including cyclic loading to quantify the remoulding behaviour. It is also critical to provide a good datum reference to the original surface of the seabed. Often data is truncated to remove very weak surface soils (or the weak slurry is removed from the box-core before testing) and yet this information is fundamental to the assessment of pipeline embedment, which affects every aspect of the subsequent pipe-soil response.

3.2

Project specific testing and interpretation

It is common now to supplement geotechnical survey investigations with project-specific model test programmes, to assess the axial and lateral pipe-soil response. Project-specific testing requires the collection of bulk samples of near-surface soils for subsequent laboratory tests and pipe-soil modelling studies (Bruton et al., 2009). The SAFEBUCK JIP reviewed historical data and conducted a number of pipe-soil interaction test programmes to provide generic guidance for future projects and improve current understanding (Bruton et al., 2006). Although some of the historical test data had limited information on some of the test parameters, the proposed formulations from this work remain sound for relatively light pipes. However, pipes that are relatively heavy have a significant influence on the cyclic response (Bruton et al., 2008). ‘Heavy’ and ‘light’ pipes are distinguished by the ratio of the pipeline weight to the seabed strength. In simple terms, values of V/Su .D <1.5 give a ‘light-pipe’ response, characterised by the pipe rising during the initial lateral breakout, while values of V/Su .D > 2.5 give a ‘heavy-pipe’response, characterised by the pipe diving with displacement. Data from the early tests has since been augmented by a large quantity of high quality project-specific tests donated by JIP participants and tests carried out by SAFEBUCK, to create a database that now spans a very wide range of pipe diameters, weights and soil conditions. A detailed review of this lateral friction response database was recently issued to participants (White & Cheuk, 2010).

780

It was also clear from this review that the quality of test data has improved significantly over recent years, with improved test equipment and testing methods, based on lessons learned from earlier tests. For example, SAFEBUCK pioneered the idea of small-scale lateral pipe-soil tests in a centrifuge and this type of test is now established as a reliable evaluation method for new projects, requiring smaller soil samples and much reduced time-scales for testing. The lessons learned have also been captured in large-scale laboratory tests and, more recently with the development of a largescale in-situ test rig called SMARTPIPE® (Hill & Jacob, 2008). 3.3

Remaining uncertainties

Nevertheless, much uncertainty remains in key areas of pipe-soil interaction response: 1. Complex lateral frictional response of pipe that is heavy in relation to the soil shear strength 2. Predicting axial friction in soft clays. 3. The influence of internal flow regimes on lateral buckling, pipe-walking and route-curve pull-out These issues are discussed in the following sections. 4

‘HEAVY’ PIPE RESPONSE

Increasing the flowline wall thickness increases its bending stiffness and capacity but this also increases the load in the pipeline due to the lateral resistance increasing with pipe weight. The restrained thermal expansion force also increases with wall thickness but the overall influence of increasing wall thickness at a lateral buckle is not obvious and depends upon the lateral friction response. If the lateral resistance is proportional to the pipe download, then the buckling becomes more severe as the wall thickness increases. However, if the lateral resistance does not change with the pipe download, then the buckling becomes less severe as the wall thickness increases. The use of pipe-in-pipe (PIP) systems has significant advantages for design over more traditional externally-insulated systems. Although PIP systems are usually more expensive per metre than external insulation systems and are much more complex to assemble offshore, they offer advantages in terms of improved thermal insulation and enhanced lateral buckling and pipe walking performance. The good thermal performance of PIP systems is well known. The ability to tailor the bending stiffness and moderate the axial load by increasing the outer pipe wall thickness (which usually experiences little thermal loading) is also known to greatly assist lateral buckling design. Finally the increased submerged weight of such systems (in comparison with externally insulated pipe) can greatly reduce or eliminate the propensity for pipe walking, due to the high weight and increased axial soil resistance. Thus they may actually offer a cost advantage. However, the use of heavy pipe-in-pipe systems introduces an extremely challenging lateral © 2011 by Taylor & Francis Group, LLC

781

Figure 2. Cyclic embedment at a lateral buckle.

cyclic response into the design process (Bruton et al., 2009). Under conditions of constant vertical load, a laterally sweeping ‘heavy’ pipeline will rapidly embed. However, in the areas where there is little cyclic movement, such as at the stationary inflection point in a lateral buckle, there is little embedment. This causes the contact pressure to vary, so that the rate of embedment and the lateral soil resistance varies along the length of the buckle. Therefore, in regions of lateral sweeping the contact force and lateral soil resistance gradually reduce as the trench gets deeper and soil berms are established at the extremes of displacement. Consequently, the friction curves formulated using the nominal pipe weight can significantly overestimate the lateral resistance at the crown of the buckle. The structural response of heavy pipe therefore required a totally new approach, to capture the reducing contact pressure along the buckle as the lobes of the buckle displace laterally and penetrate vertically. Meanwhile, the regions between the lobes experienced little lateral movement and sustained increasing contact pressure as the load is transferred to these regions. Modelling of the changing seabed elevation as the pipe sweeps and displaces soil to form a trench is extremely challenging. Figure 2 shows a section through a lateral buckle model, where the excavation of the seabed is captured at the end of each sweep of the pipe, based on predictions from experimental data that account for variation in contact pressure and the sweep amplitude. This complex modelling approach can be used to assess the cyclic response for a selected pipe weight and pipe bending stiffness. As the lateral buckle displaces cyclically in operation, the analysis defines the shape of the lateral buckle and the variation of contact pressure along it. To do this, sufficient knowledge of the rate of vertical displacement and lateral resistance for the given pipe weight and cyclic amplitude is required from project specific tests. This approach has been used to verify a simpler method of analysis, using re-formulated friction curves to suit more traditional lateral buckling models that employ an elastic seabed surface that is not deformed by cyclic loading. The re-formulated friction curves exhibit a heavy pipe response in the early cycles which gradually migrates towards a light pipe response, as the contact pressure reduces with

increasing embedment. Work is ongoing to simplify this necessarily complex approach but the ultimate solution is a better way of modelling lateral pipe soil interaction in finite element analysis (FEA), which is a key aim of the SAFEBUCK GEO JIP, described further in Section Section 7. A major challenge to designers, associated with changing operating conditions and cyclic loading, is the predicted increase in operating temperature late in field life. Often this is combined with increasing density of the production fluids as the level of produced water increases. This combination of increasing load and lateral resistance may cause a pipeline to break through an established soil berm at the apex of a lateral buckle, causing strain localisation in the pipe. Assessing this potential failure condition requires a very good understanding of the gradual growth in berm strength and the ability to model cyclic response of the lateral buckle over many operating cycles.

pipelines, is the influence of internal flow mechanisms on the axial friction response. On a relatively light pipeline with external insulation, the weight of the pipe can vary significantly as alternate slugs of gas and liquid pass along the pipeline. This frequent change in contact pressure is likely to affect the excess pore pressure in the soil around the pipeline, causing changes in the axial friction response. Observations of operating pipelines has certainly demonstrated very low levels of axial friction in operation when regular slugging occurs, compared with much higher levels of axial friction on shutdown when slugging has ceased. This is the only plausible explanation for some observed end expansion behaviour and is thought to be responsible for high levels of tension at shutdown that has resulted in route-curve instability.

6

PREDICTING PIPELINE WALKING AND END DISPLACEMENTS

5 AXIAL FRICTION IN SOFT CLAYS 6.1 Walking response of long pipelines 5.1

Drained and undrained axial response

Drained axial friction is relatively straightforward to establish using devices such as a tilt table (Najjar et al. 2003), or direct shear or ring-shear devices adapted for low stress levels. Undrained axial friction is much more complex to predict, although fast direct shear or ring shear tests can be performed. Undrained axial resistance can lead to a very low lower-bound friction value that dominates the pipewalking response. Recent large-scale axial tests (Bruton et al., 2009) showed that undrained (fast) and drained (slow) axial movements lead to dramatic differences in the axial resistance, due to the generation of excess pore pressure. Although an effective stress friction approach provides a consistent interpretation of both drained and undrained responses, it is necessary to assess and predict the development of excess pore pressure in order to calculate the axial resistance in undrained conditions. The excess pore pressure is thought to be a function of: 1. Pipe velocity – where the range of pipe velocities can span the full undrained to drained response; 2. Pipe weight – where heavier pipes are associated with higher levels of excess pore pressure; 3. Cumulative displacement – where the generation of excess pore pressure can reduce with displacement. The important transition from undrained to drained conditions is not well understood. Improving understanding is a key aim of current research, including the SAFEBUCK GEO JIP. 5.2

Influence of internal flow regime on axial resistance

Another unexpected interaction mechanism that has recently emerged, from observations of operating

© 2011 by Taylor & Francis Group, LLC

782

Lateral buckles along a pipeline can significantly influence both pipeline walking and end expansion predictions. When a pipeline undergoes lateral buckling it effectively splits the pipeline into a number of shorter lines between buckles. The pipeline expansion feeds into and out of each lateral buckle (as well as the pipeline ends) with each shutdown and restart cycle (Carr et al., 2006). As a result of differential expansion the size and shape of the buckles change. This modifies the effective axial force profiles and hence the walking response. As pipe feeds into and out of each buckle, the resistance to lateral displacement of the pipe gradually increases as soil berms build up to each side of the buckle, which can cause gradual axial displacement over long lengths of the pipeline. Meanwhile, the virtual anchors that form between each buckle shift between load and unload conditions causing additional localised walking of the pipeline that can significantly influence cumulative pipeline end expansion. Understanding of this complex cyclic interaction is improving and methods for predicting the cyclic response analytically or using FEA techniques are available (Bruton et al., 2010). However, the walking response is extremely sensitive to the axial pipe-soil resistance, Figure 3. The figure shows the typical end expansion response of a long pipeline with multiple buckles. In this example, the walking behaviour is driven by slope (making the displacement more positive), thermal transients (making the displacement more negative) and a changing buckle force profile. The figure shows the load–unload end expansion over a number of shutdown/start-up cycles; the position on load is highlighted by the thick lines. For low axial resistance (µA = 0.4), the initial pipeline expansion is approximately 0.65 m, which falls to 0.2 m on unload. Over the first 20 shutdown

gas accumulating at the top of the slope. This density variation can increase the rate of pipeline walking and significantly modify the shutdown response of lateral buckles (Bruton et al., 2010).

7

SAFEBUCK GEO AND PHASE III

The issues raised in this paper are the subject of ongoing research both for projects and for the SAFEBUCK Joint Industry Project. The current phase of SAFEBUCK GEO has two main aims:

Figure 3. Impact of axial friction on walking.

cycles, the end expansion increases rapidly. Thereafter the rate of increase slows, but there is a steady increase in end expansion on each cycle; the end expansion reaches 1.9 m after 100 cycles. Depending upon the number of cycles anticipated during the field life, anchors may be required. If the axial friction is increased (to µA = 0.8), there is a fundamental change in behaviour. The initial expansion is reduced modestly, to 0.5 m, which falls to 0.15 m on unload.The end expansion increases slightly over the next 40 shutdown cycles, to reach a maximum expansion of 1.1 m, and then begins to reduce as the thermal transients begin to dominate the response. After 100 cycles the end expansion on load is reduced to 1 m and falling. 6.2

Restraining walking – operational feedback

For long pipelines, predicting this cyclic response is extremely challenging, making the design requirements to mitigate walking extremely uncertain. This is why some project have installed anchors on some pipelines and only made provision for anchoring on others (Jayson et al., 2008). Several recent deepwater projects with and without pipeline anchors are now in operation and feedback on the walking response has gradually emerged, though Integrity Monitoring activities that are essential for pipelines that experience cyclic lateral and axial displacements (Baker et al., 2006). This includes the observation of an unexpectedly rapid pipeline walking response, which led to the identification of a multiphase flow mechanism as the main cause. During shutdowns in multiphase pipelines, the flow stream can quickly separate into gas and liquid; which can result in significant variations in submerged weight along the pipeline. This is a major influence on both lateral and axial friction response. In pipelines on steep slopes, liquids settle at the bottom of the slope with © 2011 by Taylor & Francis Group, LLC

1. To fundamentally improve the way that lateral pipesoil response is addressed in design by developing a new ‘force-resultant plasticity model’ to run inside standard software packages, which will capture experience from modelling and testing of lateral pipe-soil interaction. 2. To improve our understanding and quantify key uncertainties in predicting the axial friction response, by a review of all recent project-specific tests, supplemented by additional JIP tests. The current phase of SAFEBUCK Phase III has a number of aims, one of which is to collect and share data and lessons learned from operating pipelines. This will include the pipe-soil response observed on operating pipelines, including embedment levels, axial displacements and the detailed responses at lateral buckles. The other aim is to formalize the SAFEBUCK Design Guidance (Carr, Bruton & Cosham, 2008), including the approach to pipe-soil response, within a recognized DNV Industry Recommended Practice.

8

NOMENCLATURE

V = vertical pipe-soil force Su = soil undrained shear strength D = outside diameter of pipe (including coatings) µA = axial friction factor

783

REFERENCES Baker, J.H.A., Bruton, D.A.S. & Matheson, I.C. 2006. Monitoring and Effective Integrity Management of Laterally Buckled Flowlines in Deep Water. Offshore Technology Conference. OTC-17932. Bruton, D., White, D., Bolton, M., Cheuk C. & Carr, M. 2006. Pipe/Soil Interaction Behavior during Lateral Buckling. SPE Projects, Facilities & Construction SPEPFC106847. Bruton, D.A.S., Carr, M. & White, D. 2007. The Influence of Pipe-Soil Interaction on Lateral Buckling and Walking of Pipelines – The SAFEBUCK JIP. 6th International Offshore Site Investigation and Geotechnics Conference. Bruton, D.A.S., White, D., Carr, M. & Cheuk, C. 2008. PipeSoil Interaction During Lateral Buckling and Pipeline Walking – The SAFEBUCK JIP. Offshore Technology Conference. OTC-19589. Bruton, D.A.S., White, D., Langford, T. & Hill, A. J. 2009. Techniques for the assessment of pipe-soil interaction

forces for future deepwater developments. Offshore Technology Conference. OTC-20096. Bruton, D.A.S., Sinclair, F. & Carr, M. 2010. Lessons Learned From Observing Walking of Pipelines with Lateral Buckles, Including New Driving Mechanisms and Updated Analysis Models. Offshore Technology Conference OTC20750. Carr, M., Sinclair, F. & Bruton, D. 2006. Pipeline Walking – Understanding the Field Layout Challenges, and Analytical Solutions developed for the SAFEBUCK JIP. Offshore Technology Conference. OTC-17945. Carr, M., Bruton, D. & Cosham, A. 2008. Design Guideline. SAFEBUCK JIP. (Confidential to JIP Participants) Jayson, D., Delaporte, P., Albert, JP., Provost, M.E., Bruton, D. & Sinclair, F. 2008. Greater Plutonio Project – Subsea Flowline Design and Performance. Offshore Pipeline Technology Conference.

© 2011 by Taylor & Francis Group, LLC

Hill, A.J. & Jacob, H. 2008 In-Situ Measurement of PipeSoil Interaction in Deep Water. Proc. Offshore Technology Conference, Houston, USA. Paper OTC-19528. Najjar, S.N., Gilbert, R.B., Liedtke, E.A. & McCarron, W. 2003. Tilt Table Test for Interface Shear Resistance between Flowlines and Soils. International Conference on Ocean, Offshore and Arctic Engineering. OMAE-37499. Sinclair, F., Carr, M. Bruton, D. & Farrant, T. 2009. Design Challenges and Experience With Controlled Lateral Buckle Initiation Methods. International Conference on Ocean, Offshore and Arctic Engineering. OMAE-79434. White, D.J. & Cheuk, C.Y. 2010. SAFEBUCK Joint Industry Project – Pipe-Soil Interaction Models for Lateral Buckling Design – Phase IIa Data Review. Univ. of W. Australia Report GEO: 09497v2. (Confidential to JIP Participants)

784

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Large deformation finite element analysis of vertical penetration of pipelines in seabed S. Chatterjee, M.F. Randolph, D.J. White & D. Wang Centre for Offshore Foundations System, The University of Western Australia

ABSTRACT: A large deformation finite element analysis method has been developed based on frequent remeshing and interpolation, linked to small strain analyses performed using the finite element package, ABAQUS. Initially, in order to validate the finite element approach, a set of simulations were undertaken based on the parameters obtained from a set of centrifuge model tests. Numerical simulation results matched well with the centrifuge results, provided both the effects of strain rate and softening were taken into account. The results of the subsequent parametric study showed that the vertical capacity depends strongly on the shear strength profile of the seabed, but is best correlated to the nominal shear strength at the depth of the pipeline invert. A simple approach to allowing for the rate of loading (in respect of strain rate dependency of the shear strength) and partial softening of the soil is described. It is concluded that predicting vertical embedment of pipelines on the seabed can be erroneous if secondary effects are not taken into consideration, including strain rate and softening, local soil heave and buoyancy forces. 1

INTRODUCTION

As offshore energy extraction facilities are gradually shifting to deeper water, greater lengths of pipeline are required, both within the fields and for transporting the hydrocarbon products to land. For the most part, flowlines and export pipelines are just laid on the seabed and become partially embedded due to their self weight and the lay process. Pipes carrying oil at high temperature and pressure tend to deform axially. This is resisted by frictional forces from the seabed soil, resulting in high compressive forces in pipe, which in turn tend to buckle the pipe. Buckling is helpful in reducing excessive compressive force in the pipe but leads to high bending stresses in the pipe wall. For this reason, controlled buckling is a common design practice (Bruton et al., 2006). The lateral buckling response depends on the pipeline embedment and also on the lateral resistance provided by the seabed soil. To achieve controlled buckling, it is essential to estimate the initial vertical pipe embedment and pipe soil interaction forces. Vertical penetration of pipelines in the seabed has been widely researched. Solutions based on classical plasticity theory (Randolph & Houlsby, 1984 Murff et al., 1989, Randolph & White, 2008b) are available in the literature. Many studies (Aubeny et al., 2005; Merifield et al., 2008; Merifield et al., 2009) have been conducted with the help of small strain finite element analysis. Experimental data from centrifuge model tests are also available (Dingle et al., 2008). When a pipe is embedded into the seabed, heave occurs at the sides of the pipe. This phenomenon has been neglected in most of the theoretical studies

mentioned above. Also, dynamic pipe lay effects result in significant amount of softening in the surrounding soil (Randolph & White, 2008a). Effects of strain rate and strain softening on the vertical resistance of pipeline have mostly been ignored in previous studies. For this study, a large deformation finite element approach based on the commercial finite element software ABAQUS (Dassault Systèmes, 2007) was developed, following a similar approach to that described by Wang et al. (2010). The ‘remeshing and interpolation technique with small strain’ (RITSS: Hu & Randolph 1998) method was used, with the original shear strength of the soil modified to account for any increase due to high strain rates and any decrease due to softening (through the accumulation of plastic strain). After the results had been validated by comparing with existing centrifuge test data, a parametric study was undertaken over a wide range of soil properties.

2 2.1

Methodology

A large deformation finite element (LDFE) method based on the RITSS approach was developed. The basis of LDFE analysis using this approach is to divide the large displacements into a series of small strain analyses. Python, the in-built scripting language of ABAQUS, was used for writing pre-processing and post-processing codes. First of all, a python script was written to generate the input file of the first step. This was submitted to ABAQUS for small strain analysis

785 © 2011 by Taylor & Francis Group, LLC

LARGE DEFORMATION FINITE ELEMENT METHOD

2.3

A total stress approach, assuming an undrained soil response, was adopted for the analyses. A MohrCoulomb plasticity soil model was used with zero friction angle, which is equivalent to the Tresca failure criterion. The pre-failure soil response is assumed to be elastic, with Young’s Modulus E taken equal to 500su , where su is the undrained shear strength of the soil. Poisson’s ratio ν was chosen as 0.499 to minimise volumetric strain. This simple elasto-plastic soil model was modified to account for effects of strain rate and strain softening. The limiting shear strength in the pipe-soil interface, τmax = α sum , with α = 0.5 was assumed, where sum is the shear strength of soil at the mudline.

Figure 1. Typical initial mesh and boundary conditions.

and the output database was post-processed with the help of another python script. Nodal displacements, stresses and other parameters on integration points were recorded.The recorded nodal displacements were used to establish the new displaced boundary of the system. The whole model was then remeshed with the help of a Python script. A Fortran code was written to recover stresses and strains from the old integration points to the old nodes. The superconvergent patch recovery (SPR: Zienkiewicz & Zhu 1993) method was used for stress or strain recovery. These parameters were then interpolated from the old nodes to the new integration points using a subroutine written in Fortran. Interpolated total and plastic strains were used to modify the original shear strength to account for the effects of strain rate and strain softening on shear strength. The detailed methodology regarding this is described later in the paper. Stresses at new integration points are given as initial conditions of the remeshed structure, using the in-built ABAQUS subroutine SIGINI. With these initial conditions, the remeshed model is then submitted for another small strain analysis. The cycle is repeated until the required total displacement is achieved. The whole process is controlled by a main Fortran program, which calls other subroutines or submits Python scripts as required, with no need for user intervention during the analysis.

2.2

Model and boundary conditions

A 2-D plane strain model was used with the pipe considered as rigid and the soil as deformable. Vertical boundaries of the soil were free to move vertically but restrained horizontally. The bottom horizontal boundary was fully fixed, while the top surface was free to move. Plane strain element CPE6 with 3 vertex nodes and 3 mid-side nodes was used for soil elements. Fine meshing, with a minimum size of D/20 (where D is the pipe diameter), was used near the pipe (see Figure 1). At each step, a small displacement (1% of the diameter of the pipe) was applied to the centre of the pipe.

2.4

Effects of strain-rate and softening

Effects of strain rate and strain softening on the shear strength of soil are well accepted. The shear strength of soil is increased by increasing strain rate and decreased as it is remoulded under accumulating shear strains, depending on the soil sensitivity and ductility. The combined effects of strain-rate and softening can be incorporated by extending the simple Tresca soil model, multiplying the original shear strength by two factors (Einav & Randolph 2005, Zhou & Randolph 2007). Here, the undrained shear strength at integration points was modified prior to any step in the analysis according to

Where the first part of the equation adjusts the strength according to the strain rate effect and the second part allows for strain softening. ε1 and ε3 are the major and minor principal strains, respectively, resulting from a displacement increment, δ/D, where D is the pipe diameter. Vp is the pipe velocity. µ is the rate of strength increase per decade of strain rate, which is usually taken in the range of 0.05–0.2, γ˙ ref is the reference strain rate. Softening is taken as an exponential function of the cumulative absolute plastic shear strain ξ. Here δrem denotes the ratio of fully remoulded strength to the initial strength, hence is the inverse of sensitivity, St . The parameter ξ95 reflects the relative ductility of soil and is the value of ξ at which the soil has undergone 95% remoulding. 2.5

Comparison with centrifuge result

Initially, a set of parameters (shear strength profile of soil, unit weight of soil, pipe diameter) were chosen to match conditions in a centrifuge model test (Dingle et al., 2008). The soil used in the study was lightly over consolidated with strength at mudline, sum = 2.3 kPa, linear shear strength gradient k = 3.6 kPa/m and with

786 © 2011 by Taylor & Francis Group, LLC

Constitutive model

Figure 2. Comparison with centrifuge data. Table 1.

Case I II III IV V VI VII

Figure 3. Variation of penetration resistance with embedment for different cases.

Shear strength parameters during parametric study. Pipe Dia. m

k kPa/m

sum kPa

κ

0.8

1.00 1.00 4.00 6.25 8.00 10.0 10.0

20.0 10.0 8.00 5.00 1.60 1.00 0.50

0.04 0.08 0.40 1.00 4.00 8.00 16.0

Three values of submerged unit weight of soil (3 kN/m3 , 5 kN/m3 and 7 kN/m3 ) and three values of sensitivity of soil (1, 3 and 6) were chosen and repeated with all the cases presented in Table 1. Strain rate parameters described in first part of equation 1 were kept constant during all the simulations with: µ = 0.1; Vp / Dγ˙ ref = 5000. 4

submerged unit weight, γ
= 6.5 kN/m3 . With these parameters the pipe was vertically embedded to a depth of 0.45D. The resulting vertical reaction force V was then normalised by Dsu0 , where D is the diameter of the pipe and su0 is the original shear strength at te invert of the pipe. The normalised vertical capacity V/Dsu0 is plotted against embedment ratio w/D in Figure 2. It can be seen from Figure 2 that if effects of strain rate and softening are not considered the response does not match the centrifuge result well. If both the effects of strain rate and softening are considered, results are matching quite well with the centrifuge test data, where the pipe was penetrated at a normalised rate Vp /D = 0.015 s−1 .

4.1

PARAMETRIC STUDY

3.1 Parameters chosen Shear strength parameters, submerged unit weight and the sensitivity of the soil were varied in a parametric study. The shear strength su at any depth z is expressed by su = sum + kz, where sum is the shear strength at mudline and k is the shear strength gradient. Values of sum and k were varied so that a non-dimensionalised shear strength gradient κ (= kD/sum ) captured a wide range of values. Table 1 shows all the shear strength parameters chosen for the study.

Effects of unit weight variation

To investigate the effects of varying the submerged unit weight, the vertical response is plotted in Figures 4 and 5 for the two extreme values of κ, and for the three submerged unit weight values (γ
= 3 kN/m3 , 5 kN/m3 and 7 kN/m3 ). It is clear from these that the effect of the soil weight is much greater in the case of soils with low shear strength at the mudline, and hence high values of γ
D/sum , as would typically occur where the value of κ is high. For low γ
D/sum , the effect of unit weight is negligible. Note that for the analyses in Figure 5, the value of γ
D/su0 (normalising using the soil shear strength at pipe invert) ranges from 0.53 to 1.1, so still much greater than for the analyses in Figure 4.

787 © 2011 by Taylor & Francis Group, LLC

Effects of shear strength profile variation

The non-dimensionalised vertical reaction force is plotted against embedment ratio for different values of κ in Figure 3. For a particular submerged unit weight and sensitivity of soil, the responses are quite similar. This confirms the general approach of normalising the vertical reaction force by the shear strength at the pipe invert. It can, however, be seen that although the vertical response is more or less similar overall, the normalised penetration resistance can vary by as much as 40% at lower embedment ratios.

4.2 3

RESULTS

Figure 4. Effect of submerged unit weight on penetration resistance (κ = 0.04).

Figure 6. Best fit single power law curve for soil with St = 3 (geotechnical resistance only). Table 2. Comparison of power law coefficients for bearing capacity factor, Nc , from different studies (for 0 < w/D < 0.5).

Aubeny et al. (2005) Merifield et al. (2008) Merifield et al. (2009) Present Study

Figure 5. Effect of submerged unit weight on penetration resistance (κ = 16).

The penetration resistance during embedment has two components: the geotechnical resistance and buoyancy due to of the displaced soil. It can be expressed as

The geotechnical resistance, given by the Nc term, can be fitted by a simple power law expressions with coefficients ‘a’ and ‘b’. The buoyancy is expressed in terms of As , the (nominal) submerged cross-sectional area of the pipe and an adjustment factor, fb , to account for local heave. If Archimedes’ principle were to apply (i.e. for a completely level soil surface), the value of fb should be 1. However, as shown by Merifield et al. (2009), the value of fb is around 1.5 when heave is considered. To evaluate a best-fit power law curve to the geotechnical resistance, it is first necessary to subtract the effect of buoyancy from the total capacity. This © 2011 by Taylor & Francis Group, LLC

a

b

Comments

6.73

0.29

7.4

0.4

7.1

0.33

6.8

0.22

Wished-in-place, no strain rate effects Wished-in-place, no strain rate effects Pushed-in-place, no strain rate effects Pushed-in-place, strain rate effects

brings together the responses in Figures 4 and 5. For soils with low values of κ, if fb = 1.5 is assumed, the response merge into a single trend. For soils with high κ, a more appropriate value of fb = 1.9 was required to bring the curves together. Once the geotechnical resistance was isolated, values at a particular embedment were averaged for different κ. This gave an average vertical response curve for soil with particular strain rate and softening parameters. This curve was fitted to the power law expression as shown in Figure 6. The coefficients ‘a’ and ‘b’ are compared in Table 2 with other published values (all for fully rough pipe-soil interfaces). In this way of representing the penetration resistance, the effect of κ is neglected, other than by taking the invert soil strength, su0 . Though this approach is reasonable for embedment ratio w/D ≥ 0.1, there can be significant discrepancy for w/D < 0.1 (as is also evident from Figure 3). The power law fit can be improved for the portion w/D ≥ 0.1, although becomes less good for w/D < 0.1. For this reason, separate power laws have been fitted for the portions less than or greater than w/D = 0.1 for different κ. For w/D < 0.1, the expression

788

Table 3. Nc power law coefficients for w/D ≥ 0.1 for different κ.

κ = 0.04 κ = 0.08 κ = 0.40 κ = 1.00 κ = 4.00 κ = 8.00 κ = 16.0

a

b

6.536 6.517 6.431 6.242 6.015 5.809 5.751

0.170 0.167 0.154 0.139 0.117 0.107 0.114

Figure 8. Effect on penetration resistance of varying softening parameter ξ95 .

Figure 7. Effect on penetration resistance of varying strain rate parameter µ.

was used to get a good match with the FE data. The value of m depends on κ. For low κ, m is 0.55–0.5, whereas it is 0.35–0.25 for κ > 1. The power law fit coefficients for w/D ≥ 0.1 for different κ is presented in Table 3.

Figure 9. Effect on penetration resistance of varying sensitivity.

5

4.3 Effects of strain rate and softening Strain rate and strain softening parameters also have a marked effect on the penetration resistance. The effects of variations in strain rate parameter µ, softening parameters ξ95 and St , as described in Equation 1, were explored. Figure 7 shows the effect of varying µ whilst keeping other parameters constant. It can be seen that the penetration resistance can vary up to as much as 56% due to varying µ over the plausible range (for the adopted dimensionless pipe penetration velocity, Vp /Dγ˙ ref = 5000.). Softening parameters ξ95 and St have smaller effects on penetration resistance. As shown in Figures 8 and 9, the penetration resistance increases as ξ95 is increased, and decreases with increasing sensitivity. The penetration resistance can vary by up to 6% due to varying ξ95 , while it can vary by up to 17% for different values of sensitivity in the range 1–6.

© 2011 by Taylor & Francis Group, LLC

SUMMARY AND CONCLUSIONS

The as-laid embedment of a pipeline is extremely important for many aspects of pipeline design, in particular for the assessment of lateral buckling of deep water pipelines. In this paper, a large deformation finite element method has been developed which addresses some of the short-comings of the previous studies in this field. The LDFE method implemented is a robust approach which captures the changes in geometry, in particular the adjacent soil heave, as the pipe is pushed continuously into the soil. The method can also account for the effects of strain rate and strain softening on the shear strength of soil. The present method was validated by comparing results with data from centrifuge model tests.An excellent match between the result of this study and the centrifuge test data was obtained.

789

A detailed parametric study was performed to explore the effects of varying shear strength, buoyancy, strain rate and softening parameters on the pipeline penetration resistance. Improved curve fits were obtained for the penetration resistance for base case strain rate and strain softening parameters. These secondary parameters were also found to have an effect on the predicted vertical response. ACKNOWLEDGEMENTS This work forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established under the ARC Research Centres Program and now supported through Centre of Excellence funding from the State Government of Western Australia. The first writer is supported by an International Postgraduate Research Scholarship and a University Postgraduate Research Award from the University of Western Australia. REFERENCES Aubeny, C. P., Shi, H., & Murff, J. D. 2005. Collapse load for cylinder embedded in trench in cohesive soil. International Journal of Geomechanics, 5(4), 320–325. Bruton D. A. S., White D. J., Cheuk C. Y., Bolton M. D. and Carr M. C. 2006. Pipe-soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Proc. Offshore Technology Conf., Houston, Paper OTC 17944. Dingle, H. R. C., White, D. J. & Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout on soft clay. Canadian Geotechnical Journal, 45, 636–652. Einav, I., & Randolph, M. F. 2005. Combining upper bound and strain path methods for evaluating penetration resistance. Int. J. Numer. Meth. Engng, 63, 1991–2016.

Dassault Systèmes 2007. Abaqus analysis users’ manual, Simula Corp, Providence, RI, USA. Hu, Y., & Randolph, M. F. 1998. A practical numerical approach for large deformation problems in soil. Int. J. for Num. & Analytical Meth. in Geomechanics, 22, 327–350. Martin, C. M. & Randolph, M. F. 2006. Upper-bound analysis of lateral pile capacity in cohesive soil. Geotechnique, 56(2), 141–145. Merifield, R. S., White, D. J. & Randolph, M. F. 2008. The ultimate undrained resistance of partially embedded pipelines. Géotechnique, 58(6), 461–470. Merifield R. S., White, D. J. & Randolph, M. F. 2009. Effect of surface heave on response of partially embedded pipelines on clay. ASCE Journal of Geotechnical and Geoenvironmental Engineering, 135(6), 819–829. Murff, J.D., Wagner, D.A., & Randolph, M.F. 1989. Pipe penetration in cohesive soil. Géotechnique, 39(2), 213–229. Randolph, M. F. & Houlsby, G. T. 1984. The limiting pressure on a circular pile loaded laterally in cohesive soil. Geotechnique, 34(4), 613–623. Randolph, M. F. & White, D. J. 2008a. Pipeline embedment in deep water: process and quantitative assessment. Proc. Offshore Technology Conference, OTC19128. Randolph, M. F. & White, D. J. 2008b. Upper-bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique, 58(4), 297–301. Wang, D, White, D.J. & Randolph, M.F. 2010. Large deformation finite element analysis of pipe penetration and largeamplitude lateral displacement. Canadian Geotechnical Journal (accepted April 2009, in press for publication). Zhou, H. & Randolph, M. F. 2007. Computational techniques and shear band development for cylindrical and spherical penetrometers in strain-softening clay. International Journal of Geomechanics, 7(4), 287–295. Zienkiewicz, O. C. & Zhu, J. Z. 1993. The superconvergent patch recovery and a posterior error estimates. Part 1: The recovery technique. International Journal for Numerical Methods in Engineering, 33, 1331–1364.

790 © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Implementation of geotechnical techniques in the analysis of pipeline response G. Cumming & N. Brown J P Kenny Pty, Perth, Australia

ABSTRACT: Several design approaches can be used to analyse the interface between offshore pipelines and the seabed. This paper summarises some key areas of pipeline design that require the implementation of geotechnical techniques and the state of the art research and testing of the behaviour of that pipeline/foundation interface. The integration of the response of the pipeline to the behaviour of the foundation is a critical interface between pipeline engineers and geotechnical engineers, which requires open and productive dialogue to solve often complex interaction scenarios.

1

INTRODUCTION

The design of submarine pipeline systems to DnV OS-F101 requires that the seabed characteristics, such as uneven, unstable, subsidence, seismic activity and soil properties are factors that shall be taken into consideration for the selection of pipeline location. The seabed properties necessary for evaluating the effects of relevant loading conditions shall be determined from testing of the seabed deposits. The modelling of pipe-soil interaction in pipeline design is discussed in more detail in the recommended practices DnV RP F105, DnV RP-F109, and DnV RP-F110, for pipeline spanning, stability and global buckling respectively, where the seabed material is classified as clay or sand/gravel/rock. Pipeline engineering practice uses conservative upper or lower bounding values to characterize the pipeline response in deterministic analysis; low interface friction to conservatively predict maximum expansion or pipelay tensioner clamping requirements; high interface friction to maximize axial or bending stresses. It is not always clearly defined whether an upper or lower bound presents the most onerous case for pipeline design. This is particularly the case in global buckling design, where high axial friction increases the effective axial force in a pipeline, increasing the potential for buckling to occur, but decreases the feedin (or expansion) into buckle locations. Conversely, low axial friction decreases the effective axial force in a pipeline, decreasing the potential for buckling to occur, but increases the feed-in (or expansion) into buckle locations, increasing the consequences of buckling. The design combination of feed-in due to low friction, and initiation due to high friction, may provide a design solution that is neither economic nor practicable.

Equally the most conservative design assumptions for pipeline stabilization or pipeline embedment may similarly lead to design solutions that are neither economic nor practicable. This leads to an increasing requirement for pipeline engineers to understand more about the behaviour of the pipe-soil interface, to assess the risk inherent in a pipeline design solution. The requirement for pipeline engineers to more fully understand the pipe-soil interface is equally required where the seabed material classification does not fall into the clay or sand/gravel/rock categories.

2 2.1

General

In the design of pipeline supports, be they engineered or seabed, the initial assumption in pipeline design is that the pipeline moves, often a significant distance, and that the loads imparted by the pipeline on the support may significantly exceed the resistance that the seabed may provide. This is contrasted by traditional foundation design, where the design criteria consider that the foundation resists significant movement of the structure. This can lead to misunderstanding between pipeline and geotechnical engineers, where they are essentially talking different languages, but using the same words. Consider the example of a concrete mattress overlaid at the end of an offshore pipeline for protection. The design of the pipeline-mattress-seabed interface will see loads much greater than the pipeline-mattressseabed can resist. However, the mattress-seabed interface will have the functional requirement to be stable, under the action of the pipeline and hydrodynamic loads.

791 © 2011 by Taylor & Francis Group, LLC

PIPELINE RESPONSE

The ‘length’ of a pipe support, particularly an engineered support, may be short, even with regard to the pipeline diameter. For a pipeline supported on a seabed, the length of pipeline that is supported may vary from continuous to discrete short supports. The response of the pipeline may equally vary between long system effects or short local effects. Typically axial expansion (feed-in) / restraint is associated with a long length of pipeline. The support of pipelines at span shoulders and at the crowns of global buckles is typically a local restraint. In stability analyses, restraints such as secondary stabilization measures are generally a local response, where primary stabilization away from any local restraints is pipeline system behaviour. It is often required to consider the interaction of system and local behaviour, in particular for global buckling design.

The comparison of foundation and pipeline design for the example of lateral buckling is further discussed in White and Gaudin (2008), DNV RP-F110 (2007) and Bruton et al. (2008). For the purpose of this paper the implementation of geotechnical techniques into the consideration of pipeline response is split into three approaches: – a traditional friction factor approach, – an empirical approach, – the fundamental approach. In many ways these approaches all originate from an experimental/testing basis whether that is from in-situ testing, laboratory tests, centrifuge testing or full scale experiments. The difference is in how the idealised soil response experimental data is interfaced in to the (often idealised) pipeline response models. In the traditional friction factor approach, the results of the geotechnical testing and/or analysis are interpreted into a Coulomb, bi-linear (or multi-linear) friction factor response. In the empirical approach the geotechnical results may be interpreted to account for some of the more non-linear aspects of pipe-soil behaviour which may include passive resistance, cyclic, and load history effects. In the fundamental approach the geotechnical analysis interprets the test results to address the underlying mechanics of the pipe-soil behaviour. 2.2

Pipeline response – system or local

Offshore pipelines are long with regard to their diameter, but also with regard to the wavelength of ‘global’ response modes.

© 2011 by Taylor & Francis Group, LLC

Implementation of traditional friction factor approach

In many ways a section on the traditional friction factor approach in the proceedings of the symposium on frontiers in offshore geotechnics could seem misplaced. Many authors have highlighted the influence of loading history, loading rate and hydrodynamic interactions, on the pipe-soil interaction behaviour, and that the pipe-soil resistance is often not directly proportional to the vertical pipeline reaction. There are many reasons for the traditional friction factor approach remaining a core tool in the pipeline design ‘toolbox’:

Pipeline surveys

While pipeline route surveys are normally carried out along the total length of a planned pipeline route, it is worth noting that a pipeline route survey is not necessarily a geotechnical pipeline route survey. Guidance on geotechnical investigations for marine pipelines is further discussed in OSIF (2004), http://sig.sut.org.uk/sutosig.htm. Part of the functional requirements for these route surveys should include the assessment of seabed features including geophysical, geotechnical and geological information. The assessment of geotechnical properties typically includes information from geological databases, insitu testing, laboratory testing, and classification testing. Characteristic soil parameters – ‘design’ values – are subsequently defined. Where the design of the pipeline requires the use of model or full scale testing for the calibration or definition of input parameters for specific pipe-soil models, the collection and preservation of a substantial quantity of soil may be required. Where the design of the pipeline system requires characteristic soil parameters for soils that are not classified ‘clay’ or ‘sand’, there is limited guidance in the present range of published offshore pipeline rules and guidelines. 2.3

2.4

– The friction factor approach remains an appropriate assumption for many pipeline responses. – At conceptual/front end phases of pipeline design, it is often required to make major assumptions in lieu of little or no seabed survey information. – It is common practice in design and research to consider idealised cases, where simplifying assumptions are utilized to focus on a particular aspect of the response. – Before an experimental approach may be specified it is often necessary to bound the expected input ranges using idealised cases. – The friction factor approach can be used to quickly implement geotechnical experimental results into standard pipeline engineering tools, leading to an experimentally calibrated friction factor approach. It is often assumed that friction coefficients for pipelines and soils are historically available. However there are frequently differences in pipeline coatings, even within the same ‘type’ of coating. The typical concrete coating application method in SE Asia differs from the application method applied in European coating plants, leading to significant differences in concrete coating roughness, which may have an effect on the pipe-coating/soil interface friction.

792

Figure 2. UWAPIPE vertical displacement versus reaction. Figure 1. Typical Verley model force-displacement curve.

Where the interface friction has a significant impact on the pipeline design, laboratory shear tests are recommended for the pipe-coating/interfacing materials. 2.5 Implementation of geotechnical techniques – empirical sand model The Verley & Sotberg (1992) pipe-soil interaction model for sandy soils predicts the development of pipe penetration into the soil and the associated soil resistance that may be mobilized against horizontal pipe displacements (Figure 1). Zeitoun et al. (2009) describe the implementation of the Verley & Sotberg (1992) soil response model into the J P Kenny Simulator package that uses ABAQUS as the finite element engine. The pipe-soil model has been implemented as a FORTRAN subroutine, and could be equally extended to include the Verley and Lund (1995) soil response model for clay soils. An empirical model for partially drained conditions, which often occur in silts or calcareous sands, would require further testing to develop and implement. 2.6 Implementation of geotechnical techniques fundamental approach – plasticity model / UWAPIPE The plasticity models set out in Zhang et al. (2001, 2002a, and 2002b) provide a fundamental approach to simulate the mechanics of pipe-soil behaviour. Zhang and Erbrich (2005) present applications of the plasticity models demonstrating how the ‘traditional friction factor approach’does not capture the effect of the loading path on the minimum friction coefficient. Tian & Cassidy (2008) present the application of the plasticity models in terminology consistent with structural analysis frameworks and finite element models, referring to this model as the UWAPIPE model. The UWAPIPE model, developed by, and licensed from the University of Western Australia, has been implemented into the J P Kenny Simulator package. Examples responses of UWAPIPE, including the associated failure envelopes in vertical-horizontal (V-H) load space are shown in Figure 2 and Figure 3. © 2011 by Taylor & Francis Group, LLC

Figure 3. Pipe positions and failure envelopes.

2.7

Comparison of pipeline dynamic lateral stability response to Coulomb, Verley and Sotberg (1992) and UWAPIPE models

The key issues that are faced when designing for pipeline stability include; the acceptance criteria, the hydrodynamic loads acting on the pipeline, the pipe structural response and the pipe-soil response including resistance, liquefaction and scour as discussed in Zeitoun et al (2008). For these key issues, idealised cases are often utilized to define loads and resistances. The applicability of these idealised cases has previously been a subject of study and debate (e.g. Palmer (1996) and Teh et al. (2006)). The recently published DnV RP-F109 considers that soil resistance consists of a Coulomb component and a passive penetration contribution and that hydrodynamic load reduction may be attributed to pipe embedment and for a permeable seabed. DnV PR-F109 acknowledges the issues with calcareous soils, liquefaction and scour, and that at present a recognized model for incorporating scour and liquefaction into a generalised pipeline stability design methodology remains a subject for ongoing research.

793

increased embedment, and therefore lateral resistance, compared to the original Zhang / UWAPIPE model, which is for drained conditions. The development of further fundamental understanding of pipeline-soil-fluid behaviour is ongoing and includes research such as the STABLEpipe JIP, as discussed in Griffiths et al. (2010), and partially drained plasticity models. In addition to the resistance of the soil, these developments consider the effect of the fluid/pipe on the soil properties and the effect of the pipe on the soil resistance. Figure 4. Lateral movement comparison.

2.8

Figure 5. Vertical displacement comparison.

From a pipeline lateral stability design consideration, how do the present methods of modelling pipe-soil resistance compare when considering the response of the pipeline system to a particular hydrodynamic load history and these different resistance models? For the purpose of comparing the pipeline dynamic lateral stability response to Coulomb, Verley and Sotberg (1992) and UWAPIPE models, no load reduction, scour or liquefaction effects are considered. The input parameters for the Verley & Sotberg (1992) and UWAPIPE models are as detailed in Verley & Sotberg (1992) and Tian & Cassidy (2008), and these are compared to case using a Coulomb friction coefficient of 0.8. In Figure 4, the lateral movement for the Coulomb model and the UWAPIPE model are similar, with a significantly lower lateral displacement for the Verley and Sotberg (1992) model. This implies that the input parameters utilized in the UWAPIPE model are comparable to a Coulomb friction of around 0.8, for the pipe size and SG under consideration. The embedment obtained in the UWAPIPE and Verley and Sotberg (1992) models are significant, with a lower embedment predicted from the UWAPIPE model (Figure 5). It is noted that Cathie et al. (2005) suggest that the centrifuge test data used to develop the Zhang / UWAPIPE model gives a lower rate of increase in equivalent friction factor with increasing embedment than the test data used to develop the Verley & Sotberg model. Recent centrifuge studies to assess a partially drained response of a pipeline in calcareous soil have led to a proposed partially drained model which offers

794 © 2011 by Taylor & Francis Group, LLC

Implementation of geotechnical techniques – modelling of pipelay induced embedment and its effect on lateral resistance

Observations of pipelay-induced embedment have been documented by Lund (2000) and Westgate et al. (2009). The observations show that the embedment can be variable based on the activity of the pipelay vessel, from planned activities such as pipe end laydown, and unplanned events that reduce the lay rate. Bruton et al. (2008) and DnV OS-F110 discuss pipesoil interaction responses that influence the design of pipelines susceptible to lateral buckling, which include pipe embedment due to installation effects, its effect on lateral resistance, and the effect of axial resistance on feed-in. Cheuk and White (2008), and White and Gaudin (2008) present how centrifuge modelling techniques can be utilised to study the behaviour of the soil under dynamic pipelay load histories. This is further developed by Wang et al. (2009) who present numerical simulations of dynamic pipelay embedment, andYu and Konuk (2007) who utilize a continuum approach to simulate lateral buckling. In an operational pipeline, the as-found pipeline embedment may be considered when determining the most likely range of lateral resistances, taking into account any potential for seabed mobility. In the design phase this is less straightforward. A lateral buckle is a local feature, and the feed-in to the buckle is a global pipeline effect. Pipeline design should consider an approach that addresses both of these global and local effects. The combination of lower, upper bound and best estimate frictions for a range of potential pipelay embedments can be overly conservative. A probabilistic approach may be used to describe the potential range of pipeline responses such as the method presented in Rathbone et al. (2008). Soil responses from no pipelay effect to a full pipelay effect may therefore be addressed, capturing the range of potential lateral resistances. Engineering judgment is still required in the definition of the axial and lateral resistance probability distribution functions. For the lateral buckling case these should consider that the axial resistance is a global feature and that the lateral resistance is a local feature and the statistical distributions may be limited by the no embedment case and a maximum predicted embedment case. Engineering judgment is also required to interpret the results of the probabilistic assessment, or to communicate the risk of

Figure 6. Typical pipelay lateral friction distributions.

Figure 8. Superspan soil testing in a geotechnical centrifuge.

Figure 7. Jansz Superspan illustration Equid (2008).

buckling from the design to the operational / integrity management phases of the project. There are many challenges that are the subject of ongoing research in lateral buckling design. These include the consideration of pipe-soil model complexity (force resultant, empirical, plasticity, continuum), the requirements for, and variability of, soil testing, and how to describe appropriate idealised pipeline design cases. Even when engineered interfaces such as rock or vertical/lateral trigger systems are utilized, the small risk of an unplanned lateral buckle on the seabed may govern the pipeline design.

geotechnical and pipeline design team worked closely together to determine the interaction of pipeline loads with the potential geotechnical response at the span shoulders. In order to further understand the behaviour of the soil at the shoulder of the Jansz scarp span, detailed assessments including centrifuge testing and modelling program were undertaken (Figure 8), to determine and validate the pipe-soil interaction model to be used in the analysis of the pipeline response. The pipe-soil interaction model considered the axial and lateral soil behaviour, and in particular the effect of both the pipelay loads and operational loads on the potential embedment and vertical stiffness. The pipe-soil interaction model was interpreted into input parameters for the pipeline structural model, considering the potential for various pipelay and operational loads, and the potential range in soil conditions determined by the in-situ geotechnical tests. The resulting structural model was extensively utilized to assess the sensitivity of the base case design solution to perturbation of the constructed trench/pipe profile/location, pipelay loads and back tensions. 3

2.9 Implementation of geotechnical techniques – modelling of Jansz scarp crossing span support The challenges of the Jansz development and the crossing of the scarp are discussed in Equid (2008). Figure 7 shows an impression of the scarp and resulting pipeline span. DnV OS F105 requires that the modelling of pipe-soil interaction is considered in the detailed evaluation of free spanning pipelines, preferably by means of geotechnical tests on undisturbed soil samples, in particular where the in-situ soils differ from the typical geotechnical values given in DnV RP-F105. Borehole samples of the seabed were collected and utilized for laboratory testing. The over consolidated calcareous soil found at the Jansz scarp has very different characteristics to typical deep water soils. The

795 © 2011 by Taylor & Francis Group, LLC

SUMMARY AND CONCLUSIONS

Projects have significant benefits to gain through sound management of the risks associated pipeline geotechnical design. Mitigation strategies need to address both technical and design execution risks. Key factors that need to be addressed include: – the availability of geotechnical data, – the availability of pipe-soil models that are applicable for the soil classification, – the availability of pipeline design tools that can utilise the pipe-soil model, – the maturity of the pipeline design. A key design execution risk is management of the interaction between pipeline designers and geotechnical specialists. It must be recognized that the technical

languages used by respective parties are different and poor communication can lead to incomplete and inconsistent overall design outcomes. Recognition and mitigation of this risk is an essential element of the pipe-soil interaction design problem. A well managed design effort including clear recognition and management of technical and communication risks can lead to significant benefits from both design integrity and project execution perspectives. REFERENCES Bruton, D. A. S., White, D., Carr M. and Cheuk, J. C. Y. (2008) Pipe-Soil Interaction During Lateral Buckling and Pipeline Walking – The SAFEBUCK JIP, Proc. Offshore Technology Conference, OTC 19589, Houston, USA. Cathie, D.N. Jaeck, C., Ballard, J.C. and Wintgens J.F.:(2005) Pipeline Geotechnics: State of the Art, Proc. Int. Symp. on Frontiers in Offshore Geotech. (ISFOG) Perth, Australia. Cheuk, C. Y. and White, D. J. (2008). Centrifuge modelling of pipe penetration due to dynamic lay effects. Proc. Int. Conf. on Offshore Mech. & Arctic Engng., OMAE200857923. Estoril, Portugal. Det Norske Veritas. Offshore Standard DNV OS-F101 Submarine Pipeline Systems, October 2007. Det Norske Veritas, Recommended Practice DNV RP-F105 Free Spanning Pipelines, February 2006. Det Norske Veritas, Recommended Prac. DNV RP-F109 On-bottom Stability Design of Submarine Pipelines April 2009. Det Norske Veritas, Recommended Practice DNV RP-F110 Global Buckling of Submarine Pipelines Structural design due to high temperature/pressure., October 2007. Equid, D. (2008) Challenges of the Jansz deepwater tieback. Proc. Deep Off. Tech. Conf. (Asia-Pacific), Perth, Australia. Griffiths, T. J., White, D. J. and Cheng, L., (2010) Progress in investigating pipe-soil-fluid interaction: The STABLEpipe JIP. Proc. 20th Int. Offshore and Polar Engng. Conf., ISOPE 2010-TPC-790, Beijing, China. Lund, K. M. (2000) “Effect of increase in Pipeline Soil Penetration From Installation” Proc. Int. Conf. on Offshore Mech. & Arctic Engng. OMAE2000/PIPE-5047. Offshore Soil Investigation Forum (OSIF) 2004. Guidance notes on geotechnical investigation of marine pipelines. Rev 03. http://sig.sut.org.uk/sutosig.htm. Palmer, A.C., (1996) A Flaw in the Conventional Approach to Stability Design of Marine Pipelines. Proc. Conf. on Offshore Pipeline Technology, Amsterdam. Rathbone, A. Hakim, M. A. Cumming, G. and Tørnes, K., (2008) Reliability of lateral buckling formation from planned and unplanned buckle sites. Proc. Int. Conf. on Offshore Mech. & Arctic Engng. OMAE2008-57300, Estoril, Portugal.

© 2011 by Taylor & Francis Group, LLC

Teh, T.C., Palmer, A.C., Bolton, M.D. and Damgaard, J.S., (2006) Stability of submarine pipelines on liquefied seabed. ASCE J. of Waterway, Port and Coastal Engineering. Tian,Y. and Cassidy, M. (2008). Explicit and implicit integration algorithms for an elastoplastic pipe-soil interaction macroelement model. Proc. Int. Conf. Offshore Mech. & Arctic Engng. OMAE2008-57237, Estoril, Portugal. Verley, R.L.P. and Sotberg, T. (1992) A soil resistance model for pipelines placed on sandy soil. Proc. Int. Conf. Offshore Mech. & Arctic Engng. Alberta, Canada. Verley, R.L.P. and Lund, K. M. (1995) A soil resistance model for pipelines placed on clay soils. Proc. Int. Conf. Offshore Mech. & Arctic Engng Copenhagen, Denmark. Wang, D., White, D.J. and Randolph, M.F. (2009), Numerical simulations of dynamic embedment during pipe laying on soft clay. Proc. 28th Int. Conf. on Offshore Mechanics and Arctic Engineering, OMAE2009-79199, Honolulu, Hawaii. Westgate, Z. J., White, D. J. and Randolph, M. F. (2009), Video observations of dynamic embedment during pipelaying in soft clay”, Proc. 28th Int. Conf. on Offshore Mechanics and Arctic Eng., OMAE2009–79814, Honolulu, Hawaii. White D.J. and Gaudin, C. (2008) Simulation of seabed pipesoil interaction using geotechnical centrifuge modelling. Proc. Deep Offsh. Tech. Conf. (Asia-Pacific) Perth, Aust. Yu, S. and Konuk, I. (2007) Continuum FE modelling of lateral buckling Proc. Offshore Tech. Conf., Houston USA. Paper OTC18934. Zhang, J., Stewart, D. P. and Randolph, M. F. (2001). Centrifuge modelling of drained behaviour for pipelines shallowly embedded in calcareous sand. Int. J. Physical Modelling in Geotechnics 1, 25–39. Zhang, J., Stewart, D. P. and Randolph, M. F. (2002a). Kinematic hardening model for pipeline-soil interaction under various loading conditions Int. J. Geomech., 2(4), 419–446. Zhang, J., Stewart, D. P. and Randolph, M. F. (2002b). Modelling of shallowly embedded offshore pipelines in calcareous sand. ASCE J. Geotech. & Geoenv. Eng., 128(5), 363–371. Zhang J., and Erbrich C.T.: (2005) Stability design of untrenched pipelines – geotechnical aspects, Proc. Int. Symp. on Frontiers in Offsh. Geotech. (ISFOG) Perth, Aust. Zeitoun, H.O., Tørnes, K., Cumming, G., and Brankoviæ, M. (2008) Pipeline stability – State of the Art. Proc. 27th Int. Conf. on Offshore Mech. & Arctic Eng., OMAE2008– 57284, Estoril, Portugal. Zeitoun, H.O., Tørnes, K., Li, J., Wong, S., Brevet, R., and Willcocks, J. (2009). Advanced dynamic stability analysis, Proc. 28th Int. Conf. on Offshore Mechanics and Arctic Eng., OMAE2009–79778, Honolulu, Hawaii.

796

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Lateral soil resistance to an untrenched pipeline under the action of ocean currents F.P. Gao Key Laboratory for Hydrodynamics and Ocean Engineering, Institute of Mechanics, Chinese Academy of Sciences, Beijing, China

S.M. Yan Key Laboratory for Hydrodynamics and Ocean Engineering, Institute of Mechanics, Chinese Academy of Sciences, Beijing, China China Petroleum Pipeline Engineering Corporation, Langfang, China

E.Y. Zhang CNOOC Research Center, Beijing, China

Y.X. Wu Key Laboratory for Hydrodynamics and Ocean Engineering, Institute of Mechanics, Chinese Academy of Sciences, Beijing, China

X. Jia CNOOC Research Center, Beijing, China

ABSTRACT: A plane-strain finite element model is proposed to investigate the pipe-soil interaction mechanisms for the partially embedded pipe with two kinds of constraint conditions, i.e. freely-laid pipe and anti-rolling pipe. The numerical model is verified with updated mechanical-actuator experiments. The magnitude of lateralsoil-resistance coefficient for the examined anti-rolling pipes is much lager than that for the freely-laid pipes, indicating the end-constraint condition affects significantly the lateral stability of the untrenched pipeline in ocean currents. Parametric study indicates that the variation of lateral-soil-resistance coefficient with the dimensionless submerged weight of pipe is affected greatly by the internal friction angle of soil, pipe-soil friction coefficient, etc. 1

INTRODUCTION

To avoid the occurrence of pipeline on-bottom (lateral) instability, i.e. the breakout of the pipe from its original site, the seabed must provide enough soil resistance to balance the hydrodynamic loads upon the untrenched pipeline. For pipeline geotechnical engineers, one of the main concerns for pipeline on-bottom stability design is to properly predict the ultimate soil resistance in the severe ocean environments, and to further determine the thickness of coating layers based on nominal pipe weight (Det Norske Veritas 2007). In the past few decades, the pipe-soil interactions have attracted much interest from pipeline researchers and designers. Numerous experimental studies on wave-induced pipe instability have been carried out with 1g mechanical actuators (e.g., Wagner et al. 1987; Palmer et al. 1988), with centrifugal pipe-soil interaction tests on calcareous sand (e.g. Zhang et al. 2002), and with flume hydrodynamic simulations (e.g. Gao et al. 2003; Teh et al. 2003). Several empirical

© 2011 by Taylor & Francis Group, LLC

797

“pipe-soil” or “wave-pipe-soil” interaction models were developed to improve the conventional Coulomb friction theory. Some reviews on pipeline geotechnics and pipe-soil interactions have been made recently by Cathie et al. (2005), White and Randolph (2007), etc. Note that the aforementioned studies mainly focused on the pipeline on-bottom stability subjected to ocean waves. As the oil and gas exploitation moving into deeper waters, ocean current becomes the prevailing hydrodynamic load for on-bottom stability of submarine pipelines.Although the pipe on-bottom stability in currents seems less complicated than in waves, till now, the underlying physical mechanism has not been well revealed (Gao et al. 2007). To further explore the mechanism of pipeline onbottom stability in ocean currents, a plane-strain finite element model is proposed and verified with the mechanical-actuator tests. The ultimate lateral soil resistance to the untrenched pipes with two kinds of constraint conditions, i.e. freely-laid pipes and anti-rolling pipes, is investigated numerically.

Figure 1. Typical plane-strain finite element mesh (not in scale) and boundary conditions for pipe lateral stability analyses.

2

DEVELOPMENT OF A PLANE-STRAIN PIPE-SOIL INTERACTION MODEL

2.1 The finite element model 2.1.1 Finite element mesh and boundary conditions As the length of a submarine pipeline is much larger than its diameter, the pipeline lateral stability can be treated as a plane-strain problem. A plane-strain finite element model is proposed for simulating the breakout of the pipeline from its original site. The typical finite element mesh is illustrated in Figure 1. The boundary conditions are set as follows: (1) at the left and right boundary, no displacement in the x direction takes place; (2) the bottom boundary is fixed, i.e. the displacement and rotation are not permitted; (3) at the pipe-soil interface, the contact-pair algorithm provided in the ABAQUS software (Hibbitt, Karlsson and Sorensen Inc, 2006) is adopted to simulate the moving pipe along the deformable soil. The non-contact soil surface is treated as a free boundary. In the numerical modelling of the pipeline losing onbottom stability, it is crucial to properly describe the contact conditions between the pipe and the neighbouring soil. The pipe-soil friction is defined by the Penalty Function with the advantage that it guarantees the positive definiteness of sparse matrix in the calculation. In order to avoid large distortion of finite elements causing the calculation misconvergence, the self-adaptive mesh technology is employed. To obtain high calculation efficiency, the finite element mesh gets more refined at closer proximity to the pipe. Based on the results of a series of trial calculations, the width of the numerical model is set as 17.5D and the depth as 5D, and the pipeline is located at x = 7.5D (Dis the pipeline diameter), see Figure 1. 2.1.2 The end-constraint and the simulation of ocean current loading on the pipeline For a long-distance laid pipeline, the on-bottom stability of the pipeline at its separate sections is different. Due to the constraints from risers and the pipeline own anti-torsion rigidity, the pipeline movement is neither purely parallel nor purely rotational. As such, the following two end-constraint conditions are taken account in the present study: Case I: Anti-rolling pipe. © 2011 by Taylor & Francis Group, LLC

Pipeline’s rolling is restricted, but pipeline can move freely in horizontal and vertical directions; Case II: Freely-laid pipe. The pipe may rotate around its axis without any end constraint. When a pipeline is laid on the seabed under the action of ocean currents, there exists a dynamic balance between the submerged weight of the pipe, the hydrodynamics forces (including the horizontal drag force FD and the vertical lift force FL ), and the soil resistances. When the ultimate lateral soil resistance is not able to balance the horizontal drag force, the pipe would breakout from its original site, i.e. the lateral instability occurs. To efficiently simulate the ocean currents induced hydrodynamic loads upon a submarine pipe-line is crucial for evaluating pipeline lateral on-bottom stability. According to Morison’s equation, the horizontal and lift (vertical) components of the steady flow induced horizontal drag force and vertical lift force are expressed as FD = 0.5CD ρw DU 2 , FL = 0.5CL ρw DU 2 , respectively. Herein, CD is the drag coefficient, CL is the lift coefficient, is the mass density of water, U is the effective water particle velocity. The variations of the drag and lift coefficients, CD and CL , with the Reynolds number (Re) for various values of pipe surface roughness have been obtained by Jones (1978). The resultant hydrodynamics force upon the pipe is obliquely upwards with the inclination angle:

Referring to the experimental results by Jones (1978), the inclination angle (θ) is approximately between 530 − 570 . It is therefore reasonable to apply an inclined force in the θ direction to simulate the hydrodynamic loads on the pipe in steady ocean currents. Constitutive model for soils and the material properties The sandy soil under drained conditions can be essentially assumed to behave as an elastic c-φ material (e.g. Mohr-Coulomb or D-P material). The seabed soil is simulated with the well-known Drucker-Prager (D-P) elastoplasticity constitutive model. In the simulations, the parameters of soil are chosen as follows: Young’s modulus E = 0.18 MPa, Poisson’s ratio ν = 0.32, the cohesion c = 0, the buoyant unit weight of the soil γ
= 9.3 × 103 N/m3 , the values of soil internal friction angle φ are various for the parametric study in Section 3.1. As aforementioned, the pipe is treated as a rigid cylinder with outer diameter D = 0.15 m (same as the test pipes). The submerge weight of the pipe per meter (WS ) and the pipe-soil friction coefficient (µ) are various for parametric studies in Section 3. Due to that the stiffness of the steel pipeline with concrete cover is normally larger than that of the soil, the wall of the pipeline is regarded as a rigid cylinder in this finite element analysis.

798

2.1.3

Figure 2. Experimental setup for pipe lateral stability.

2.2 Verification of numerical model To verify the proposed numerical model, an updated experimental facility by employing the mechanicalactuator simulation method has recently been designed and constructed, as depicted in Fig. 2. The facility mainly consists of a sand box with glass wall, a mechanical-actuator, and the measurement system, etc.. In the sand box (2 m long, 0.5 m wide and 0.6 m deep), a saturated sand-bed with certain relative density can be prepared by employing the sand-raining technique. In the mechanical actuator system, a stepper motor was capable of generating inclined force onto the test pipe via a cable passing through a fixed pulley, for simulating steady currents induced drag force and lift force on the pipeline. Meanwhile, a lifter was used to adjust the inclination angle, which was maintained in the range of 530 ∼ 570 according to the above analyses. Figure 3(a) illustrates typical development of lateral soil resistance and the corresponding vertical pipesoil contact force for an anti-rolling pipe when losing lateral stability. With the increase of horizontal displacement (Sx ) during the pipe losing lateral stability, the horizontal lateral soil resistance (FH ) increases gradually to its maximum value (Fu = 0.10 kN/m) when the additional settlement is nearly fully developed according to the experimental observation. Meanwhile, the corresponding vertical pipe-soil contact force (WS − FH tanθ) decreases gradually to its minimum value (0.085 kN/m). The FEM numerical results match well with the test results. Figure 3(b) shows the numerical results of plastic deformation beneath the anti-rolling pipe while losing lateral stability. It is indicated that the shear band is distributed underneath the deformed soil layer; meanwhile, the soil just in front of the moving pipe upheaves obviously (see Figure 3b). The variation of ultimate lateral soil resistance (Fu ) with the vertical pipe-soil contact force (WS − Fu tanθ) is given in Figure 4, indicating the numerical and the experimental results are quite comparable. The ultimate lateral soil resistance increases linearly with the vertical pipe-soil contact force. The proposed FEM model is capable of predicting the lateral resistance for the untrenched pipeline on-bottom instability. © 2011 by Taylor & Francis Group, LLC

Figure 3. (a) Development of lateral soil resistance and the corresponding vertical pipe-soil contact force for an anti-rolling pipe when losing lateral stability: Comparison between numerical and experimental results; (b) Plastic deformation beneath the anti-rolling pipe while losing lateral stability (D = 0.15 m, µ = 0.7, WS = 0.225 kN/m, φ = 26.7◦ ).

Figure 4. Variation of ultimate lateral soil resistance with vertical pipe-soil contact force: Comparison between the numerical and the experimental results (D = 0.15 m, µ = 0.7, φ = 26.7◦ ).

799

3

NUMERICAL RESULTS AND ANALYSES

In the process of a pipeline losing lateral stability under the action of ocean currents, the soil plastic deformation beneath the untrenched pipeline may be

Figure 5. Plastic deformation beneath the pipe while losing lateral stability: (a) Anti-rolling pipe; (b) Freely-laid pipe (D = 0.15 m, Ws = 0.439 kN/m, µ = 0.7, φ = 20◦ ).

created due to the intensive pipe-soil interaction. Figure 5(a) and (b) illustrate the plastic strain in the proximity of an anti-rolling pipeline and that of a freely-laid pipeline, respectively. It is indicated that, the end-constraint condition has much influence on the distribution of the plastic strain zone in the soil. For the anti-rolling pipeline, an obvious shear strain band may be formed in the underlying soil layer, and soil upheave occurs in front of the moving pipeline (see Figure 5(a)). Nevertheless, for the freely-laid pipeline, the smaller plastic-strain zone is created just underneath the rolling pipeline (see Figure 5(b)). As discussed above, many factors influencing the pipe-soil interaction could be incorporated in the proposed finite element model. In the following sections, the effects of soil internal friction angle and the pipesoil friction coefficient on the on-bottom stability of the pipelines with two kinds of end-constraint will be further investigated numerically.

3.1

Figure 6. Lateral stability of anti-rolling pipes for various values of internal friction angle: (a) em /D vs. G; (b) η vs. G(D = 0.15m, µ = 0.7).

horizontal lateral soil resistance to the corresponding vertical pipe-soil contact force, i.e.

The commonly-used dimensionless submerged weight (G) of the pipe is

Effects of soil internal friction angle

For better understanding the pipe-soil interaction mechanism for on-bottom stability, a lateral-soilresistance coefficient (η) is proposed, whose physical meaning is the ratio of the ultimate value of the © 2011 by Taylor & Francis Group, LLC

800

where γ
is the buoyant unit weight of the saturated sand. Both the experimental and numerical results show that, in addition to the initial embedment due to selfweight of the pipe in the process of losing lateral stability, some additional settlement may be developed while the horizontal lateral soil resistance increases gradually to its maximum value. Figure 6(a) and (b) give the variation of maximum pipe settlement (em /D. with the dimensionless pipe submerged weight (G) and that of the corresponding lateral soil resistance coefficient (η) with

Figure 7. Lateral stability of freely-laid pipes for various values of internal friction angle: (a) em /D vs. G; (b) η vs. G (D = 0.15 m, µ = 0.7).

G for various values of soil internal frictional angles for the Case of anti-rolling pipeline with a given diameter (D = 0.15 m). The maximum pipeline settlements (em /D in the process of pipeline losing stability increase approximately linearly with the increase of G. For the same value of G, em /D increases with the decrease of soil internal friction angle, especially for the larger pipeline submerged weights (see Figure 6(a)). The lateral-soil-resistance coefficient (η) decreases gradually to a constant value with the increase of G. The effect of soil internal friction angle on η gets more significant with increasing submerged weight of the pipeline. Similarly, the variation of em /D with G and that of η with G for the case of freely-laid pipelines are given in Figure 7(a) and (b). Compared with the case of anti-rolling pipelines (see Figure 6), the relationships between em /D and G for the freely-laid pipes follow similar trends, but the maximum settlements are somewhat less in magnitude. Unlike the case of antirolling pipe, the effect of φ on the variation of η with G for the freely-laid pipeline is different, i.e. η decreases with the increase of φ for a fixed value of G (e.g.

Figure 8. Effects of pipe-soil friction coefficients on the lateral stability of anti-rolling pipes: (a) em /D vs. G; (b) η vs. G (D = 0.15 m, φ = 26.70 ).

G > 1.0, see Figure 7(b)). This may attribute to that the pipe settles shallower into the soil with bigger internal friction angle, and that the freely-laid pipe tends to roll away from its original site. Note that the range of η for the examined anti-rolling pipes is between 1.0 ∼ 2.0 (see Figure 6(b)), but that for the freely-laid pipes only between 0.2∼0.3 (see Figure 7(b)). Therefore, the end-constraints have significant influence on the lateral stability of the untrenched pipeline in ocean currents. 3.2

The submarine pipeline is usually constructed with concrete cover. As imagined, the pipe-soil friction coefficient may affect the lateral stability of the pipeline. Figure 8 (a) and (b) give the variation of em /D with G and that of η with G for various values of pipe-soil friction coefficient (µ), respectively. For the case of anti-rolling pipes, the increase of µ brings an increase of maximum settlement in the process of pipe losing lateral stability (see Figure 8(a)). The effect of

801 © 2011 by Taylor & Francis Group, LLC

Effects of pipe-soil friction coefficient

pipe-soil friction coefficient is more obvious for the smaller value of G. Its effect on the variation of η with G gets less with the decrease of µ (see Figure 8(b)). 4

CONCLUDING REMARKS

As the offshore oil & gas exploitation goes into deeper waters, ocean current becomes the prevailing hydrodynamic load for on-bottom stability of submarine pipelines. In this paper, a finite element model is proposed and verified with the updated mechanicalactuator experiments. Parametric study is made to investigate the pipe-soil interaction mechanism for the current-induced pipeline lateral instability. The following conclusions can be drawn: • The finite element model can effectively simulate

the behaviour of pipeline losing lateral stability in ocean currents under two end-constraint conditions, i.e. the anti-rolling pipe and the freely-laid pipe. The ultimate lateral soil resistance can be obtained from the load vs. displacement curve. • A lateral-soil-resistance coefficient (η) is presented for better understanding pipe-soil interaction mechanism. The value of η decreases gradually to a constant with the increase of G. The magnitude of η for the examined anti-rolling pipes is much lager than that for the freely-laid pipes, indicating the end-constraint condition affects significantly the lateral stability of the untrenched pipeline in ocean currents. • For a certain case of end constraint (anti-rolling or freely-laid), the variation of η with G is affected by various parameters, including soil internal friction angle, pipe-soil friction coefficient, etc. The effect of pipe-soil friction coefficient is more obvious for the smaller value of G. • When evaluating the capacity of lateral resistance, it would be beneficial to further examine and get correlation with the maximum pipeline penetration (including initial and additional settlement) and the development of the plastic-strain zone beneath the pipeline.

REFERENCES Cathie, D.N., Jaeck, C., Ballard J-C. & Wintgens, J-F., 2005. Pipeline geotechnics – state of the art. In: Frontiers in Offshore Geotechnics: ISFOG 2005. Eds: Gourvenec S. and Cassidy M., New York: Taylor & Francis, 95–114. Det Norske Veritas, 2007. On-Bottom Stability Design of Submarine Pipelines. Recommended Practice, DNV-RPF109. Gao, F.P., Gu, X.Y. and Jeng, D.S., 2003. Physical Modeling of Untrenched Submarine Pipeline Instability. Ocean Engineering, 30 (10): 1283–1304. (SCI, EI) Gao, F.P., Yan, S.M., Yang, B. and Wu, Y. X., 2007. Ocean currents induced pipeline lateral stability on sandy seabed. Journal of Engineering Mechanics,ASCE, 133(10): 10861092. Hibbitt, Karlsson and Sorensen Inc, 2006. ABAQUS Theory Manual, Version 6.5–1. Jones, W.T., 1978. On-bottom pipeline stability in steady water currents.Journal of Petroleum Technology, Vol. 30, 475–484. Palmer, A. C., Steenfelt, J. S. and Jacobsen, V., 1988. Lateral resistance of marine pipelines on sand. Proceedings of 20th Annual Offshore Technology Conference, OTC 5853, 399–408. Teh, T.C. and Palmer, A.C. and Damgaard, J.S., 2003. Experimental study of marine pipelines on unstable and liquefied seabed. Coastal Engineering, 50, 1–2: 1–17. Wagner, D. A., Murff, J. D., Brennodden, H., and Sveggen, O., 1987. Pipe-soil interaction model. Proceedings of Nineteenth Annual Offshore Technology Conference, OTC 5504, 181–190. White, D.J. and Randolph, M.F., 2007. Seabed Characterisation and Models for Pipeline-Soil Interaction. Proceedings of the Seventeenth International Offshore and Polar Engineering Conference, Lisbon, 758–769. Zhang, J., Stewart, D. P., Randolph, M. F., 2002. Modeling of shallowly embedded offshore pipelines in calcareous sand. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Vol. 128, 363–371.

ACKNOWLEDGEMENTS The work reported herein is financially supported by Knowledge Innovation Program of the Chinese Academy of Sciences (No. KJCX2-YW-L02) and China National S&T Major Project (No. 2008ZX05026-005).

© 2011 by Taylor & Francis Group, LLC

802

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Vertical cyclic testing of model steel catenary riser at large scale T.E. Langford & V.M. Meyer Norwegian Geotechnical Institute, Norway

ABSTRACT: Deepwater offshore developments often use steel catenary risers to carry oil and gas between subsea facilities and floating platforms or vessels. Multidisciplinary research on steel catenary risers has investigated the interaction between these long cylindrical structural elements and the often soft and compressible seabed. The seabed is typically included within a structural model as an equivalent vertical stiffness, or hyperbolic curve, which may be based upon model riser tests or numerical simulations. Although these models may be sufficiently complex to capture the non-linearity of a single load phase, the vertical loading itself is variable throughout the lifetime of a riser. A large scale model has been subjected to vertical cyclic loading to demonstrate how set-up of a high plasticity soil leads to large regains in strength and stiffness, likely to be the result of thixotropy. Details of the test performance and results are presented, together with a discussion of the differences between the tests and other recent work.

1

INTRODUCTION

2.2

Offshore oil and gas developments in deepwater locations often use Steel Catenary Risers (SCRs) to link seabed facilities with floating platforms. The interaction of SCRs with typically soft seabed soils is an important research area in offshore geotechnics. Recent work (e.g. Hodder et al., 2009) investigated the effects of reconsolidation between cyclic phases of vertical loading. Such work relates to locations where seasonal storms are prevalent, with relatively benign loading conditions in between, and have been modeled in the centrifuge using kaolin. The tests described herein permit comparison with large-scale 1g model tests in high plasticity West African clay.

Soil properties

The soil for the current test program was high plasticity West African clay with natural water content around 150%. The liquid and plastic limits of the clay were measured to be 160% and 60% respectively, giving a plasticity index IP = 100%. The clay was consolidated using a vacuum applied within a rubber membrane at a stress of 15 kPa. The typical clay depth during testing was around 1.25 D. Mini T-bar (25 mm × 120 mm) penetration tests were performed immediately prior to the riser model tests in order to evaluate the shear strength profile. The T-bar tests were performed on the submerged soil at a penetration rate of 2 cm/s. The undrained shear strength, su , has been interpreted from the T-bar resistance using a factor NT-bar = 10.5. Figure 1 shows the

2 TEST PROGRAM 2.1 Test apparatus The pipe/riser model test apparatus at NGI uses a biaxial rig with hydraulic actuators and 50 kN maximum load capacity. The tests are executed using an MTS FlexTest SE control system and MOOG servo valves. For the tests presented herein, the model was subjected to vertical loading only using the vertical actuator. Further details of the apparatus are described by Langford et al. (2007). The test bin has plan dimensions of 3.6 × 1.75 m and has space for 6 footprints, 3 of which were used for the testing presented herein. The test riser section has a length of 1300 mm and diameter of 174 mm. It is finished with a roughened polypropylene coating. The base of the pipe is fitted with filter-capped holes which are in turn linked to pore-pressure transducers.

© 2011 by Taylor & Francis Group, LLC

803

Figure 1. Undrained shear strength from T-bar tests.

Table 1.

Summary of tests.

Penetration (z/D) Time between cyclic phases Test Max Min (days) 2-1 0.5 2-2 0.5 2-3 0.5 ∗

0.0 0.0 0.0

1 (36∗ ) 0.1 16–18

Additional cyclic phase after 36 days. Figure 2. Initial penetration to 0.5 D.

interpreted shear strength profiles, together with socalled ‘corrected’ profiles which are based upon the average of the measurements. This correction is based upon a non-linear increase in T-bar until it reaches a steady-state value (White & Randolph, 2007). Different interface coefficients are used ranging from α = 0.0 (smooth) to α = 1.0 (rough) to define the penetration factor Np for smaller penetrations according to Aubeny et al. (2005). In summary, su is inferred to be approximately 5 kPa at the surface, increasing to around 7 kPa at the base of the test bin. 2.3 Test parameters Each test footprint was subjected to 3 cyclic episodes, each comprising 20 cycles of vertical loading. Penetration to 0.25 D was performed using a constant displacement rate of 0.5 mm/s. Subsequent vertical cyclic tests were performed to consistent maximum and minimum embedments of 0.5 D and 0.0 D respectively. Cyclic penetration was performed according to a displacement controlled sine function with a cyclic period of 350 seconds (0.0029 Hz) giving an average displacement rate of 0.5 mm/s. The test program is summarized in Table 1. The delay between cyclic phases was 0.1 to 18 days for which the pipe was fixed at 0.25 D (i.e. above the surface of the trench). These delays allowed for set-up of the soil before retesting. Following completion of the third cyclic phase for Test 2-3, the rig was shifted back to the position of Test 2-1 and an additional cyclic phase was performed after a longer delay of 36 days. Successive cyclic test phases are given suffices A, B and C (and D for Test 2-1) within the paper. Pore pressures were measured at 3 positions along the pipe invert: U1 and U3 were at either end of the model whilst U2 was at the midpoint. 3 3.1

RESULTS AND INTERPRETATION Initial penetration phase

The initial penetration of the model riser into the soil is shown in Figure 2 and comprises monotonic loading to 0.25 D and further penetration to 0.5 D as part of the first cyclic phase. There was a small delay between these two phases which caused a drop in load at constant penetration, followed by an increase in load due to the short set-up period.

Figure 3. Vertical cyclic penetration for 1st cyclic phase.

A reduction in test rate towards maximum penetration occurs with the sine-time displacement loading function. This gives a corresponding reduction in soil resistance which can be explained by the change in soil strength with rate of strain. Test results show the contribution from soil resistance, qs , where the buoyancy component, qb , has been subtracted from the total resistance, qt , as suggested by Hodder et al. (2009). 3.2

Plots of soil resistance versus normalised penetration are shown on Figure 3 for the first cyclic phase of each test footprint (2-1-A, 2-2-A and 2-3-A). The results show the same response in each case whereby the soil resistance for penetration and extraction reduces for each cycle. It can also be seen that the soil appears to detach from the pipe and re-penetrates at a normalized penetration of around 0.35 to 0.45 D, increasing during the course of each test phase. The soil resistance for the 2nd cycle is around 65% of that for the 1st cycle, and reduces to as low as 10% of the original value after 20 cycles. The second and third cyclic phases for Test 2-1 (2-2-B and 2-2-C) are shown on Figure 4. The trench depth remains around 0.40 to 0.45 D for these phases, and the range of measured resistance during each phase is much smaller than for the first phase where the clay was originally intact. Similar results were obtained for Tests 2-1 and 2-3 respectively.

804 © 2011 by Taylor & Francis Group, LLC

Cyclic penetration

Table 2.

Summary of soil resistance values.

Test / phase

Time (days)

2-1-A 2-1-B 2-1-C 2-1-D 2-2-A 2-2-B 2-2-C 2-3-A 2-3-B 2-3-C

–

Soil resistance, qs (kPa)

Figure 4. Vertical cyclic penetration for 3 cyclic phases of Test 2-2; scale on y-axis for left hand plot is 4 times larger than for middle and right hand plots.

1 1 36 – 0.1 0.1 – 18 16

Cycle N

Cycle N + 19

qs,N+1 / q∗s,N

17.8 2.5 2.2 3.9 17.8 4.6 3.4 17.7 5.2 3.7

1.9 1.5 1.4 1.2 4.4 3.2 2.2 3.8 2.6 1.8

– 1.32 1.45 2.79 – 1.06 1.05 – 1.38 1.44

∗

Where ‘N’ is last cycle of previous phase and ‘N + 1’ is first cycle of current phase.

Figure 5. Change in pore pressure for 3 cyclic phases of Test 2-2; scale on y-axis for left hand plot is 5 times larger than for middle and right hand plots. Figure 7. Definition of normalised secant stiffness, Ksec .

response’ after 10 cycles. In contrast, the results from the current tests show that the soil resistance does not stabilize even after 20 cycles. The reason for this difference is unclear. From earlier T-bar testing the sensitivity (St ) of the soil is judged to be between 2 and 3 (Langford & Aubeny, 2008b). However, the observed pipe model test behavior suggests that water may be becoming entrained within the soil body during tests, thereby causing further reduction of the resistance. Table 2 summarises values of soil resistance for the first and last cycle of each test phase. The peak soil resistance for the first cycle is around 18 kPa for all three footprints, which suggests a homogeneous clay bed. However, the value after 20 cycles varies between 1.9 and 4.4 kPa, where the lowest ‘remoulded’ strength is given for the footprint closest to the tank wall where boundary effects may have some influence.

Figure 6. Change in peak soil resistance per cycle.

Figure 5 shows the measured change in pore pressure from Test 2-2. These data are not currently used in the interpretation, however general agreement between the changes in soil resistance and pore pressure are evident from the tests. Figure 6 shows the change in soil resistance per cycle for the three cyclic phases in each test. The Figure has been arranged so that the test with the shortest set-up period (2-2, 0.1 days) is on the left and that with longest set-up period (2-3, 16–18 days) is on the right. Hodder et al. (2009) presented results in kaolin where the soil resistance reached a ‘steady remoulded

3.3

The method used to derive the normalised cyclic stiffness, Ksec , is shown in Figure 7, and may be defined by the following equation:

805 © 2011 by Taylor & Francis Group, LLC

Normalized secant stiffness

Figure 8. Normalised secant stiffness, Ksec , versus normalised extraction, z/D, for Test 2-1; scale on y-axis for left hand plot is 10 times larger than for middle and right hand plots.

Figure 10. Normalised secant stiffness, Ksec , at z/D = 0.025 versus cycle number.

Figure 9. Normalised secant stiffness, Ksec , versus normalised extraction, z/D, for the last cycle of each test (N = 20).

where Vs is the change in unit pipe-soil load (kN/m), z is the change in displacement (m) and qs,initial is the initial pipe bearing capacity (kN/m2 ). The normalised secant stiffness has been evaluated for each cycle at different normalized penetrations. The resulting values for the different phases of Test 2-1 are shown on Figure 8. The results show that the normalised secant stiffness is highly non-linear when compared to the normalised penetration. Furthermore, Ksec drops rapidly during the first cyclic phase before reaching a range which is more typical of the values for the 2 subsequent cyclic phases. Figure 9 shows contours of Ksec from the different tests, as derived from the last cycle of each phase. There is a general trend of reduction in Ksec during the 3 phases for each test; however the curves plot remarkably close together. The range in cyclic secant stiffness between the different tests is similar to that for the soil resistance shown previously. Figure 10 shows the change in Ksec per cycle for the three cyclic phases in each test, based on a normalized extraction of z/D = 0.025. There is an increase in Ksec after each set-up period followed by a trend of general decrease throughout each cyclic episode. This trend is the opposite to that presented by Hodder et al.

© 2011 by Taylor & Francis Group, LLC

806

Figure 11. Comparison of results for Test 2-3 with data given by Hodder et al. (2009).

(2009) for a test series in kaolin where there was also no contact stress between the riser model and soil during the set-up periods. Figure 11 shows an example of the results from Hodder et al. (2009) where Ksec at the end of each phase increased from one phase to the next. Inspection of the graphs for soil resistance versus normalised displacement reveals the same trend, where the measured resistance at the end of one phase is greater than that preceding it. Since the remoulded strength is linked to the water content, it is therefore reasonable to assume that the water content has decreased during the test series. This could be due to the properties of the kaolin or stress regime within the centrifuge. It should be noted that the set-up factor for kaolin differs from natural clays. As reported by Andersen & Jostad (2004), kaolin is not thixotropic, but it may give higher setup than natural clays because the effective stress after redistribution of pore pressures is higher and because of faster global pore pressure dissipation. For the current tests, once the soil has been remoulded, and possibly experienced an increase in water content, the stresses near the seabed are much lower than those used for the initial consolidation, and the general trend of decrease soil resistance and normalised secant stiffness is not unexpected. If many more phases were performed, or if the number of cycles

Table 3.

Summary of normalised secant stiffness values. Normalised secant stiffness, Ksec (−)

Test / phase

Time (days)

2-1-A 2-1-B 2-1-C 2-1-D 2-2-A 2-2-B 2-2-C 2-3-A 2-3-B 2-3-C

– 1 1 36 – 0.1 0.1 – 18 16

Cycle N

Cycle N + 19

Ksec,N+1 / K∗sec,N

57.7 8.1 7.2 11.6 59.1 12.1 9.5 54.1 13.9 10.6

5.7 4.8 4.7 3.7 11.5 9.2 5.9 9.9 6.9 5.5

– 1.42 1.51 2.50 – 1.05 1.03 – 1.40 1.53

Figure 13. Effect of time delay between cyclic phases on qs and Ksec ; ratio of values from the 1st cycle of phase p to values from 20th cycle of phase p.

∗

Where ‘N’ is last cycle of previous phase and ‘N + 1’ is first cycle of current phase.

Figure 12. Effect of time delay between cyclic phases on qs and Ksec ; ratio of values from the 1st cycle of phase p+1 to values from 20th cycle of phase p.

was increased, it is possible that values may stabilize. Different results may be expected where there is a contact stress between the riser and seabed during the set-up periods. Table 3 is similar in format to Table 2, but rather presents values of Ksec at the start and end of each cyclic phase, giving the increase due to set-up in each case.

The figure shows a clear trend of increasing soil resistance and normalised secant stiffness with time, which is as expected.There are some issues that require further attention. For instance, the values for 1 day setup are very similar to those for 16-18 days set-up. The additional phase for Test 2-1 (2-1-D) included a longer 36 day set-up period. The associated increase in soil resistance and normalised secant stiffness is greater than would be expected from the trends for the other test phases. If the location for Test 2-1 is affected by the tank boundaries, it is possible that the horizontal and mean stresses have been increased by the deformed soil pushing on the tank walls, thereby giving a higher degree of set-up. This effect may have been less relevant for the 1 day set-up periods for phases 2-1-B and 2-1-C, since these would have been dominated by thixotropy. Nonetheless, this phenomenon will be explored during future tests. Figure 13 presents the results in a different way, where the relative decrease in soil resistance or normalised secant stiffness is plotted versus the corresponding time delay for set-up. This ratio can be considered analogous with the development of sensitivity in the soil. For reference, the ratios based on the soil resistance for the first cyclic phase in each case vary from around 4 to 9, using the values in Table 2. The corresponding ratios for Ksec range from 5 to 10. Once again, the results for the additional test phase at 2-1 (2-1-D) for 36 days set-up seem to lie above the trend for the other tests. This anomaly warrants further investigation at a later stage. 3.5

3.4 Time delay effects The time delay between cyclic phases allows set-up of the soil. For the current tests, since the set-up periods are relatively short, the set-up is likely to be dominated by thixotropy, which can be significant for high plasticity soils. Set-up due to changes in effective stress may be more relevant for longer set-up periods. Figure 12 presents the results from Tables 2 and 3 in terms of set-up. Also shown are typical thixotropy test results for high plasticity clays (Ip = 75 to 90%) taken from Andersen & Jostad (2004).

© 2011 by Taylor & Francis Group, LLC

807

Comparison with earlier work

Langford & Aubeny (2008a) present results from vertical displacement-controlled tests with set-up periods in between cyclic phases. There were also some important differences between the tests: • The displacement range was much smaller (0.01 D

instead of 0.25 D) • The set-up periods were limited between 1 and

4 hours • The riser model was kept in contact with the soil

during the set-up periods

• The soil consolidation stress was lower, giving a

secant stiffness. This increase seems to be related to the length of set-up period in a manner analogous to thixotropy in high plasticity clays.

lower shear strength profile Nonetheless, similar trends were obtained as for the current testing. The soil resistance dropped significantly throughout the test series. The magnitude of this decrease was far greater than that shown for the current tests, although this could be due to the very small range of displacement. However, there were also significant relative increases in measured resistance after the short set-up periods. Furthermore, the values of normalised secant stiffness at the end of the cyclic phases were very similar in spite of the decrease in soil resistance. Load-controlled tests from the same test bin also indicated significant set-up of the soil between cyclic test phases, in this case resulting in a reduction of permanent displacement per cycle. 4

CONCLUSIONS

Large scale vertical cyclic riser model tests were performed to investigate the performance of a high plasticity West African clay. The tests included set-up periods of different durations to allow the remoulded soil to regain strength between cyclic test phases. The tests showed the following trends • There is good agreement between the load-

displacement curves for the initial penetration phase of each test, which suggests a homogeneous soil profile. • The cyclic test phases reveal incomplete stabilization of the soil resistance, which differs to the trend shown by recent centrifuge tests with kaolin where stabilization occurred after 10 cycles. Incomplete stabilization suggests that water may have become entrained within the soil body during the pipe model tests. • There is a general trend for decrease in soil resistance and normalised secant stiffness from the first cyclic episode to the subsequent episodes. • The set-up periods of between 0.1 and 36 days allow a partial increase of soil resistance and normalised

© 2011 by Taylor & Francis Group, LLC

ACKNOWLEDGEMENTS The testing presented herein was partially funded by the Norwegian Research Council. The authors are grateful for help from colleagues in the laboratory and workshop at NGI, Oslo. Chuck Aubeny from Texas A&M provided helpful advice during planning of the test program and subsequent interpretation of results. REFERENCES Andersen, K.H. & Jostad, H.P. 2004. Shear strength along inside of suction anchor skirt wall in clay. OTC16844, Offshore Technology Conference, 3–6 May 2004, Houston, Texas, USA. Richardson: OTC. Hodder, M., White, D.L. & Cassidy, M. 2009. Effect of remoulding and reconsolidation on the touchdown stiffness of a steel catenary riser. OTC19871, Offshore Technology Conference, 4–7 May 2009, Houston, Texas, USA. Richardson: OTC. Aubeny, C.P., Shi, H. & Murff, J.D. 2005. Collapse loads for a steel cylinder embedded in trench in cohesive soil. Int. J. Geomechanics, ASCE 5(4): 320–325. Langford, T.E., Dyvik, R. & Cleave, R. 2007. Offshore pipeline and riser geotechnical model testing: practice and interpretation. Proc. Conf. on Offshore Mechanics and Arctic Engineering (OMAE), San Diego, California, USA. New York: ASME International. Langford, T.E. & Aubeny, C.P., 2008a. Model tests for steel catenary riser in marine clay. OTC19495, Offshore Technology Conference, 5–8 May 2008, Houston, Texas, USA. Richardson: OTC. Langford, T.E. & Aubeny, C.P., 2008b. Large scale soilriser model testing on high plasticity clay. Proc. 18th Int. Offshore & Polar Engineering Conference (ISOPE), Vancouver, Canada, Vol. 2, pp. 80–86. Cupertino: ISOPE. White D.J. & Randolph M.F., 2007. Seabed characterisation and models for pipeline-soil interaction, Proc. 17th Int. Offshore & Polar Engineering Conference (ISOPE), Lisbon, Portugal. Cupertino: ISOPE.

808

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Kupe gas project pipeline – optimisation of discrete rock berm design shore approach B.L. Larsson Technip Oceania Pty. Ltd., Perth, WA, Australia

ABSTRACT: The Kupe Gas Project, located 30 km offshore the NW Taranaki Peninsula of the New Zealand North island consists of a Wellhead Platform, 30 km of subsea 12

pipeline and a 6

service umbilical connected to the onshore processing plant and export pipeline. This paper describes the optimisation of the near shore approach rock dumping engineering design and operations executed to provide a long term solution for the rigid pipeline and unarmoured umbilical crossing of an extensive and culturally sensitive boulder field in very shallow water conditions. The use of discrete rock berms constructed at regular predetermined intervals provides a robust solution for crossing the boulder zone. By controlling the elevation of the ocean floor, the risks to product arising from point loading, excessive free-spans and long term erosion and stability problems have been managed. 1

INTRODUCTION

1.2

1.1 Kupe Gas Project The Kupe Gas Project consists of an unmanned Wellhead Platform, 30 km of subsea 12.75

pipeline and a 6

service umbilical with an onshore processing plant, located in the Taranaki geological basin on the West side of the North Island of New Zealand. Origin Energy Resources (Kupe) Limited is the nominated operator for the Joint Venture partners Origin Energy, Genesis, New Zealand Oil & Gas and Mitsui. The project, commissioned in March this year, will provide 15% of the domestic gas consumption of New Zealand. The Engineering, Procurement, Installation and Commissioning contract was executed by the Technip-Origin Energy Alliance, see Figure 1.

Scope of paper

The emphasis of this paper is the near shore crossing discrete rock berm design and construction.The design provided the Project with a technically and economically feasible solution for the near shore approach and crossing of a 2 km wide boulder field. 2 2.1

SUBSEA TRANSPORTATION SYSTEM Rigid export pipeline

The majority of the 12.75

(323.9 mm) line pipe is DNV 450 MPa with 22.2 mm Wall Thickness (WT) and 2.2 mm 3LPP coating. For the HDD crossing the WT increased to 25.4 mm and for the boulder field section, the coating increased to 6 mm. 2.2

Umbilical

The unarmoured umbilical is designed and supplied by Duco – a Technip umbilical manufacturing company. The service umbilical includes MEG and chemical injection lines, 3-phase 33 kV power and fibre optics sheathed in HDPE plastic. 3

Figure 1. Kupe field. © 2011 by Taylor & Francis Group, LLC

NEAR SHORE APPROACH

The subsea HDD shore crossing exits 2 km from the shoreline in sand at a water depth of 14 m (LAT). The near shore approach was categorised as the route alignment from the HDD subsea exit to the end of an extensive boulder field. Here, the pipeline and umbilical traverses a rock boulder zone in a water depth between 14 m and 22 m over a length of approximately 2 km.

809

Figure 3. Shoreline boulders.

Figure 2. The Taranaki coastline.

Table 1.

Table 2.

Metocean data.

Description

Water Depth U∗s U∗cwd Hs Hmax

Soil As-built data. Location

Unit –

1YR KP0

1YR KP27

100YR KP0

100YR KP27

m m/s m/s m m

34.0 1.09 0.44 6.38 11.86

14.0 1.67 28.0 4.71 8.75

34.0 1.88 0.76 9.17 17.05

14.0 2.25 0.42 7.6 13.14∗∗

Description

From-KP To-KP Soil description

Grey sands

0

14.4

Sand/Peat

14.4

17.6

Boulder field 25.6

28.0

∗

( ) Note 1: Significant velocity (Us ) and Current velocity (Ucwd ) taken at 1.0 m Above Sea bed (ASB). (∗∗ ) Note 2: The back-analysed maximum wave height derived from the statistical data is capped at 12.2 m due to seabed bathymetry capping Hmax in more shallow waters.

4 WEATHER Situated on the South side of the Taranaki peninsula the field is exposed to the weather arriving from the Southern Ocean, and subject to severe (winter) storms, see Figure 2 and Table 1. 5

GEOLOGY

The shoreline is formed by 40 m high cliffs, and subject to an average yearly erosion rate of 1 m. The seabed is generally covered by boulders resting on top of an underlying siltstone base over the 4 kms closest to the shore, with only isolated areas of sand with less rock boulders. The majority of the boulders are 200–400 mm, but boulders up to 2.5 m in diameter are present. Boulders are generally made up of andesite,formed from lahar lava flows and erosion of the coastline, see Figure 3. 5.1

Dark grey fine to course sand. Generally very loose to dense to very dense within the trenching depth. Grey sand as per above. Lenses of peat and silt. High undrained shear strength >180 kPa generally of limited thickness. Heavy black (Iron) sands Underlying siltstone Substrata. Isolated patches of superimposed sand. The majority of the boulders are 200–400 mm ranging up to approx. 2.5 m in diameter.

HDD exit (2 km offshore) to 34 m water depth at the WHP (30 km offshore), see Table 2. Cone penetration tests performed along the route alignment indicated soil strength of <2 MPa. The near shore approach seabed was further characterised in terms of extent of boulder cover as an area percentage. Significant rock boulders were charted. The pipeline traverses over 2 km through a boulder field with the soil substrata subject to: – 70% of area covered by 0–1 m boulders – 20% of area covered by 1–2 m boulders – 10% of area covered by >2 m boulders

Geotechnical data

Offshore the boulder field section, the seabed is predominantly sand and silty-sand, with isolated patches of dense sand and peat. The seabed is relatively flat with a gradual decline from 14 m water depth at the © 2011 by Taylor & Francis Group, LLC

6

LOCAL CONSENT

The Alliance had a clear objective to keep an open dialogue and work together with the local population in

810

the Taranaki region, accounting for regional employment, financial and environmental considerations. Attention was given to how the ancestral owners of the land, the Iwi people commonly known as Maori of New Zealand and their ethnic, beliefs and religion could be respected without excessive intrusion by the Project. As the rock boulders are considered to be the ancestors for the local Iwi people, the extended HDD shore crossing in combination with the rock dumping solution allowed the Alliance to install the pipeline production system with a minimum of disturbance. 7

DESIGN CRITERIA

7.1 Route alignment The final route alignment accounts for optimised alignment vs. dominant design weather direction, depth of boulder field and ensured that the HDD subsea exited in an area of sand.

In short all the excavation options were subject to: – A constraint on availability of suitable equipment; – A risk of excavation works not being completed within one summer causing delay to the offshore campaign; – A potential for extensive scour creating free-spans and a risk of boulders being displaced by storms in an uncontrolled way upon the pipe or umbilical product. 8.2 Infill methods To avoid excavation, the alternative was to use infill methods, dominated either by pre-lay mattresses or pre-pipeline installation rock dumping options. Due to the extent of the boulder field, the prelay mattress option was considered not to be technically, nor financially viable for the project. 9

ROCK BERM DESIGN CONCEPT

9.1 Design approach 7.2 Boulder field requirements The near shore boulder field represents a significant risk to the pipeline and umbilical during installation and throughout field life, e.g. point loading, mechanical damage, deformation and over stress of product around rock outcrop, increased risk of free-spans development etc. 7.3 Environmental loading The offshore bathymetry of the Kupe field is quite complex, with a dominant S-SW long period swell semi-aligned with the pipeline with potential for a more wind-driven wave loading approaching as a quartering sea. The tidal current is orientated almost perpendicular to the pipeline due to strong currents through the Cook Sound in W-E direction. 8

NEAR SHORE APPROACH SOLUTIONS

A number of options were evaluated before the Project decided in favour of the discrete rock berms. The various options can be grouped into either clearance or infill methodology. 8.1 Clearance methods Initially a number of excavation options were assessed, and included barge and surface excavator spreads, suction-hopper dredger, excavator modified for subsea use or blasting. The blasting alternative was dismissed quickly, as it would have had the largest environmental impact and would be inconsistent with the objectives of the Alliance of mutual respect for the Iwi people’s beliefs and religion. © 2011 by Taylor & Francis Group, LLC

Typically, in the past, rock dumping has been performed as a continuous rock berm. Either as a prepipelay operation to provide a more level seabed or for stabilisation purposes post-pipelay to provide protection against scour and related free-spanning, stability in storm conditions and protection against fishing and trawling gear. Based on the technical, economical and sociological/ethnical assessments and interfaces, the Project decided to proceed with a near shore approach based on rock dumping. The construction of a series of discrete rock berms effectively addresses a number of constraints including: – Less disturbance of the seabed and the Iwi ancestral rocks; – The elevated and globally level seafloor, i.e. well defined allowable difference in height between each adjacent rock berm, reduce the required pre lay and post lay rectification of free-span; – Height control of rock dumping is less critical with discrete berms, whereas only the shorter alignment of a rock berm providing the actual support for the pipeline is in contact with the product; – Known dynamic free-span allows the structural integrity including fatigue issues to be assessed and addressed during the detailed engineering phase ahead of pipeline installation; – Product seabed stability improves since a spanning pipeline generates minimal lift and hence less mattress weight requirements; – Less pipeline/umbilical length subject to abrasion; – Lower overall cost. 9.2 Detailed design Rather than implementing an extensive hydro dynamic physical modelling, the Project has worked closely with the design consultant Coastal Engineering Solutions based in Melbourne, Australia, with the

811

Offshore Engineering Division from Technip’s Specialist Centre in Aberdeen, UK, reviewing the results. The winter between pipelay installation and rock berm construction allowed the rock berm materials to consolidate prior to pipelay. Rather than the actual rock berm design itself, the allowable mattress spacing and pipeline free-span defines the individual distance between two adjacent rock berms. The detailed design included: – Rock stability assessment; – Identifying the rock sizing and grading curves (D15 / D85); – Assessment of discrete rock berm vs. a continuous rock berm including stabilisation methodology; – Assessment of implications for pipeline upheaval and/or lateral buckling; – Hydrocarbon flow assurance; – Pipeline design for dynamically loaded pipeline; – Umbilical design incorporating dynamic loading; – Fatigue analysis for pipeline & umbilical; – Fishing and trawling gear interaction; – Rock quality specification; – Rock quantity assessment.

rock particles were subject to an abrasion test and a requirement for cubical, compact unit-rock shape. Not more than 10% of the rock particles were allowed to have an elongated, brick like shape. During construction, samples were taken from every 10th truck leaving the quarry by employing the Queen Mary and Westfield (QMW) abrasion test, the elongation check was performed by visual monitoring and recording at discrete intervals.

10.4

With the rock berms being fully immersed in seawater there were limits set on absorption of water for both individual stones and a sample average. Even though not subject to freeze and thaw cycles, the chemical composition in combination with the absorption was controlled to reduce the risk of excessive erosive deterioration potential due to chemical reactions over the long term, generally related to (smectite) clay minerals.

11 10

ROCK SPECIFICATION

To fulfil the design criteria, there were also specific requirements on the rock aggregate itself. These criteria arise from International and local New Zealand/ Australian Standards, industry research findings, design specific criteria and Technip Group experience. The combined criteria reflect the requirements of the detailed design and define the quality of the rock aggregate in terms of geometric shape, chemical composition, density and gradation etc.

DISCRETE ROCK BERM IMPLICATIONS

The decision to construct discrete rock berms also imposed additional engineering requirements for the design of the production system. Arising from the use of discrete rock berms, as opposed to continuous support, additional engineering should include assessment of:

10.1 Rock aggregate density The minimum density of the rock aggregate particles accounts for the 100YR RP wave load condition. The unit-rock particle size vs. specific gravity needs to be assessed carefully, as it quickly impacts the construction methods and rock aggregate size. 10.2 Edginess To improve stability and unit-rock particle interlocking, the rock aggregate is an angular quarried and crushed rock with a certain edginess, rather than a smooth rounded shape of each rock itself. By implementing a series of selected mesh sizes and percentages from four (4) defined more narrow spanning ranges, the overall rock gradation ensured a rather steep gradation with a certain minimum of larger rocks to ensure the overall stability if subject to the design 100YR RP wave condition. 10.3 Shape To maintain the long term integrity of the rock berms, in addition to the basic shape criteria and density, the © 2011 by Taylor & Francis Group, LLC

Chemical composition

812

– Risk for excessive point loading at contact points exerted either during the lay, or long term field life; – Predefinition of criteria for pipelay installation including layback and contact loads.Attention to the mattress design including exit/entry of the product at either edge of a stabilisation mattress, and in turn mattress installation landing velocity limitation; – Use of static design criteria is required such that if subject to hydro-dynamic loading or other lateral loading the system is not moved beyond the elevated rock berm support; – Pipeline and umbilical lay tolerances to account for lay tolerances and temporary stabilisation. Width of rock berm to accommodate mattresses; – Pipeline potential to move or oscillate, could lead to both local coating damage, with increased risk of local corrosion or fatigue damage, with the potential to cause a pipeline / umbilical failure; – Shape and construction of the stabilisation mattresses vs. rock aggregate shape support taking into account the geometry of the product. Consider coating thickness increase; – Structural integrity criteria, including fatigue considerations, for pipeline and umbilical when used for a dynamic application; – Piggy-back of light-weight product umbilical to pipeline, allows increased predefined free-spans. Design to account for strap/saddle spacing vs. dynamic loading including fatigue and local point

– The created predefined free-spans results in pipeline and umbilical product being exposed to dynamic loading and fatigue issues; – By securing the pipeline with concrete stability mattresses on top of the discrete rock berms and using the pipeline as carrier pipe for the umbilical, the long term structural integrity is maintained for the whole production system; – Rock unit shape and static stability measures accounts for pipe cross section to prevent local buckling or coating damage during installation and permanent design alignment; – Design of the stabilisation mattresses and other stabilisation measures shall reflect the geometry and spacing of the discrete rock berms; – Reduced pre-pipelay span rectification with rock berms constructed within predefined height tolerance; – Less pipeline/umbilical areas subject to abrasion; – Pipeline stability is improved since a spanning pipeline generates minimal lift and hence mattress weight requirements are reduced;

Figure 4. Rock berm route alignment.

Figure 5. Rock berm cross section.

Particular to the discrete rock berm design, it is concluded that:

loading. Here, the rock berm distance was determined by pipeline allowable free-span accounting for VIV, fatigue and ultimate structural integrity.

12

– The use of discrete rock berms provides a robust solution for crossing boulder zones and provides system stability. By controlling the elevation of the ocean floor, the risk of point loading, excessive freespans and long term erosion and stability problems for the pipe can be managed; – Discrete rock berms allow increased control of project cost and construction duration; – The solution provides a predefined and controlled free-span situation with reduced post pipelay rectification; – Careful design of rock-unit size and gradation vs. inherent specific gravity for rock material that can be sourced close to site is essential for viability of solution; – Toughness, abrasion resistance and durability requirements of the rock aggregate to reflect site environment and design life; – Height control of rock dumping is less critical with discrete individual berms, than for a continuous rock berm.

ROCK BERM CONSTRUCTION

The rock is installed as discrete rock berms at intervals coinciding with the permissible free-spans. The berms are oriented perpendicular to the alignment route with a 6 m × 2.5 m stability mattress at each rock berm. The rock berm flat top surface is sized for lay tolerances and mattress size, see Figure 4 and Figure 5. A total of 86 rock berms were constructed using a side dump vessel with survey capability allowing monitoring, for practical purposes, down to ±10 cm of rock berm level and geometry. During the construction of the rock berms, particular attention was paid to: – Limit sea state for rock dumping activities; – Transversally offset rock berm widened to maintain lay corridor at top of rock berm; – Control of rock berm top level to reduce need of post-lay free-span rectification.

ACKNOWLEDGEMENTS 13

CONCLUSION

A careful review of installation criteria and associated weather forecasting services available was carried out. For the installation duration, temporary stabilisation measures were assessed. Particular aspects for the discrete rock berm design and interaction with pipeline & umbilical systems that can be highlighted are:

Technip and the Alliance would like to direct a special thank you to Peter Riedel of Coastal Engineering Solutions who provided the detailed design, and Technip OED Aberdeen, in particular Alasdair Maconochie and John Oliphant for the geotechnical advice during the project engineering phase.

– The pipeline expansion is relatively unaffected by the rock berm support at the cold-end of the production system; © 2011 by Taylor & Francis Group, LLC

813

PREFERENCES, SYMBOLS AND UNITS – 3LPP: 3 Layer Poly-Propylene – EHU: Electro Hydraulic Umbilical

– – – – – – – – –

HDD: Horizontal Directional Drill WHP: Wellhead Platform KP: Kilometre Point Lahar: Volcanic type of mudflow or landslide composed of pyroclastic material and water LAT: Lowest Astronomical Tide OD: Outer Diameter RP: Return Period WT: Wall Thickness YR: Year

REFERENCES ABS. 2006. Guide for Building and Classing Subsea Pipeline Systems. Houston: American Bureau of Shipping API 1104. 2005. Welding of Pipelines and Related Facilities, 20th Edition. Washington D.C.: American Petroleum Institute API-RP-1111. 1999. Design, Construction, Operation and Maintenance of Offshore Hydrocarbon Pipelines, 3rd Edition.Washington D.C.: American Petroleum Institute API Specification 5L. 2004. Specification for Line Pipe, 43rd Edition. Washington D.C.: American Petroleum Institute ASME Section VIIl Div 2. 2004. Rules for Construction of a Pressure Vessel. New York: American Society of Mechanical Engineers ASME B16.5. 2003. Pipe Flanges and Flange Fittings. New York: American Society of Mechanical Engineers Brown, G. et al. 2004. Reliability Based Assessment of Minimum Reelable Wall Thickness for Reeling, Proceedings of International Pipeline Conference, IPC04-0733. Calgary: American Society of Mechanical Engineers BS 7910. 2005. Guide to Methods for Assessing the Acceptability of Flaws in Metallic Structures. London: British Standard Institution DNV OS-F101. 2000. Submarine Pipeline Systems. Oslo: Det Norske Veritas DNV CN 30.5. 2000. Environmental Conditions. Oslo: Det Norske Veritas

© 2011 by Taylor & Francis Group, LLC

814

DNV RP-E305. 1988. Recommend Practice for On-Bottom Stability of Pipelines. Oslo: Det Norske Veritas DNV RP-F111. 2006. Interference between Trawlgear and Pipelines. Oslo: Det Norske Veritas DNV RP-F105. 2006. Free Spanning Pipelines. Oslo: Det Norske Veritas DNV RP-F107. 2002. Protection of Pipelines. Oslo: Det Norske Veritas Kupe Project. 2005. Final Report for the Geophysical and Geotechnical Pipeline Route Survey, South Taranaki Bight, New Zealand, Document Number P0071 Rev 0. New Plymouth: Kupe Project HR Wallingford. 2005. Kupe Metocean Design Criteria, Kupe Field, New Zealand, Assessment of Pipeline Exposure Risk, Technical Note EBR4048/01, Document Number 5510-02. Wallingford: HR Wallingford Metocean Solutions. 2006. Wave and current Design values for the Kupe Pipeline Doc. No. DESIGN-DATA-01 Rev. B. New Plymouth: Metocean Solutions Metocean Solutions. 2006. Summer season waves and currents for the Kupe pipeline route Doc. No. DESIGN-DATA02 Rev. B. New Plymouth: Metocean Solutions CIRIA. 1991. Manual on the use of rock in coastal engineering and shoreline engineering. London: Construction Industry Research and Information Association CIRIA. 1995. Special Publication 83, CUR Report 154 – Manual on the Use of Rock in Coastal and Shoreline Engineering. London: Construction Industry Research and Information Association Rance, P.J et al. 1970. The threshold of movement of loose material in oscillatory flow, Proc. 12th International Conf. on Coastal Engineering. Washington: International Conference on Coastal Engineering Soulsby, R. 1997. Dynamics of Marine Sands – A manual for practical applications. London: Thomas Telford Publications U.S.Army Coastal Engineering Research Centre. 1984. Shore Protection Manual. Washington D.C: Department of the Army, Corps of Engineers

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Model test studies on soil restraint to pipelines buried in sand R. Liu & S.W. Yan School of Civil Engineering, Tianjin University, Tianjin, China

J. Chu School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

ABSTRACT: The buckling of submarine pipelines may occur due to the action of axial soil frictional force caused by relative movement of soil and pipeline, which is induced by the thermal and internal pressure. The likelihood of occurrence of this buckling phenomenon is largely determined by the capability of the soil to resist pipeline movements. A series of large-scale model tests were carried out to facilitate the establishment of substantial data base for a variety of burial pipeline topologies. Results show that the soil resistance depends on the pipe diameters and the depth of cover. According to the uplift test results, the force-displacement topologies with smaller depth of cover are greatly different from those with larger depth of cover. The results of the lateral sliding and axial pull out tests show that the soil resistance initially increases before a peak value is reached and then keeps the same level. For the same covered depth, the lateral soil resistance is more than twice that for uplift.

1

INTRODUCTIONS

Vertical buckling of buried submarine pipeline caused by high temperature and pressure is an important issue endangering the safety and stability of pipeline (Nielsen et al. 1990, Guijt 1990, Liu et al. 2005). In-service hydrocarbons must be transported at high temperature and pressure to ease the flow and prevent solidification of the wax fraction. Thermal stress together with Poisson effect will cause the steel pipe to expand longitudinally. If such expansion is resisted, for example by frictional affects of the foundation soil over a kilometer or so of pipeline; compressive axial stress will be set up in the pipe-wall. When the value exceeds the constraint of foundation soil on the pipeline, sudden deformation will occur to release internal stress, which is similar to the sudden deformation of strut due to stability problems. The upheaval buckling may cause the pipeline destroyed suddenly, as a result, the effective prediction method and protection measure against this phenomenon are important parts in the design of buried submarine pipelines. From the analysis above, it can be known that under a given temperature and pressure design condition, the occurrence of buckling is largely determined on the soil resistance acting on the pipeline. However, the soil resistance depends on a good many factors, such as direction of pipeline movement, amplitude of pipeline buckling, burial depth of pipeline and soil properties. Unfortunately, there is no theoretical method which can determine the soil resistance accurately up to now. As a result, model tests and numerical analysis are both the primary means to research the interaction between the soil and pipelines. Many researchers have done © 2011 by Taylor & Francis Group, LLC

some researches in this area since the early eighties of the last century. The first published work in the field of pipe-soil interaction surfaced in 1981. To design the pipeline for lateral stability and determine the winch capacity for pulling the pipeline, Anand and Agarwal (1981) carried out small-scale model and large-scale prototype experimental studies to determine the frictional resistance between the concrete-coated pipes and the soil in the lateral as well as the longitudinal directions. Taylor et al. (1985; 1989) chose sand as the supporting medium in view of North Sea conditions and carried out the pull-out tests and axial friction tests. In paper Boer et al. (1986) described the results of full scale pull-out tests for a 2 m long test section of a 12.75 inch O.D. concrete coated pipe covered with gravel. Horizontal and vertical pull out tests for pipeline buried in sand and soft clay were performed by Friedmann in 1986. Schaminee et al. presented the results of a fullscale laboratory test program on the uplift and axial resistance of a 4 inch pipe embedded in cohesive or cohesionless soil in 1990. Finch (1999) confirmed that soft clay backfill can be effectively modeled as a frictional material when considering uplift resistance, and derivation of applicable axial friction factors for coated pipelines in sand. Schupp et al. (2006) described a plane strain pipe unburied tests in loose dry sand and initial small scale three-dimensional buckling tests. White et al. (2008) presented a limit equilibrium solution for the uplift resistance of pipes and plate anchors buried in sand based on the model tests. Some centrifuge model tests were conducted by Dickin (1994), Moradi and Craig (1998), Bransby et al. (2001, 2002), Palmer et al. (2003) and Cheuka et al. (2007) to assess

815

Figure 1.

Model test tank scheme and Pull-out Topology.

the uplift capacity pipelines. Based on the tests, the force-displacement relationships of the pipe for different embedment depths had been produced and a semi-empirical design formula had been proposed. However, there are great differences among the results calculated by these various formulas. The differences are owing to the following reasons. One is that the understandings of the soil failure mechanism are at variance, the other is that the soil properties are quite different in different region. This paper chooses fine sand as the soil medium in view of Bohai Gulf geotechnical conditions and carries out a series of large-scale model tests to facilitate the establishment of substantial data base for a variety of burial pipeline topologies. Pipe segments with diameter of 30 mm, 50 mm and 80 mm are used respectively. The pipe segments are buried in different depth-todiameter ratios between 1 and 9. The uplift, lateral and axial resistances are recorded during the tests.

2 2.1

MODEL TESTS Design of experiment model

The purpose of the tests was to obtain the forcedisplacement relationship under different test conditions. To do this, the pipe segments were pulled out in vertical, lateral and axial directions, buried at different depths. The test data could be applied to determine the soil resistance in upheaval analysis of submarine pipelines. The model test tank dimensions were 1 m × 1 m × 3 m, which was assembled by PVC plates. In order to pull out the pipeline in three different directions and eliminate end effects (Tran 1994), two “L” shape notches were slotted symmetrically on each side of the tank. Sliding retaining device was used to prevent the sand from flowing out of the hole with pipeline moving. An organic-glass-plate was installed on one side of the tank to facilitate the observation of soil deformation during the pipe moving. Photograph of the device are shown in Figure 1. Stainless steels pipe segments with diameter of 30 mm, 50 mm, 80 mm and length of 1200 mm were selected to simulate the real pipelines in the test. All the test pipe segments were sealed on ends and extend out of the tank with 100 mm for each side to eliminate the disturbance of pipe ends to soil during the pipe moving. The pipe segments were buried in different depth-to-diameter ratios between 1 and 9. To record © 2011 by Taylor & Francis Group, LLC

Figure 2. Silty sand grading curve.

the peak values of soil resistance during pipe moving, it required that the movement speed of the pipe should be smooth and slow enough and the force acting on the pipe should be steady and continuous. Therefore, the electromotor with a secondary reducer was adopted to control the force application. The rotate speed of the electromotor was controlled at 0.06 mm/s during the test. The force was transferred to the pipe by the steel rope with a group of dynamic and static pulleys. Figure 1 is the sketch of the force acting as pipe moves in different directions.

2.2

Soil sample preparation

Sand was chosen as the soil medium in view of Bohai Gulf conditions and a sieve analysis identified the requisite fine sand (ref. to Fig. 2). Dry condition testing was employed for convenience. According to sieve analysis results, the average particle size of sand used in the test was 0.248 mm, and its non-uniform coefficient was 3.16, the curvature coefficient was 0.95, which belongs to uniform fine sand according to the soil classification code. Laboratory data showed that the maximum and minimum dry density of the sand is 1.76 g/cm3 and 1.57 g/cm3 separately. The fine sand was compacted to a relative density (Dr) equal to 0.5, and the corresponding dry density was 1.66 g/cm3 . The natural repose angle of the sand was 32◦ and internal friction angle was 35◦ .

816

2.3

Data recording

Pipeline displacement and soil resistance were the main measurement data in the test. The displacement transducers were installed on both ends of the pipe in accordance with pipe movement direction. Electric pressure sensors were installed on the steel rope which transfers the force in the test. A KYOWA-PCD300A type dynamic data processor had been used. During the test, the displacement transducers, pressure sensor and dynamic data processor were all connected to a computer, which could record the displacement and force data simultaneously by special data processing software.

2.4 Tests procedure The weight of the pipes was measured before the test. Sand was filled into the test tank layer by layer. For each filling layer, sand was compacted to a target bulk density of 16.6 kN/m3 . As the filling depth reaches a required level, a horizontal trench was then excavated to 100 mm depth to the tank bottom where the pipe emplaced. During vertical pull-out test, firstly measured the pull-out force without covered sand and took it as initial value. Secondly, a trench was dug to the design burial depth with 30◦ sidewall. The pipe was laid and loose sand was filled without compaction. Finally, a vertical force was applied to the pipe and the pipe move upwards slowly and smoothly. The time path curve of vertical pull vs. displacement was recorded until the pull out force remains the same level. During lateral pull-out test, firstly, placed the lateral pull out setup. Secondly, a trench was dug to the design burial depth with 30◦ sidewall. The pipe was laid and loose sand was filled without compaction. Finally, applied a lateral pull-out force to the pipe and keep it moving.The lateral pull and the displacement of the pipe were recorded simultaneously until the force approached to a constant value. During axial pull out test, firstly, the axial pull out setup was placed on both ends of the pipe to realize changing the movement direction during the test. Secondly, a trench was dug to the design burial depth with 30◦ sidewall. The pipe was laid and loose sand was filled without compaction. Finally, the axial pull out force and the pipe displacement had been recorded in the process of the pipeline moving back and forth for two cycles. Diameters of the pipe were 30 mm, 50 mm, 80 mm and the depth-to-diameter ratios are between 1 and 9 in the test. The weights of the pipe were 2.8 kg, 6.3 kg and 13.45 kg, separately. For all three movement directions, 81 group tests had been carried out.

Figure 3. Vertical Pull-out Tests Results for H /D =2.

Figure 4. Vertical Pull-out Tests Results for H /D = 8.

Figure 5. Lateral Pull-out Tests Results for H /D = 2.

3 TEST RESULTS ANALYSIS 3.1 Influence of diameter to the soil resistance Averaged vertical pull-out characteristics are illustrated in Figures 3–4 for cases of H /D = 2, 8 and D = 30 mm, 50 mm and 80 mm respectively. The vertical force is denoted by Fv and pipe vertical displacement by Sv. Figures 3–4 indicate that the soil resistance follows a similar procedure for pipes of different diameters. The loci show that only small deformations are onset up to the maximum pullout state, deflection then increasing rapidly down the post maximum falling branch. The maximum pull-out values are indicative of the mechanical effect of pipeline diameter and burial depth. The larger diameter, the greater soil resistance can be reached and relatively the greater displacement is needed. In the case of H /D = 2, the pipe displacement corresponding with maximum soil resistance are 0.3 mm, 1.5 mm, 3.2 mm © 2011 by Taylor & Francis Group, LLC

for D = 30 mm, 50 mm and 80 mm respectively. In the case of H /D = 8, the pipe displacement corresponding with maximum soil resistance increase to 0.375 mm, 3.6 mm, 8.0 mm for D = 30 mm, 50 mm and 80 mm respectively. Averaged lateral pull-out characteristics are illustrated in Figure 5 and Figure 6 for cases of H /D = 2, 9 and D = 30 mm, 50 mm and 80 mm respectively. The lateral force is denoted by FL and pipe lateral displacement by SL . Figures 5–6 indicate that the soil resistance loci of lateral pull out tests are quite different from that of vertical pull out tests. Lateral test results show that the soil resistance initially increases before a peak value is reached and then keeps the same level. The larger the pipe diameter is, the greater the soil resistance reached. And such changes decrease with the pipe depth of cover increasing. For smaller depth of cover, the pipe displacement corresponding to the

817

Figure 6. Lateral Pull-out Tests Results for H /D = 9.

Figure 9. Vertical Pull-out Tests Results for D = 80 mm.

Figure 7. Axial friction force for D = 30 mm and H /D = 8. Figure 10. Lateral Pull-out Tests Results for D = 80 mm.

soil resistance initially increases before a peak value is reached and then keeps the same level. Comparing Figure 7 to Figure 8, it can be seen that the pipe displacement corresponding to the maximum soil resistance increases with the pipe diameter. Due to soil disturbances, the soil resistance in axial direction is reduced after reversed pull out tests. Under cyclic reversal, the soil resistance for the pipe with large diameter decreases faster than the pipe with small diameter. For example, the soil resistance after two times reciprocation decreases to 81 percent of that after the first pull-out test for the case of D = 30 mm. And the decrease ratio falls to 69 percent for the case of D = 80 mm.

Figure 8. Axial friction force for D = 80 mm and H /D = 8.

maximum soil resistance is approximately 0.1D for the pipe of different diameters, which is much larger then the one corresponding to the vertical maximum soil resistance. For smaller depth of cover, the lateral soil resistance is more than twice that for uplift. Since the depth of cover will enlarge this change of soil resistance, under thermal and internal pressure actions the vertical buckling is particularly of interest with respect to entrenched submarine pipelines. Pipelines in operation usually experience cyclic temperature changes which lead to pipeline expansive and shrink along the axial direction. In order to study the influence of pipeline extending and shrinking on the soil resistance, the pipe is pulled back and forth twice during the axial pull out test. The axial resistances before and after loading in such way are recorded as shown in Figures 7–8. The axial force is denoted by FA and pipe axial displacement by SA . Figures 7–8 indicate that the © 2011 by Taylor & Francis Group, LLC

3.2

Influence of burial depth to soil resistance

Averaged vertical and lateral pull-out characteristics are illustrated in Figure 9 and Figure 10 respectively. Figure 9 shows that the vertical soil resistance increases obviously with the pipe burial depth. However, there is no linear proportional relationship between the soil maximum resistance and depth-todiameter ratios. When the depth-to-diameter ratio is less than 5, the soil resistance initially increases rapidly before a peak value is reached and decreases to a residual level. When depth-to-diameter ratio is greater than 5, the soil resistance has no significant trend of decrease after its peak. Figure 10 illustrates that the lateral soil resistance initially increases before a peak value is reached and then keeps the same level for all depth-to-diameter ratio. The soil resistance increases

818

Figure 13. Soil failure modes for H /D = 2 and H /D = 8. Figure 11. Axial friction force for D = 50 mm and H /D = 2.

Figure 12. Axial friction force for D = 50 mm and H /D = 8.

significantly with pipe burial depth. For the same depth of cover, the lateral soil resistance is more than twice that for uplift. The axial pull-out characteristic is illustrated in Figures 11–12 for cases ofH /D = 2, 8 and D = 50 mm respectively. The results of the axial pull out tests show that the soil resistance initially increases before a peak value is reached and then keeps the same level with the pipe displacement developed. Comparing Figure 11 to Figure 12, it can be seen that the pipe displacement corresponding to the maximum soil resistance increases with the depth of cover. Due to soil disturbances, the soil resistance in axial direction was reduced after reversed pull out tests and the soil resistance decrease faster with larger depth of cover than the one with smaller depth of cover. For example, soil resistance reduces to 74 percent of its initial value for the case of H /D = 2 and reduces to 70% for the case of H /D = 8. 3.3 Soil failure mode analysis In order to observe soil deformation as pipe moving, dyed sands are laid between different backfill sand layers and black straight lines can be seen before test (ref. to Fig. 13). According to the uplift test results, the soil failure modes with smaller depth of cover are greatly different from those with larger covered depth. The smaller © 2011 by Taylor & Francis Group, LLC

cases show that the pipe moving up mobilizes the soil wedge above it extending to the surface and upheave the soil surface within a certain ranger over the pipe. This type of soil failure mechanism is called shallow failure. Figure 13a is a photo of soil deformation corresponding to the maximum soil resistance for the case of D = 50 mm and H /D = 2, which shows that black sign lines between the pipe and soil surface all curve with the pipe moving up. The influence region looks like a reverse trapezoid. The soil failure mechanism above the pipe belongs to shallow failure. With the burial depth increasing, the upheaval phenomenon of soil surface fades away. For the case of H /D > 5, the displacement on the soil surface is hardly observed and a very localized failure mechanism appears in close proximity to the moving pipe, which can be called larger failure. Figure 13b is a photo of soil deformation corresponding to the maximum soil resistance for the case of D = 30 mm and H /D = 8, which shows that movement of the pipeline just affect the region about five times of the pipeline diameter, and the shape of the affected region is like rectangular. Therefore, a larger failure occurred in soil over the pipeline. Based on the test data, H /D = 5 can be taken as transition from smaller to larger failure mechanisms for burial pipelines.

4

CONCLUSIONS

Model tests have been employed to facilitate the establishment of a substantial data base for a variety of pipeline-burial relationships. Pipe segments with diameter of 30 mm, 50 mm and 80 mm are used respectively. The pipe segments are buried in different depth-to-diameter ratios between 1 and 9. The uplift, lateral and axial resistances are recorded during the tests. Based on 81 tests, it can be concluded that: The vertical pull-out tests show that the soil resistance depends on the pipe diameter and depth of cover. The force-displacement relationships with small pipe diameter and shallow depth of cover are greatly different from those with large pipe diameter and deep depth of cover. In both cases, the soil resistance initially increases rapidly before a peak value is reached. In the former case, the soil resistance decreases to a residual level, whereas in the latter case the soil resistance remains the same level.

819

The results of the lateral sliding tests show that the soil resistance initially increases before a peak value is reached and then keeps the same level. The displacement corresponding to peak resistance increases with pipe diameter. For the same depth of cover, the lateral soil resistance is more than twice that for uplift and increase with pipe diameter, which indicates that in practice, buried pipeline usually occurs vertical buckling rather than lateral buckling under thermal and internal pressure action. The results of the axial pull-out tests also show that the soil resistance initially increases before a peak value is reached and then keeps the same level. The displacement corresponding to peak resistance increases with pipe diameter. The soil resistance decreases with cyclic reversal. For the conservation, the residual axial soil resistance should be used in pipeline upheaval analyzing. According to the uplift test results, the forcedisplacement relationships with smaller depth of cover are different from those with larger depth of cover. The soil deformation also depicted the different between the pipe with smaller covered depth and larger covered depth. Based on the test data, H /D = 5 can be taken as the limit to divide the soil failure mode into shallow failure and deep failure. ACKNOWLEDGMENTS The work described in this paper was funded by China National Natural Science Foundation (No. 40776055). REFERENCES Anand S. and Agarwal, S.L., 1981, Field and laboratory studies for evaluating submarine pipeline frictional resistance,Transactions ofASME, Journal of Energy Resources Technology, 103: 50–254. Boer S., Hulsbergen C.H., Richards, D.M. et al, 1986, Buckling considerations in the design of the gravel cover for a high temperature oil line, Proc. 18th OTC, Houston,Texas, May, 5294: 1–8. Bransby M.F., Brunning P., and Newson T.A., 2001, Numerical and centrifuge modelling of the upheaval resistance of buried pipelines Proceedings of the International Conference on Offshore Mechanics and Arctic Engineering, 4(6): 265–273. Bransby M.F., Newson T.A., and Brunning P., 2002, The upheaval capacity of pipelines in jetted clay backfill, International Journal of Offshore and Polar Engineering, 12(4): 280–287.

© 2011 by Taylor & Francis Group, LLC

Cheuk C.Y., Take W.A., Bolton M.D., Oliveira J.R.M.S., 2007, Soil restraint on buckling oil and gas pipelines buried in lumpy clay fill, Engineering Structures, 29: 973–982. Dickin, E. A., 1994, Uplift resistance of buried pipelines in sand,Soils and Found, 34(2): 41–48. Finch, M., 1999, Upheaval buckling and floatation of rigid pipelines: the influence of recent geotechnical research on the current state of the art, Proceedings of the Annual Offshore Technology Conference, 1(5): 27–43. Guijt, J., 1990, Upheaval buckling of offshore pipeline: overview and introduction, In proceedings of the 22nd Annual OTC, Houston, Texas, 4: 573–578. Liu R., Yan S.W. and Sun G.M., 2005, Improvement of the Method for Marine Pipeline UpheavalAnalysis under Thermal Stress£¬Journal of Tianjin University, 38(2): 124–128. Moradi, M., and Craig, W. H., 1998, Observation of upheaval buckling of buried pipelines, Centrifuge 98, Kimura, Kusakabe and Takemura(eds): 693–698. Nielsen, N.J.R., Lyngberg, B. and Pedersen, P.T., 1990, Upheaval buckling failures of insulated burial pipelinesa case story, In proceedings of the 22nd Annual OTC, Houston, Texas, 4: 581–592. Palmer A. C., White D. J., Baumgard A. J., et al. 2003, Uplift resistance of buried submarine pipelines: comparison between centrifuge modeling and full-scale tests, Geotechnique, 53(10); 877–883. Peng L.C., 1978,Stress Analysis Methods for Underground Pipe Lines: Part 2, Pipeline Industry: 65–75. Schaminee, P.E.L., Zorn, N.F. and Schotman, G.J.M., 1990, Soil response for pipeline upheaval buckling analyses: Full-scale laboratory tests and modeling. OTC 6486, 22nd annual otc, Houston, texas, may 7–10: 563–572. Schupp, J., Eacott, N., Byrne, B.W. et al, 2006, Pipeline unburial behaviour in loose sand, Proceedings of the International Conference on Offshore Mechanics and Arctic Engineering, 2006(6). Taylor, N., Richardson, D., and Gan, A.B., 1985, On submarine pipeline frictional characteristics in the presence of buckling, Proc. 4th International Symposium on Offshore Mechanics and Arctic Engineering, ASME, Dallas, Texas, February: 508–515. Taylor, N., Tran, V.C., and Richardson, D., 1989, Interface modeling for upheaval subsea pipeline buckling, In proceedings of 4th International Conference on Computational Methods and Experimental Measurements, Capri, Italy, Springer-Verlag, May: 269–282. Tran, V., 1994, Imperfect upheaval buckling of subsea pipelines,PhD Thesis, Sheffield Hallam University. White, D.J., Cheuk, C.Y. and Bolton, M.D., 2008, The uplift resistance of pipes and plate anchors buried in sand, Geotechnique, 58(12): 771–779.

820

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Pipe-soil interaction on clay with a variable shear strength profile D.R. Morrow & M.F. Bransby Civil Engineering Department, University of Dundee, Dundee, Angus, UK

ABSTRACT: Pipe-soil interaction is an important consideration in pipeline design for sub-sea oil and gas developments, particularly when pipeline burial is not undertaken as is commonly the case in a deep water setting. This paper presents the results of a suite of small strain finite difference analysis, undertaken to investigate the influence of shear strength gradient and a shear strength crust on pipeline penetration under vertical loading. Both the shear strength gradient and the presence of shear strength crusts were found to influence pipeline penetration resistance significantly. In particular, some geometries of shear strength crust resulted in a punchthrough mechanism significantly increasing pipeline penetration for a given vertical load, relative to an equivalent linear increasing shear strength gradient.

1

INTRODUCTION

1.1 Background Subsea pipelines fulfill a range of important functions in the development of offshore oil and gas fields. For example, infield pipelines are used to link wells and longer distance export pipelines are used to transport products ashore or to a central offshore facility. Ancillary pipelines may also be present, providing water or gas injection to the reservoir or transporting additives. Within a deepwater setting it is typical for these subsea pipelines to remain on the seabed through the course of their design life, increasing the importance of understanding pipe-soil interaction (Bruton et al., 2006; Perinet and Fraser, 2006). Pipelines in shallower water settings may also remain on the seabed if consideration of protection and other requirements indicates no need to undertake burial. Perhaps the earliest method of considering pipe-soil interaction, and the most commonly used in design, was the development of empirical methods based on the results of model testing (e.g. Brennodden et al. 1986; Verley and Lund, 1995). More recently, further work has been undertaken using this approach specifically for deep water developments, as described by Bruton et al. (2006), and utilizing centrifuge model testing (Hodder et al., 2008). Hill and Jacob (2008) also describe the migration of “model testing” to the field, or even to site specific testing. In addition to model testing based approaches, there has been progress in understanding pipe-soil interaction using analytical techniques, such as finite element analysis (Aubeny et al., 2005; Merifield et al., 2008) and upper bound plasticity calculations (Randolph and White, 2008). Useful information on failure mechanism has also been gained using image processing techniques in conjunction with centrifuge © 2011 by Taylor & Francis Group, LLC

model testing (Dingle et al., 2008). With literature largely confined to a flat seabed case some preliminary investigations into the effects of seabed slopes were undertaken by Morrow and Bransby (2009). Clay soils present in subsea development areas are likely to have increasing shear strength with depth, as a result of normal consolidation from self weight. This is discussed by Puech et al. (2005) for Gulf of Guinea soils and Yun et al (2006) for Gulf of Mexico soils. Increase in shear strength with depth may also be associated with various over-consolidation events, although these will not be specifically considered within the scope of this paper. An example of a somewhat usual shear strength profile that can be present in deepwater development areas is the presence of a near surface increased shear strength “crust” as described by Ehlers et al. (2005), Puech et al. (2005), and by Kuo and Bolton (2009). This type of shear strength profile may be significant for pipe-soil interaction and initial investigations will be presented here. Aubeny et al. (2005) undertook some investigations into the influence of shear strength gradient, comparing pipeline penetration for a uniform shear strength and a shear strength gradient with a zero strength intercept at mudline. However, it does not appear that the shear strength variation associated with crusts has been investigated previously with regards to pipeline penetration resistance. A similar design problem, albeit with some differences in geometry, is the problem of a strip footing on an undrained soil. For this problem a linearly increasing shear strength gradient has been investigated for surface foundations (Davis and Booker, 1973) and embedded foundations (Bransby and Yun, 2009). Again it appears that a shear strength crust has not been considered specifically.

821

Figure 2. Representation of a shear strength crust.

Figure 1. Problem definition.

1.2

Scope of study and problem definition

The analysis and results presented in this paper are part of a wider study into pipe soil interaction on clay seabeds. The aim of this element of the study is to investigate the influence of a range of variable undrained shear strength profiles on the vertical penetration resistance of a pipeline. This paper will initially considered shear strength profiles with a linear variation in strength with depth before being extended to consider the influence of shear strength crusts. As is implicit with the use of an undrained shear term, this problem is considered at a rate of loading that provokes an undrained soil response. A pipeline of a given diameter, D, embedded in a clay seabed to depth zp is considered (Figure 1a). The seabed can be uniform strength (Figure 1b), have a linear increasing shear strength profile with a zero strength intercept at mudline (Figure 1c), or have a linearly increasing shear strength but a non-zero shear strength at mudline (Figure 1d). To increase generality of the results from the analyses it is useful to express resistance to penetration in terms of the dimensionless group R/su .D, where R is the vertical bearing capacity, su is the undrained shear strength and D is the diameter of the pipeline. When shear strength varies with depth, the shear strength at the base of the pipeline will be used, suzp at depth zp . This approach is consistent with Aubeny (2005). To quantify the different linear shear strength gradients (Figure 1b-d), the shear strength at mudline, su0 was divided by suzp . For example su0 /suzp = 0.5 represents the case where the shear strength intercept at mudline is 50% of the shear strength at the pipeline embedment depth. For this problem this provides an easier visualization of the non-uniformity of the soil strength than the more conventional used dimensionless term kD/suzp , where k is the gradient of shear strength change with depth. Note that the nonuniformity will vary with pipeline penetration and suzp © 2011 by Taylor & Francis Group, LLC

increases for all cases except that with uniform soil strength (Figure 1b). A representation of a shear strength crust for use in analysis is shown in Figure 2. This treats the crust as a departure from a given linear increasing shear strength gradient. The geometry of the crust is captured by a steep positive linear shear strength gradient extending to a peak of suct , at a depth zcp . The shear strength then returns to the underlying shear strength gradient using a negative shear strength gradient, intercepting at a depth zc . Although only a limited number of examples of crusts are considered in this paper, this approach should be relatively robust in representing a wide range of shear strength crusts. A total of 50 analyses were undertaken in this study to investigate the influence of shear strength gradient. These analyze investigated two interface conditions, perfectly smooth and perfectly rough, five “wished in place” embedment depths from 0.1D to 0.5D, and five shear strength gradients. The shear strength gradients investigated were the extremes of a constant shear strength and a gradient with an intercept at mudline of suo = 0. Intermediate shear strength gradients of su0 /suzp = 0.25, su0 /suzp = 0.5 and su0 /suzp = 0.75 were also analyzed. For the case of a shear strength crust. there is very little generality with a large range of possible cases, at least from a geometrical standpoint. In addition there is little discussion in the literature on common geometries, with only a limited number of examples reported in the public domain (e.g. Ehlers et al., 2005). This paper presents analysis for a limited number of cases with the aim of highlighting some of the issues related to shear strength crusts and pipe-soil interaction. Further work is ongoing. A total of sixteen analyses where undertaken to consider shear strength crusts. Both completely rough and smooth interface conditions were considered for one embedment depth, 0.3D. The crust was taken as being symmetrical, i.e. zcp = zc /2 (see Figure 2), and could therefore be defined in terms of just two parameters, zcp and suct . The depth to crust peak, zcp was expressed in dimensionless form, with respect to pipeline diameter, and suct was described as a multiple of the

822

underlying shear strength at zcp . Two strength ratios were considered: five and ten. Four crust depths were analyzed, zcp /D = 1, 0.433, 0.3 and 0.2 was selected for the smooth interface case and zcp /D = 1, 0.5, 0.3 and 0.2 for the rough interface. 2

METHODOLOGY

The analysis presented in this paper was undertaken using the Finite Difference code FLAC (Fast Lagrangian Analysis of Continua) Version 6.0 (Itasca, 2008). Analysis was undertaken under plane strain conditions in a small strain calculation mode. Computational time was reduced by considering half a pipeline around a central axis of symmetry. Grid density was selected from grid convergence studies. Grid density requirements varied with embedment depth and the highest grid densities were required for zp = 0.1D and 0.5D. The overall boundary dimensions were selected to ensure they did not affect the calculated failure load. The width of the grid also needed to be widened significantly to accommodate a punch-through mechanism for the shear strength crust analysis. For example a 0.65 m by 0.65 m grid was required for a 0.3D embedment with D = 0.3 m. However for later analysis this was extended laterally to 1.20 m due to the larger failure mechanism associated with punch-through. The pipeline shape was formed from a section of grid above the soil surface, before being wished in place to the selected depth of embedment. The interface between the pipeline and the soil was controlled by interface elements (considering perfectly rough and smooth conditions) and the pipeline shape was given rigid behavior by application of a uniform fixed velocity (or displacement) boundary. A series of displacement controlled analysis were undertaken. The associated load-displacement behavior was calculated by summing the vertical nodal reaction forces on the pipe-soil interface. Each analysis was run until a constant load was reached representing the ultimate capacity. The seabed soil was modeled as a linear elastic perfectly plastic material with a Tresca yield criterion. Initially, a constant shear strength was assigned, but subsequent analyses were undertaken with a variation in shear strength with depth as previously described. Bulk Modulus (K) and Shear Modulus (G) were assigned based on a Poisson’s Ratio of ν = 0.49 and a Young’s Modulus of E = 200.su . On this basis, elastic stiffness parameters also varied with depth in line with changing shear strength. 3

RESULTS AND DISCUSSION

3.1 Linear increasing shear strength Results were expressed in terms of the dimensionless group R/su .D (see Figure 3 and 4). There was reasonable agreement between the results for the constant shear strength case and previous work by Aubeny et al. © 2011 by Taylor & Francis Group, LLC

823

Figure 3. R/su .D against z/D for a smooth interface condition.

Figure 4. R/su .D against z/D for a rough interface condition.

(2005) and Merifield et al. (2008). Analysis results were also normalized by the capacity for the constant shear strength case, Rconstant , which simplified interpretation. Figure 5 presents results for the case of a smooth interface condition and Figure 6 shows results for the rough interface case. The relationship between shear strength gradient and peak resistance to vertical penetration is relatively complex, with the shear

Figure 5. Change in vertical capacity for different shear strength gradients and embedments for a pipeline with a smooth interface.

Figure 6. Change in vertical capacity for different shear strength gradients and embedments for a pipeline with a rough interface.

strength gradient having a clear influence. However, this influence varies with both depth of embedment and interface condition. For the smooth interface condition, larger variations in shear strength generally result in reduced resistance to pipeline penetration for larger embedment. However, at shallower depth, resistance to penetration is generally increased. In contrast, for the rough interface condition, a larger shear strength gradient increases the resistance for all but the deepest embedments and steepest strength gradients. Additional insight into the reasons for these results can be gained by considering the failure mechanisms associated with variation of the shear strength gradient. Figures 7a to 7e show changes in failure mechanism associated with variation in shear strength gradient © 2011 by Taylor & Francis Group, LLC

824

Figure 7. Displacement vector plots at failure for a smooth pipeline with zp /D = 0.3.

Figure 8. Displacement vector plots at failure for zp /D = 0.1 and smooth interface. Uniform shear strength (left) and suo /suzp = 0 (right).

Figure 9. Displacement vector plots at failure for zp /D = 0.3 and rough interface. Uniform shear strength (left) and suo /suzp = 0 (right).

for a pipeline embedded at zp = 0.3D with a smooth interface. With a smaller value of suo /suzp there is an increase in the shear strength below the base of the pipeline and a decrease above. Figure 7 shows that non-uniformity produces a change in the failure mechanism as it becomes preferential for the mechanism to migrate upwards into the lower strength material. With a smooth interface condition this effect will reduce the resistance at larger embedments (Figure 3 and 5) At shallower depths the shear strength below the base of the pipeline increase rapidly, for given gradient, in conjunction with a relatively small zone of lower shear strength material above the pipeline base. Figure 8 provides an example of the deformation mechanisms for a shallowly embedded pipeline. Relatively high shear strength material beneath the pipeline must be sheared, consistent with a greater resistance then for the constant shear strength case. For the rough interface condition, increased variation in shear strength gradient increases the resistance for all but the deepest embedments and steepest gradients (Figure 4 and 6). A rough interface mechanism (Figure 9) is larger and mobilizes soil at a greater depth then the equivalent smooth interface behavior, which is consistent with the increased failure loads (Figure 4 and 6). 3.2 Shear strength crusts The influence of shear strength crusts was investigated and comparison made with the results previously described for linear increasing shear strength gradients (Figure 10). This shows the normalized pipeline capacity with the peak strength of the crust at different depths for four conditions: (i) for rough and smooth interface conditions, and; (ii) for a crust where the shear strength © 2011 by Taylor & Francis Group, LLC

Figure 10. Normalized penetration resistance against crust depth for a pipeline with zp = 0.3D.

Figure 11. Failure mechanism showing punch-through: smooth interface; zp /D = 0.3; zcp /D = 0.43; x10 crust.

is 5 or 10 times the strength of the underlying shear strength gradient. Note that the penetration resistance is normalized by the strength at the embedment depth, zp , which varies with crust depth. When the crust peak, zcp , is relatively deep, for example 1.0D in the data shown in Figure 10, the vertical bearing capacity is identical to that in soil with a linear increasing shear strength profile extending to infinite depth (the solid lines on Figure 10). This is because the bearing capacity is dominated by the strength and gradient in the upper part of the crust. In addition, the failure mechanism was identical to those previously seen (Figure 7 and 9). The results for a crust peak depth of 0.43D and 0.5D shows a reduction in resistance. This can be attributed to a punch-through mechanism, which can be seen in Figure 11. The failure mechanism extends into the lower shear strength material below the crust resulting in a very large deformation mechanism, but a smaller bearing capacity. The analysis for shallower crust depths (e.g. zcp /D = 0.3, 0.2 in Figure 10) also give reduced resistance relative to the case of constant linear increasing gradient, with the biggest reduction occurring when

825

the pipeline embedment depth coincides with the depth of the crust peak. This case in particular highlights that the failure mechanism can encompasses a large proportion of lower shear strength material than is characterized by the shear strength at the base of the pipeline. The case with zcp /D = 0.2, is perhaps the most complex. Normalized bearing capacity is slightly larger than the case with zcp /D = 0.3, albeit significantly below the resistance for zcp /D = 1. This can be attributed to a mechanism that encompasses lower strength material below zp and higher strength material above zp . The mechanism was shallower as it intercepted the increasing shear strength gradient below the crust, and this also contributes to the relative increase in resistance. As well as the crust depth being important, the strength of the crust peak compared to the underlying shear strength gradient was also critical, especially close to the onset of punch-through. When there was a larger strength difference, punch through was more marked. This is likely to relate to an increased size mechanism, including proportional weaker soil encompassed by the mechanism, associated with a stronger crust. 4

FURTHER CONSIDERATIONS

The analysis presented in this paper incorporates a number of simplifications as described in section 2. These simplifications may need to be addressed or otherwise quantified for specific pipelines or soil conditions. In addition a number of uncertainties remain with regards to the origin and properties of shear strength crusts incuding time dependent variation. A limited number of examples have been presented in this paper and work is currently ongoing to consider a wider range of crusts and produce generalized design guidance. 5

CONCLUSIONS

This paper reports the results of a study investigating the influence of clays with variable shear strength profiles on pipe-soil interaction. Various linear increasing profiles were considered, both as an area of interest in itself and to provide background material for subsequent consideration of shear strength crusts. In addition, results have been shown for analyses with shear strength crusts of different strength relative to the underlying strength gradient, variation of the depth to crust peak, and variation in the interface conditions. The presence of a shear strength gradient was seen to influence vertical bearing capacity, particularly for the rough interface conditions. However, the presence of a shear strength crust had the most dramatic effect. When the crust was deep the behavior was dominated by the upper shear strength gradient, but as the crust peak got closer to the base of the pipeline a punchthrough mechanism was observed. © 2011 by Taylor & Francis Group, LLC

A number of examples have been presented in this paper, where the bearing capacity cannot be calculated accurately using a single shear strength without consideration of either shear strength gradient and/or the geometry of the shear strength crust. REFERENCES Aubeny, C., Shi.,H & Murff, J. 2005. Collapse load for cylinder embedded in trench in cohesive soil. International Journal of Geomechanics 5(4): 320–325. Bransby, M.F. & Yun, G.J. 2009. The undrained capacity of skirted strip foundations under combined loading. Geotechnique 59(2): 115–125. Brennodden, H., Sveggen, O., Wagner, D.A. & Murff, J.D. 1986. Full scale pipe-soil interaction tests. Proceedings of OTC 1986. OTC Paper No. 5338. Bruton, D., White, D., Check,C., Bolton, M. & Carr, M. 2006. Pipe soil interaction behavior during lateral buckling including large amplitude cyclic displacement tests by the SAFEBUCK JIP. Proceedings of OTC 2006. OTC Paper No. 17944. Davis, E.H. & Booker, J.R. 1973. The effect of increasing shear strength with depth on the bearing capacity of clays. Geotechnique 23(4): 551–563. Dingle, H.R.C., White, D.J. & Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout on soft clay. Canadian Geotechnical Journal, 45(5): 636–652. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origin of near-seafloor “crust zones” in deepwater. Proceedings of ISFOG 2005: 927–933. Hill, A.J. & Jacob, H. 2008. In-situ measurement of pipe-soil interaction in deep water. Proceedings of OTC 2008. OTC Paper No. 19528. Hodder, M.S., Cassidy, M.J. & Barret, D. 2008. Undrained response of shallow pipelines subjected to combined loading. Proceedings of ICOF 2008: 897–908. Itasca Consulting Group. 2008. FLAC – Fast Lagrangian Analysis of Continua – Users Guide. Kuo, M.Y.H. & Bolton, M.D. 2009. Soil characterization of deep sea west African clays: is biology a source of mechanical strength. Proceedings of ISOPE 2009: 488–494. Merifield, R., White, D.J. & Randolph, M.F. 2008. The ultimate undrained resistance of partially embedded pipelines. Geotechnique 58(6): 461–470 Morrow, D.R. & Bransby, M.F. 2009. The influence of slope on the stability of pipelines subjected to horizontal and vertical loading on clay seabeds. Proceedings of OMAE 2009. Perinet, D. & Fraser, I. 2006. Mitigation methods for deepwater pipeline instability induced by pressure and temperature variation. Proceedings of OTC 2006. OTC Paper No. 17815 Puech, A, Colliat, J.L., Nauroy,J-F. & Menier,J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments. Proceedings of ISFOG 2005: 1047–1053. Randolph, M.F. & White, D.J. 2008. Upper bound yield envelopes for pipelines at shallow embedment depths in clay. Geotechnique 58(4): 213–229. Verley, R.L.P. & Lund, K.M. 1995 A soils resistance model for pipeline place on clay soils. Proceedings of OMAE 95, 225–232. Yun, T.S., Narsilio, G.A. & Santamarina, J.C. 2006. Physical characterization of core samples recovered from Gulf of Mexico. Marine and Petroleum Geology 23: 893–900.

826

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Sweeping behaviour of shallowly-embedded pipeline during cyclic lateral movement T. Takatani Department of Civil Engineering, Maizuru Nat’l College of Technology, Maizuru, Kyoto, Japan

ABSTRACT: A non-linear finite element analysis based on an effective stress theory for pipeline-seabed interaction problem was carried out in order to simulate the sweeping behaviour of a shallowly-embedded pipeline on carbonate sandy soil subjected to cyclic lateral movement under a constant vertical load. Pipeline sweeping behaviour during cyclic lateral movement was numerically investigated in terms of the initial depth of pipeline, the carbonate soil conditions, the amplitude and frequency of cyclic lateral movement under constant vertical loading. The large-amplitude and cyclic pipe-soil interaction is discussed. Pipeline sweeping behaviour greatly depends on the amplitude and frequency of cyclic lateral movement.

1

INTRODUCTION

Shallowly-embedded offshore pipelines are directly exposed to the vertical and horizontal forces induced by the hydrodynamic environment. The cyclic movement of pipeline due to both drag and lift forces caused by waves and currents will lead to a large deformation of pipeline. On the other hand, the seabed soil berms created at the extremities of the cyclic lateral sweeping range and the remoulding of the seabed soil generated by the cyclic lateral movement of a pipeline have recently investigated in the offshore pipeline engineering field. Recently, Dingle et al. (2008) observed the deformation mechanism during cyclic lateral movement of pipeline through some centrifuge tests, and the in-flight images which indicate pipeline breakout and large amplitude sweeping were evaluated using Particle Image Velocimetry (PIV) analysis. White and Cheuk (2008) proposed a simplified modelling of cyclic lateral pipe-soil interaction, based on the accumulation and deposition of berm materials. Hodder et al. (2008) conducted a series of centrifuge model tests to evaluate a riser-soil interaction within the touchdown zone of a steel catenary riser during pipeline laying process. Pipeline sweeping behaviour due to cyclic lateral movement strongly depends on a pipe-soil interaction. It is, therefore, very important to investigate the largeamplitude and cyclic pipe-soil interaction behaviour using experimental tests and numerical analyses in order to accurately evaluate the pipeline sweeping behaviour during cyclic lateral movement. The purpose of this paper is to investigate the pipeline sweeping behaviour during cyclic lateral movement, focusing on the large-amplitude and cyclic pipe-soil interaction behaviour during severe storm © 2011 by Taylor & Francis Group, LLC

condition. In this paper, a two-dimensional non-linear finite element analysis based on an effective stress theory is employed for a pipe-soil interaction problem. 2

CYCLIC LATERAL PIPELINE-SEABED INTERACTION ANALYSIS

An advanced numerical analysis in this paper is two-dimensional dynamic non-linear finite element method (Takatani et al., 2005, 2008) based on the effective stress theory in order to simulate a sweeping behaviour of shallowly-embedded pipeline on carbonate sandy soil under undrained condition. In this finite element analysis, a non-linear relationship between shear stress and shear strain of soil element is accurately expressed by a multi shear spring model (Towhata and Ishihara, 1985) and the Masing rule for loading and unloading curves is employed so as to adjust the amplitude of hysteresis damping for the multi shear spring model. Also the cyclic mobility model (Iai et al., 1990), which is of a generalized plasticity-multiple mechanism type, is adapted to simulate excess pore water pressure. Pore fluid is assumed to be imcompressible, and also the viscous boundary technique (Lysmer et al., 1969) is used to create the infinite of seabed soil in this analysis. There are three governing equations of a kinematic equation between soil and pipeline, pore water input/output balance equation in each pore fluid element, and dynamic water pressure wave propagating equation for pore fluid. Pore water pressure can be expressed by an increment of volumetric strain of soil skeleton because of undrained condition, and also dynamic water pressure wave propagating equation for pore fluid can be represented by a technique that the effect of pore fluid

827

Figure 1. Fem mesh for pipeline-seabed interaction analysis (z0 = 0.1 m.)

existence can be taken into consideration by applying an additional mass of each pore fluid element to the soil-structure kinematic equation. In this paper, the sweeping behaviour of a shallowlyembedded pipeline resting on carbonate sandy soil is investigated from a view point of carbonate sandy soil characteristics. Pipe is assumed to be 1.0 m diameter and both 0.1 m and 0.25 m for its initial depth, and the seabed is assumed to be a carbonate soil. Pipeline sweeping behaviour during cyclic lateral movement is numerically investigated in terms of the stiffness of joint element between pipe and seabed soil, the initial depth of pipeline, the carbonate soil conditions, the amplitude and frequency of cyclic lateral movement under a constant vertical loading. Figure 1 shows a finite element mesh for an unburied offshore pipeline-seabed interaction analysis considering a liquefaction phenomenon in the seabed around the pipeline. The joint element is used at contact area between pipeline and sand layer to represent a slip phenomenon at contact area between pipeline and seabed surface (Takatani et al., 2005). The initial depth of pipeline, z0 , is 0.1 m as shown in Figure 1.The model of initial depth of pipeline, z0 = 0.25 m, is described in Figure 1, too. The analytical domain is 7 m × 10 m and is assumed to be a carbonate sand layer. Both 6node triangle and 8-node square elements are used in this mesh, and also the Selective Reduced Integration method (Hughes, 1980) by which each soil element integration can be separately evaluated for both the volumetric and deviation components is employed in order to make an accurate evaluation for each soil element integration. At every incremental time step, the coordinate of each nodal point is renewed according to the soil deformation, and the stress loading of each element is re-evaluated by a self-weighted analysis result at every time step. This finite element analysis with a coarse mesh in the vicinity of the pipeline shown in

© 2011 by Taylor & Francis Group, LLC

Table 1.

828

Material properties of carbonate sand (Dr = 60%).

Initial shear modulus, Gma (kPa) Elastic tangent bulk modulus of soil skeleton, Kma (kPa) Friction angle, φf (degree) Phase transformation line angle, φp (degree) Material parameters for dilatancy S1 w1 c1 p1 p2

48,395 126,207 38.16 28 0.005 4.634 1.548 0.500 1.037

Figure 1 can be carried out with sufficient accuracy (Ozutsumi, 2003). The mechanical properties for carbonate soils with the relative densities Dr = 60% and 80% are evaluated from the liquefaction resistance curves, that is, the effective stress ratio vs. the number of cycles. The material properties of carbonate soil for the relative density Dr = 60% is shown in Table 1, and also are obtained from the liquefaction resistance curve for carbonate sand indicated in Figure 2(a) (Aramaki, 1997). Figure 2(b) shows the liquefaction resistance curve for carbonate sand with the relative density Dr = 80%. As a reference, the liquefaction resistance curves for Toyoura sand are indicated, too. Five parameters shown in Table 1 which specify the dilatancy are determined by the back-fitting technique to the liquefaction resistance curves of the carbonate sand obtained from the laboratory test. In general, the liquefaction resistance curve is determined by combining the laboratory test data and the bearing capacity test data at the sites for taking the in-situ conditions of soils into account (Morita et al., 1997). Before the cyclic lateral movement of pipeline, the self-weighted analysis for pipeline-seabed interaction

Figure 2. Liquefaction resistance curve for carbonate sand (Aramaki, 1997).

problem is carried out under the completely drained condition to obtain the initial effective stress of each soil element. In this numerical pipeline-seabed interaction analysis, a strain space plasticity approach is assumed to be used for cyclic mobility in order to represent the realistic hysteretic damping factor under cyclic loading. In this approach, actual cyclic shear mechanism is decomposed into a set of one dimensional virtual simple shear mechanism. Material properties of dilatancy S1 , w1 , c1 , p1 and p1 shown in Table 1 are five parameters to define the cumulative volumetric strain of plastic nature for representing cyclic mobility. These parameters define the correlation between the liquefaction front parameter (Iai et al., 1990) and the normalized shear work. The liquefaction front parameter is given by a function of shear work, and Towhata and Ishihara (1985) obtained the correlation between the shear work and the excess pore pressure, and also concluded that the correlation is independent of the shear stress paths with or without the rotation of principal stress axes. 3

NUMERICAL RESULTS

In the sweeping behaviour analysis of a shallowlyembedded pipeline subjected to cyclic lateral movement, the vertical load, V , is assumed to be maintained a constant value during cyclic lateral movement, H .

© 2011 by Taylor & Francis Group, LLC

Before cyclic lateral movement, the static analysis subjected to a constant vertical load, V , is carried out, and then both a constant vertical load, V , and cyclic lateral movement, H , operate at the centre of pipeline as shown in Figure 1. Figure 3 shows seabed deformation around a pipeline after 1,000 cyclic lateral movements for two initial depths of pipeline z0 = 0.1 m and 0.25 m in the relative density Dr = 80%, the frequency f = 0.5 Hz of cyclic lateral movement and the constant vertical load V = 4 kN/m. The unit tangential stiffness for normal and shear directions, Kn and Ks , f or a joint element are used 1.0 × 106 (kN/m) and 1.0 × 105 (kN/m), respectively. The friction angle of joint element is assumed to be 25 degree in this analysis. It can be observed from these figures that the settlement of pipeline increases with the number of cycles, and that greatly depends on the amplitude of cyclic lateral movement, H . The larger the amplitude of cyclic lateral movement becomes, the more widely and deeply the seabed surface around a pipeline will be excavated by the cyclic lateral movement of pipeline. Pipeline settlement after 1,000 cyclic lateral movements in the initial depth of pipeline, z0 = 0.1 m, is much larger than that in z0 = 0.25 m for each amplitude of lateral movement, H . This is because that the soil resistance force to pipeline increases with the initial depth of pipeline, z0 . Although the seabed deformation after 1,000 cyclic lateral movements in the relative density Dr = 60% for each amplitude of cyclic lateral movement, H , is not illustrated in this paper on account of space consideration, pipeline settlement slightly increases with the decrease of the relative density, Dr. Figure 4 shows pipeline sweeping behaviour during cyclic lateral movement in Dr = 80%, H = ±0.05 m and V = 4 kN/m. Pipeline settlement behaviour in the frequency f = 2.0 Hz of cyclic lateral movement is indicated in Figure 4(a). It can be seen from this figure that the large settlement in pipeline sweeping behaviour occurs within several cyclic lateral movements at initial cyclic stage and then gradually approaches a certain value with the number of cycles. Figure 4(b) illustrates pipeline settlement behaviour during 1,000 cyclic lateral movements in the frequencies f = 0.5, 1.0 and 2.0 Hz. Pipeline settlement increases with the frequency, f , of cyclic lateral movement, and also each settlement behaviour seems to increase with the number of cycles, N , after 1,000 cycles. It should be noted that pipeline settlement greatly depends on the frequency, f , of cyclic lateral movement, H . Table 2 indicates pipeline settlement after 1,000 cyclic lateral movements for each amplitude of cyclic lateral movement, H , in the relative density Dr = 60%, the initial depths of pipeline, z0 = 0.1 m and 0.25 m, the cyclic frequencies f = 0.5, 1.0 and 2.0 Hz. It can be seen from this table that pipeline settlement has a tendency to increase with the amplitude, H , and the cyclic frequency, f , of cyclic lateral movement. On the other hand, Table 3 illustrates pipeline settlement results after 1,000 cyclic lateral movements

829

Figure 3. Seabed deformation around a pipeline behaviour after1,000 cyclic lateral movements (Dr = 80%, f = 0.5 Hz, V = 4 kN/m).

© 2011 by Taylor & Francis Group, LLC

830

Figure 4. Pipeline behaviour during cyclic lateral movement (Dr = 80%, H = ±0.05m, V = 4 kN/m). Table 2. Pipeline settlement after 1,000 cyclic lateral movements (Dr = 60%, V = 4 kN/m)

Table 4. Pipeline settlement after 1,000 cyclic lateral movements (f = 2.0 Hz, H =±0.1 m)

Settlement (m) after 1,000 cycles f = 1.0 Hz

f = 2.0 Hz

(a) z0 = 0.1 m ±0.05 −1.292 ±0.1 −1.859 ±0.2 −2.784 ±0.5 −4.761

−1.346 −1.974 −2.928 −5.548

−0.997 −1.756 −3.238 −5.839

(b) z0 = 0.25 m ±0.05 −0.697 ±0.1 −0.875 ±0.2 −1.247 ±0.5 −2.571

−0.787 −1.114 −1.882 −3.641

−0.892 −1.518 −2.419 −5.415

H (m)

f = 0.5 Hz

Table 3. Pipeline settlement after 1,000 cyclic lateral movements (Dr = 80%, V = 4 kN/m) Settlement (m) after 1,000 cycles f = 1.0 Hz

f = 2.0 Hz

(a) z0 = 0.1 m ±0.05 −1.054 ±0.1 −1.244 ±0.2 −1.954 ±0.5 −3.620

−0.988 −1.382 −2.221 −4.137

−1.032 −1.549 −2.488 −4.137

(b) z0 = 0.25 m ±0.05 −0.426 ±0.1 −0.615 ±0.2 −1.018 ±0.5 −1.916

−0.514 −0.830 −1.408 −2.872

−0.678 −1.148 −2.076 −4.364

H (m)

f = 0.5 Hz

for each amplitude of lateral movement, H , in the relative density Dr = 80%. It can be observed from Tables 2 and 3 that pipeline settlement increases with the decrease of the relative density, Dr. This is because that the soil resistance force to pipeline increases with the relative density, Dr. © 2011 by Taylor & Francis Group, LLC

V (kN/m)

Dr = 60%

Dr = 80%

(a) z0 = 0.1 m 0.5 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0

−0.844 −1.071 −1.223 −1.386 −1.756 −1.787 −1.800 −2.088 −2.273

−0.996 −1.161 −1.174 −1.309 −1.549 −1.817 −1.843 −1.891 −2.138

(b) z0 = 0.25 m 0.5 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0

−0.659 −0.804 −0.988 −1.172 −1.518 −1.846 −1.657 −1.800 −2.000

−0.539 −0.623 −0.993 −0.979 −1.148 −1.319 −1.351 −1.521 −1.564

Table 4 shows the effect of constant vertical load, V , on the pipeline settlement in the initial pipeline depths, z0 = 0.1 m and 0.25 m. It can be seen from this table that pipeline settlement increases with the constant vertical load, V . As the soil resistance force to pipeline increases with the initial depth of pipeline, pipeline settlement may trend to decrease if the initial pipeline depth, z0 , is large and the constant vertical load, V , is small. Table 5 indicates the effect of joint element stiffness on the pipeline settlement in the initial pipeline depths, z0 = 0.1 m and 0.25 m, the relative densities, Dr = 60% and 80%. It can be clearly found from this table that the settlement of pipeline increases with the stiffness of joint element. This is because the pipeline movement is directly transmitted to the seabed soil around pipeline with the increase of joint element stiffness. Although this effect of joint element stiffness is not

831

Table 5. Pipeline settlement after 1,000 cyclic lateral movements (H = ±0.1 m, f = 2.0 Hz, V = 4 kN/m). Dr = 60%

Dr = 80%

Kn , Ks (kN/m)

z0 = 0.1 m z0 = 0.25 m z0 = 0.1 m z0 = 0.25 m

108 ,107 106 ,105 104 ,103 103 ,102 102 ,101

−1.756 −1.751 −1.513 −1.273 −0.547

−1.583 −1.518 −1.209 −1.134 −0.604

−1.537 −1.249 −1.053 −0.851 −0.392

−1.202 −1.148 −0.858 −0.670 −0.383

REFERENCES

mentioned in detail in this paper on account of space consideration, Takatani (2005) discussed the effect of joint element stiffness on the pipeline behaviour during cyclic loading. 4

CONCLUSIONS

The advanced finite element analysis based on an effective stress theory was conducted to simulate the sweeping behaviour of a shallowly-embedded pipeline on carbonate sandy soil during cyclic lateral movement. In summary, the following conclusions can be made based on the results presented in this paper. 1. Pipeline settlement greatly depends on the amplitude, H , and frequency, f , of cyclic lateral movement, the initial pipeline depth, z0 , the vertical load, V , and the stiffness, Kn and Ks , of joint element between pipeline and seabed surface. 2. Because the soil resistance force to pipeline increases with the initial depth of pipeline, the settlement of pipeline after 1,000 cyclic lateral movements in the initial depth of pipeline,z0 = 0.1 m, is much larger than that in z0 = 0.25 m for each amplitude of cyclic lateral movement. 3. Pipeline settlement after 1,000 cyclic lateral movements in the relative density Dr = 80% is smaller than that in Dr = 60%, because the soil resistance force to pipeline increases with the relative density. In this paper, the sweeping behaviour of a shallowlyembedded pipeline on carbonate sandy soil was investigated under undrained condition. In future, there seems to be a need for further consideration on this analytical condition in the pipeline-seabed interaction problem. Although the pore pressure accumulation in

© 2011 by Taylor & Francis Group, LLC

the seabed around a pipeline is not presented in this paper due to the limited space, it is necessary for an intensive study on the effect of pore pressure accumulation on the pipeline sweeping behaviour because it is the most important role in the design of offshore pipeline. In addition, further investigation may be needed to simulate these phenomena mentioned above and make some concrete conclusions.

Aramaki, N. 1997. Undrained cyclic and monotonic triaxial behaviour of crushable carbonate soil, Dr. Eng. Thesis, Yamaguchi University. Cheuk, C.Y., White, D.J. and Bolton, M.D. 2008. Uplift mechanism of pipes buried in sand, J Geotech & Geoenvironmental Eng, Vol.134, No.2, pp.154–163. Dingle, H.R.C., White, D.J. and Gaudin, C. 2008. Mechanism of pipe embedment and lateral breakout on soft clay, Canadian Geotech J, Vol.45, No.5, pp.636–652. Hodder, M.S., White, D.J. and Cassidy, M.J. 2008. Centrifuge modeling of riser-soil stiffness degradation in the touchdown zone of a steel catenary riser, Proc 27th Int Conf Offshore Mech Arctic Eng (OMAE), ASME, Estoril, Portugal, CD, OMAE2008-57302. Hughes, T.J.R. 1980. Generalization of selective integration procedures to anisotropic and nonlinear media, Int J Num Meth Eng, Vol.15, pp.1413–1418. Iai, S., Matsunaga, Y. and Kameoka, T. 1990. Strain space plasticity model for cyclic mobility, Report of Port and Harbour Research Institute, Vol.29, No.4, pp.27–56. Lysmer, J. and Kuhlemeyer, R.L. 1969. Finite dynamic model for infinite media,J Eng Mech Div, ASCE, No.EM4, pp.859–877. Morita,T, Iai, S, Liu, H, Ichii, K, and Sato,Y. 1997. Simplified Method to Determine Parameter of FLIP, Material of Port and Harbour Research Institute, No 869. Ozutsumi, O. 2003. Numerical analysis on seismic damage estimation for soil-structure system on liquefied area, Dr. Eng. Thesis, Kyoto University. Takatani, T. 2005. Pipeline-seabed interaction analysis subjected to horizontal cyclic loading, Proc Int Symp Frontiers Offshore Geomech, Perth, pp.629–635. Takatani, T. and Kaya, T. 2008. Unburied offshore pipeline stability analysis based on non-linear relationship between pipeline and carbonate soil, Proc 18th Int Offshore Polar Eng Conf, Vancouver, Canada, Vol.2, pp.176–185. Towhata, I. and Ishihara, K. 1985. Modelling soil behaviour under principal stress axes rotation, Proc 5th Int Conf Num Method Geomech., Nagoya, pp.523–530. White, D.J. and Cheuk, C.Y. 2008. Modelling the soil resistance on seabed pipelines during large cycles of lateral movement,Marine Structures, Vol.21, No.1, pp.59–79.

832

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Advanced nonlinear hysteretic seabed model for dynamic fatigue analysis of steel catenary risers I.H.Y. Ting, M. Kimiaei & M.F. Randolph Centre for Offshore Foundation Systems, The University of Western Australia, Perth

ABSTRACT: Fatigue of steel catenary risers (SCRs) in the touchdown zone (TDZ) remains one of the greatest challenges in designing SCRs. This is because of the nonlinear pipe-soil interaction in the TDZ and the random cyclic motions to which the SCRs are subjected. The soil parameters used to model the interaction can also significantly affect the TDZ fatigue response of SCRs. Most traditional fatigue design approaches for SCRs are based on an assumed linear stiffness for the seabed and tend to provide very conservative results, particularly in deep water. The response of SCRs as a result of hysteretic nonlinear pipe-soil interaction is difficult to predict and improved understanding of this interaction is therefore vital for more accurate fatigue design. In this paper, the results of a parametric study are presented, using a nonlinear pipe-soil interaction model to investigate how the main geotechnical parameters influence the fatigue life of SCRs. The main parameters considered are: maximum normalised stiffness, soil suction ratio and shear strength gradient. The paper summarises the relative effects of these parameters on the fatigue life, concluding that the shear strength gradient and soil suction ratio and are the most critical parameters affecting the SCR fatigue life. 1

INTRODUCTION

As the offshore industry continues to progress developments in deepwater fields, steel catenary risers (SCRs) are often the preferred riser option for subsea tieback to floating platforms. This is due to their conceptual simplicity, ease of construction and installation and simple interface with the flowlines. Modelling of pipesoil interaction in the touchdown zone (TDZ), where the riser meets the seabed, is one of the critical challenges for SCRs, since it has a significant effect on the fatigue life in that zone. The interaction exhibits complex behaviour and is highly nonlinear in response to the random cyclic motions to which the SCR is subjected (Grealish et al. 2007). The pipe-soil interaction in the TDZ is critical for accurate estimation of the fatigue life. Traditionally fatigue assessment of SCRs is carried out using linear soil springs, with or without damping, despite the awareness of the nonlinear interaction in the TDZ (Clukey et al. 2004). Linear pipe-soil interaction models ignore or simplify much of the fundamental response of soil, such as variation of the secant stiffness depending on the amplitude of cyclic displacement, suction during uplift and softening of soil under cycling motions. In spite of this, linear springs are adopted partly because of previous software limitations and also partly because linear solutions greatly simplify fatigue study. The limited understanding of the actual nonlinear interaction has also restricted development of appropriate pipe-soil interaction models, although a number of models have been proposed recently © 2011 by Taylor & Francis Group, LLC

833

(Aubeny & Biscontin 2009, Randolph & Quiggin 2009). It is important, however, not to restrict the modelling of such effects to linear approximation, in particular for ultra deepwater developments, as it can lead to a conservative design approach (Clukey et al. 2007, Grealish et al. 2007). It is therefore important to model the interaction accurately to minimise conservatism in fatigue response predictions for SCRs. Significant research programmes, laboratory testing and field-scale experiments such as the STRIDE and CARISIMA JIPs (Bridge et al. 2004) have helped to understand the interaction and provide a basis for nonlinear soil models. The soil parameters used for SCR analysis can have a significant effect on riser response especially on the predicted fatigue life (Bridge et al. 2004). It is therefore important to study the sensitivity of fatigue results to soil input parameters. The soil model presented by Randolph & Quiggin (2009) has been used in this study. This soil model is based on a hyperbolic secant stiffness formulation such as proposed by Bridge et al. (2004) with asymptotic limiting penetration and uplift resistance. It is able to capture the variation of stiffness with the amplitude of cyclic displacements. The results of a parametric study of fatigue damage based on this nonlinear pipe-soil interaction model are presented here. A series of dynamic riser response analyses were carried out to investigate how the main geotechnical parameters influence the fatigue life of SCRs in the TDZ. The main parameters considered are: maximum normalized stiffness, soil suction ratio and shear strength gradient. The paper summarises how these

input parameters influence the fatigue results of an example deepwater SCR.

The key soil input parameters, and the manner in which they affect the response, are described below.

2

2.1

PIPE-SOIL INTERACTION MODEL

The hysteretic nonlinear pipe-soil interaction model presented by Randolph and Quiggin (2009) is used in this study. As shown in Fig. 1, in general there are four different penetration modes in this model: not in contact, initial penetration, uplift and repenetration. The primary input data into the model are the pipe diameter, seabed soil shear strength profile (assuming a linear strength profile with mudline intercept sum and gradient ρ) and soil density. Additional parameters include: normalised maximum stiffness of the pipe-soil response, suction resistance ratio (relative to penetration resistance at the given embedment), normalised suction decay distance and normalised repenetration offset (which controls the incremental additional embedment with each cycle). These various input parameters are used to define the non-linear hyperbolic functions that model the seabed resistance force as a function of the penetration history as detailed by Randolph and Quiggin (2009). Function parameters are updated each time a penetration reversal occurs allowing the model to capture the hysteretic behaviour of the seabed response and the increasing penetration of the pipe under vertical cyclic loading. No attempt is made to model softening of soil due to remoulding directly, although incremental embedment occurs during load controlled cycles. The model is drawn from a number of research programmes such as reported by Bridge et al. (2004) and has been calibrated against model tests reported by Aubeny et al. (2008). It has also been calibrated against field-scale experiments carried out at Watchet harbour, UK.

Soil shear strength gradient, ρ

The pipe-soil response is closely linked to the shear strength profile, which is assumed to vary linearly with depth in the present implementation of the model. For deepwater applications, it is reasonable to assume that the strength intercept at the mudline, sum , is zero, since any small positive value would rapidly be eliminated due to remoulding. The strength profile is therefore controlled entirely by the rate of change of shear strength with depth (ρ). For foundation and anchor design, strength gradients are typically in the range 1.5 to 2 kPa/m over depths in the 1 to 50 m range (and even deeper). However, many deepwater sediments show shear strengths of 5 to 15 kPa at a depth of about 0.5 m, and hence high strength gradients over that depth range, before the strength gradient reverts to the more typical 1.5 to 2 kPa/m. As such, shear strength gradients for SCR design may range typically from 1.5 to 30 kPa/m. 2.2

Normalised maximum stiffness, Kmax

The normalised maximum stiffness, Kmax , controls the stiffness of the pipe-soil response during initial penetration mode, uplift mode and repenetration mode. It also controls how fast the resistance asymptotically approaches its limiting value. A higher value leads to greater stiffness and translates to increase in fatigue damage. The soil stiffness is given by:

where Pu is the ultimate penetration resistance (force per unit length of riser), z is the penetration of the riser invert below the seabed and D is the riser diameter. The ultimate resistance (net of the buoyancy correction) is expressed as

where Nc is a bearing capacity factor (expressed as a power law function of normalised embedment as Nc = a(z/D)b and su,invert is the shear strength at the current depth of the riser invert. The hyperbolic response linking the initial stiff gradient to the ultimate limiting resistance is expressed as

where Figure 1. Different penetration modes in nonlinear pipe-soil interaction model by Randolph & Quiggin (2009). © 2011 by Taylor & Francis Group, LLC

834

It may be seen that substitution of Eq. 4 into Eq. 3 leads to Eq. 1 as z reduces towards zero, and Eq. 2 as z becomes large. Similar approaches, using changes in penetration, are used for reversals of displacement from penetration to uplift and vice versa. Typical values of normalised maximum stiffness for soft sediments have been shown to lie in range 150 to 250 (Bridge et al. 2004, Clukey et al. 2005).

2.3 Suction ratio, fsuc The suction ratio controls the ultimate suction (or uplift) resistance of the model, which is given by:

It should be noted that in reality the suction resistance will depend on a variety of factors, in particular the rate at which the pipe is lifted up, the length of time over which upward motion is sustained and the recent history of cyclic motion (Bridge et al. 2004); use of a constant fsuc in Eq. 5 is therefore a simplification, reflecting the very limited experimental data currently available. For single uplift motions of a pipe that has been undisturbed for a period, values of fsuc between 0.5 to 1 may be appropriate. However, for fatigue studies or other applications with many cycles of loading, values in the range 0 to 0.3 are recommended (Randolph & Quiggin 2009).

3

Figure 2. General view of the case study in this paper. Table 1.

CASE STUDY

The example SCR configuration used for the parametric studies in this paper is the same configuration adopted by Kimiaei et al. (2010). A general view of the configuration is shown in Figure 2. It comprises a 9 inch (0.228 m) diameter SCR, with 1 inch (25.4 mm) wall thickness, submerged weight of 1.01 kN/m and bending stiffness, EI, of 17.7 MNm2 . The riser is 1600 m long and is hung in 1000 m water depth from the pontoon of a semi-submersible platform. The mean departure angle of the riser is 10◦ as measured from the downward vertical. A flexjoint with stiffness of 10 kNm/deg is incorporated at the connection of the riser to the platform. The riser touchdown point (TDP) is approximately 1170 m arc length measured from the top connection while the TDZ during dynamic motions varies between 1130 m and 1200 m arc length. The soil properties adopted for this study are characteristic of typical soft clay offshore sediments such as in the Gulf of Mexico. The default set of soil model parameters, referred to as ‘base case’ here, are presented in Table 1. In order to investigate the sensitivity of the results to key input parameters, nine different load cases were considered in addition to the base case; in each case only one of the soil parameters (Kmax , fsuc or ρ) were changed, relative to the values for the base case, as presented in Table 2. © 2011 by Taylor & Francis Group, LLC

Soil parameters for the base case.

Parameter

Symbol

Value

Pipe diameter Mudline shear strength Shear strength gradient Saturated soil density Power law parameter Power law parameter Soil buoyancy factor Normalized maximum stiffness Suction ratio Suction decay parameter Repenetration parameter

D sum ρ ρsoil a b fb Kmax fsuc λsuc λrep

0.228 m 0 kPa 1.5 kPa/m 1.5 te/m3 6 0.25 1.5 200 0.2 1 0.3

Table 2.

Load cases used for the parametric studies.

Load case

Symbol

Value

1 2 3 4 5 6 7 8 9

Kmax Kmax fsuc fsuc fsuc ρ ρ ρ ρ

100 300 0 0.4 0.6 3 kPa/m 5 kPa/m 10 kPa/m 20 kPa/m

The analysis software OrcaFlex (Orcina 2009) was used for dynamic time domain response analysis of the SCR. The nonlinear pipe-soil interaction model described previously has been incorporated in OrcaFlex (Randolph & Quiggin 2009). For fatigue damage calculations, deterministic regular fatigue

835

analysis, together with a Palmgren-Miner approach, was used, since this is one of the most widely accepted methods for fatigue design of offshore facilities. In this method, environmental loads on the system were represented by a wave scatter table comprising a number of regular wave packets (giving wave heights, wave periods and total number of wave occurrences). The cyclic stress ranges in the SCR (as relevant for the fatigue analysis in the TDZ) were obtained from the numerical simulations and then, using the standard stress-cycle (S-N) approach, the corresponding fatigue damage for each wave packet was determined. The overall fatigue damage of the system is then obtained from the summation of fatigue damage due to each individual wave packets. The example wave scatter table used in this study, broadly representative of wave data in deepwater Gulf of Mexico, is presented inTable 3. In this study, loading time histories for the riser dynamic response analyses comprised a series of 15 consecutive wave packets selected from Table 3, but with the order varies as detailed later. Note that the wave packets tabulated represent an original set of 30 storms with irregular wave scatter table (covering total exposure time of 20 years). In order to reduce the computational efforts only a 15 wave packet regular wave scatter table was adopted. The main objective here is to provide a relative assessment of how the fatigue life is affected by different orders of waves and variations in key soil properties, rather than obtain absolute estimates of fatigue life. Results of riser fatigue response analyses, which are influenced by nonlinear pipe-soil interaction behaviour, are sensitive to the order of the wave packets in each loading time history (Kimiaei et al. 2010). In this paper, results of only two following sample loading sequences (LS) have been studied:

Table 3. Wave scatter table. Wave packet no.

Wave height (m)

Wave Period (s)

Number of annual wave occurrences

1 2 3 4 6 7 8 10 11 12 14 15 16 18 19

1 1 1 1 3 3 3 8 8 8 13 13 13 18 18

3 8 13 18 8 13 18 8 13 18 8 13 18 13 18

24603799 4584144 770892 288213 1170825 88515 15607 28572 5042 572 868 440 44 63 4

Figure 3. Fatigue damage profiles for different normalised maximum stiffness.

LS1: Loading time history comprising wave packets 4, 8, 12, 16, 19, 3, 7, 11, 15, 18, 2, 6, 10, 14 and 1, sequentially. LS2: Loading time history for each wave packet (1, 2, 3, 4, 6, 7, 8, 10, 11, 12, 14, 15, 16, 18 and 19) separately. Measure fatigue damage for each wave packet in these separate analyses. In each case, the fatigue damage for each wave packet in the loading time history was determined and the total fatigue damage for the system then determined by summation. Note that only sufficient wave cycles were analysed for each wave packet to reach a steady state response (generally 20 cycles) and then the total fatigue damage per year obtained using the stabilised cyclic stress range obtained in the final cycles. 4 4.1

NUMERICAL RESULTS Fatigue results for LS1

Results of 20-year life fatigue damage for LS1 due to different normalized maximum stiffness (Kmax ), suction ratio (fsuc ) and soil shear strength gradient (ρ) over © 2011 by Taylor & Francis Group, LLC

836

Figure 4. Fatigue damage profiles for different suction ratios.

the touch down zone, are shown in Figures 3, 4 and 5, respectively. The figures show consistency of the results in respect of the influence of the three parameters on the SCR fatigue damage. In Figure 3 it is seen that the maximum fatigue damage increases slightly with increase of the maximum normalised stiffness and the location of the maximum fatigue damage is largely

Figure 5. Fatigue damage profiles for different shear strength gradients.

Figure 7. Suction ratio effect on predicted fatigue life.

Figure 8. Soil shear strength gradient effect on predicted fatigue life.

Figure 6. Normalized maximum stiffness effect on predicted fatigue life.

unaffected (remaining close to the initial TDP, which is around 1170 m arc length). Figure 4 shows that the maximum fatigue damage increases as the suction ratio increases, while the location of maximum damage shifts towards the floating platform. For a realistic maximum suction ratio of 0.4, the maximum damage is increased by 50% Figure 5 shows that the maximum fatigue damage increases markedly as the shear strength gradient is increased. It is also observed that as the shear strength gradient increases the location of maximum damage shifts away from the floating platform. Figures 6, 7 and 8 show how the estimated fatigue life of the system is affected by each parameter. Note that the fatigue life is dictated by the most critical section of the riser in the touchdown zone, equivalent to 20 years divided by the maximum 20-year life fatigue damage. From Figure 6 it is clear that the normalised maximum stiffness has no significant effect on the fatigue life of the system. Increasing the normalised maximum stiffness from 100 to 300 only leads to a reduction of approximately 10 years in the predicted fatigue life of the SCR. This translates to only 5% drop in predicted fatigue life. Figure 7 shows a more significant effect of the suction ratio on the SCR fatigue life. Increasing the suction ratio from 0 to 0.6 shows a reduction in fatigue life from around 220 years to 120 years, i.e. a 45% © 2011 by Taylor & Francis Group, LLC

reduction. A similar reduction, from 180 years to 100 years, occurs as the soil shear strength gradient is increased from 1.5 to 10 kPa/m, as shown in Figure 8. The flattening of the curve for shear strength gradients higher than 10 kPa/m suggests that further increase in the shear strength gradient has a more limited effect on the fatigue life. Note that a more realistic upper limit of ∼0.3 on the suction ratio would limit the reduction in fatigue life to about 25% (see Figure 7). 4.2

Effect of loading sequence on fatigue life

The effect of the loading sequence on fatigue life was explored by (a) running LS1 in reverse order and (b) running LS2 (i.e. each wave packet separately). These analyses showed that the original LS1 gave the highest damage, with the fatigue life increased by 10% (reverse LS1) and 15% (LS2). 4.3

Equivalent linear stiffness

The fatigue analyses using the non-linear model gave a fatigue life between 120 and 220 years. It is of interest to note what linear seabed stiffness, k (kN/m/m or kPa) would give a similar fatigue life. This is shown in Figure 9, from which it may be seen that a linear stiffness of 10 to 50 kPa would give the similar range of fatigue life. This is considerably lower than is commonly adopted in fatigue design studies.

837

It is recommended that nonlinear pipe-soil interaction should be taken into account in order to allow reliable, but economic, fatigue design of SCRs.

ACKNOWLEDGEMENTS

Figure 9. Variation of fatigue life with linear seabed stiffness.

5

CONCLUSIONS

The nonlinear soil model recently incorporated in the OrcaFlex software for dynamic analysis of risers was used to conduct a parametric study to investigate the influence of three key soil parameters on the fatigue response of a typical deepwater SCR. The key soil parameters explored were: the normalised maximum stiffness, the suction ratio that determines the maximum uplift resistance, and the soil shear strength gradient. Results of the sensitivity analysis show that: • The normalised maximum stiffness has no signifi-

cant effect on the maximum fatigue damage or its location along the riser. • Increasing the suction ratio or shear strength gradient lead to a decrease in fatigue life. • Fatigue life is more sensitive to the soil shear strength gradient than to the suction ratio. • Increasing the suction ratio shifts the point of maximum fatigue damage towards the floating platform whereas increasing the shear strength gradient shifts it away from the floating platform. It was also found that, while the ordering of the wave packets had some effect on the fatigue life, this was of secondary importance (less than 15%). Analyses using linear seabed stiffness showed that a low stiffness, in the range 10 to 50 kPa, was required to obtain a similar range of fatigue life to that predicted using the nonlinear pipe-soil model. The results from this fatigue parametric study confirm that the soil parameters in the nonlinear soil model can have a significant effect on the SCR fatigue damage in the touchdown zone. However, the fatigue life obtained for typical soil parameters is rather greater than would be obtained using typical linear seabed stiffness values as currently adopted in design.

© 2011 by Taylor & Francis Group, LLC

This work forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established under the Australian Research Council’s Research Centres Program and now supported by the State Government of Western Australia as a Centre of Excellence. The authors would also like to thank Orcina Ltd for their technical support for this study.

REFERENCES Aubeny, C.P. & Biscontin, G. 2009. Seafloor-Riser Interaction Model. Int. J. of Geomechanics, ASCE, 9(3), 133–141. Aubeny, C.P., Gaudin, C. & Randolph, M.F. 2008. Cyclic tests of model pipe in kaolin. Proc. Offshore Technology Conference, Houston, Paper OTC19494. Bridge, C., Laver, K., Clukey, E. & Evans, T. 2004. Steel catenary riser touchdown point vertical interaction models. Proc. Offshore Technology Conference. Houston, Texas, Paper OTC16628. Bridge, C. & Willis, N. 2001. Steel catenary risers – results and conclusions from large scale simulations of seabed interaction, 2H Offshore Engineering Ltd, Woking, Surrey, UK Clukey, E., Haustermans, L. & Dyvik, R. 2005. Model tests to simulate riser-soil interaction effects in touchdown point region. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, ISFOG, Perth, Australia, 651–658. Clukey, E., Ghosh, R, Mokarala, P. & Dixon, M. 2007. Steel catenary riser (SCR) design issues at touch down area, Proc. 17th Int. Conf. on Offshore and Polar Engineering, ISOPE, Lisbon, 814–819. Grealish, F., Kavanagh, K., Connaire, A. & Batty, P. 2007. Advanced nonlinear analysis methodologies for SCRs. Proc. Offshore Technology Conference 2007. Houston, Texas, Paper OTC18922. Kimiaei, M., Randolph, M.F. & Ting, I. 2010. A parametric study on effects of environmental loadings on fatigue life of steel catenary risers. Proc. 29th Int. Conf. on Ocean, Offshore and Arctic Eng., Shanghai, Paper OMAE201021153. Orcina 2009. OrcaFlex User Manual, www.orcina.com, UK. Randolph, M.F. & Quiggin, P. 2009. Non-linear hysteretic seabed model for catenary pipeline contact. Proc. 28th Int. Conf. on Offshore Mech. and Arctic Eng., Honolulu, USA, Paper OMAE2009-79259.

838

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Mobilization distance in uplift resistance modeling of pipelines J. Wang & S.K. Haigh University of Cambridge, UK

N.I. Thusyanthan & S. Mesmar KW Ltd.

ABSTRACT: Upheaval buckling (UHB) of pipelines is a phenomenon by which buried pipelines buckle due to the increased temperature and pressure of operational conditions. UHB is resisted by the resistance of the soil cover. This paper presents a series of experiments designed to investigate the mobilization required for soil cover to provide its peak uplift resistance. It is shown that the DNV-RP-F110 code recommended mobilization is unconservative, leading to overly stiff force-displacement response, especially for higher H/D ratios. As buckling is a stiffness governed behavior, underestimation of peak mobilisation and hence overly stiff force-displacement response will lead to unconservative designs. Based on test results from this research and data available from literature, a new equation in terms of H/D is proposed for predicting peak mobilization.

1

INTRODUCTION

Pipeline networks are instrumental for transporting hot crude oil from offshore platforms to onshore refineries. At shallow water sites (water depth ≤ 15 metres), the trench-and-burial method is typically adopted for pipeline laying projects and the excavated material during trenching is used as primary backfill. One of the main purposes of burial is to preventthe pipeline from heaving upwards, the result of a phenomenon known as upheaval buckling (UHB). UHB is a thermally induced structural effect. The operating conditions of high temperature and large internal pressure, which are significantly above the ambient seabed conditions at first laying, lead to thermal extension along the pipeline. These axial movements are restricted by the frictional forces at the soil-pipeline interface. Large compressive forces are then developed, which can cause the pipeline to buckle globally if lateral restricting forces are inadequate. At locations where the pipeline profile features an over-bend, the most likely buckling mode is for the pipeline to heave upwards through the backfill soil, hence the name upheaval buckling. Once UHB has initiated, additional upward movement of the pipeline would lead to reduced axial compressive force. At the same time, the imperfection curvature would increase, making it easier for the pipeline to buckle. Hence, the stability of the pipeline would depend on the interaction of these two effects. The Current understanding ofuplift resistanceis based on analyses and experimental work (Vesic, 1971; Rowe and Davis, 1982; Hobbs, 1984; Randolph and © 2011 by Taylor & Francis Group, LLC

839

Houlsby, 1984; Trautmann et al., 1985; Palmer et al., 1990; Schaminée et al., 1990; Dickin, 1994; Croll, 1997; Moradiand Craig, 1998; Baumgard, 2000; White et al., 2001; Bransby et al., 2001; and Cheuk, 2005) by researchers on both structural and geotechnical fronts. Most research effort has been directed at identifying the maximum available uplift resistance in granular soils. Little has been concluded, however, on the full uplift force-displacement regime. Small-scale centrifuge tests are widely used as a modeling technique for uplift resistance. Comparison between model and full-scale experiments shows good agreement on the maximum available uplift resistance, but there are significant discrepancies in the dimensionless mobilization displacement (Palmer et al., 2003). Current design guidelines and practice for pipelines are more empirical than analytical. Improved design efficiency would arise from a better understanding of the deformation mechanism during the uplift event and hence a more robust theoretical basis for the prediction of the mobilization of available uplift resistance with pipe upward displacement. A series of both full scale (1 g) and centrifuge (30 g) experiments have been conducted at the Schofield Centre, University of Cambridge, to model the uplift response of pipelines buried in saturated sand. One of the key objectives of this research is to investigate the mobilization distance to peak uplift resistance. The Particle Image Velocimetry (PIV) technique (White et al. 2003) has been employed to reveal the true deformation mechanism involved in the uplift process. This research paper will elaborate on these findings.

Table 1.

Parameters for the DNV tri-linear design curve.

Soil type

Parameter

Range

Loose Sand (H/D range 3.5 to 7.5)

fp δf ∗ α β

∈ ∈ ∈ =

[0.1, 0.3] [0.5%, 0.8%] H [0.75, 0.85] 0.2

Medium/Dense Sand (pre-peak) (H/D range 2 to 8)

fp δf ∗ α β

∈ ∈ ∈ =

[0.4, 0.6] [0.5%, 0.8%] H [0.65, 0.75] 0.2

Rock (H/D range 2 to 8)

fp δf ∗ α β

∈ ∈ ∈ =

[0.5, 0.8] [20 mm, 30 mm] 0.35 D 0.2

Figure 2. Trautmann-Pedersen vertical slip surface model.

where σ configuration is the standard deviation for the survey accuracy of the pipeline configuration and has a minimum value of 0.025 m. However, this minimum value would lead to γUR = 0.925 for cohesionless backfill, i.e. less conservative than without applying this safety factor, which contradicts its original intention. The Trautmann-Pedersen (Pedersen & Jensen, 1988) Vertical Slip Surface Model (Figure 2) and the associated design equation is adopted by DNV to provide an estimate for Rmax :

*: Recommended values from DNV

where: D is the external diameter of the pipe; H is distance between the soil surface and the pipe crown; γ
is the submerged unit weight of the soil; and fp is the dimensionless Pedersen Uplift Resistance Factor. The recommended range of fp values for different types of cohesionless soils is also shown in Table 1. The limited range of H/D ratio for which Equation 2 applies should be noted (Table 1). For design scenarios with H/D ratios below 1.0, the typical current practice is to force fp = 0, and limiting Rmax arbitrarily to the weight of the soil cover alone:

Figure 1. Tri-linear uplift force-displacement model with global safety factor applied.

2

REVIEW ON CURRENT DESIGN PRACTICE

The current industry design practice for uplift resistance modeling is prescribed in Appendix B of the offshore design code, DNV-RP-F110. Description of soil models for both cohesive (clay) and cohesionless (sand and rock) soils are given. This research has concentrated on the latter soil type only. The DNV design code states that the characteristic response of cohesionless soils to UHB can be approximated by a normalized tri-linear uplift forcedisplacement curve (Figure 1), with Rmax and δf being the maximum available uplift resistance per meter along the pipeline and the corresponding mobilization distance respectively. The geometry of this tri-linear characteristic curve can be accurately defined by three parameters: α, β, and δf .The prescribed ranges of these parameters for different types of cohesionless backfill are summarized in Table 1. To obtain the more conservative design curve, a global safety factor, γ UR , must be applied as shown in Figure 1. For cohesionless soils, this global safety factor, γ UR , is given by:

The tri-linear design curve can then be applied. This extra-conservatism at low H/D ratios can lead to large quantities of rock dump material being required as secondary backfill, which can cost millions of dollars. The major reason behind this conservatism is the absence of available data at these low H/D ratios. Recent research (Wang et al., 2010; Thusyanthan et al., 2010) suggests that Equation 2 still provides a good estimate for Rmax at H/D ratios below 1.0 in both sand and rock. 3

RESEARCH OBJECTIVES

A series of full-scale pipeline uplift resistance tests have been devised and conducted at the Schofield © 2011 by Taylor & Francis Group, LLC

840

Table 2.

Figure 3. Full-scale plane-strain test tank in (a) schematic plot, and (b) photograph of complete set-up.

Centre, University of Cambridge. The test series can be further divided into two categories: 1. Plane-strain testing in saturated sand 2. Full-scale field testing in moist sand

Test No.

Backfill

H/D

Pipe Diameter (mm)

1 2 3 4 5 6 7 8 9a 9b

Sand Sand Sand Sand Sand Sand Sand Sand Sand Sand

0.1 0.4 0.5 1 2 3.5 0.5 1.0 6 8

100 100 100 100 100 100 258 258 200 200

(W) × 850 mm (H). More details of the test tank can be found in Wang et al., (2010). Two model pipes with external diameters of 100 mm and 258 mm with PTFE front and back faceswere manufactured. The PTFE material has a very low friction angle, which minimizes the end effects in 2D plain strain modeling. Either pipe will be connected to the actuator via a 12 mm diameter aluminum rod. The cross sectional area of the rod represents 1.03% of the projected area of the 100 mm diameter model pipe and 0.40% of the 258 mm diameter model pipe. Hence its effect on the measured uplift resistance is negligible. Fraction E sand of relative density ID = 35% (loose), median diameter D50 of 0.15 mm and saturated unit weight γ sat of 18.5 kN/m3 was used as backfill. The Particle Image Velocimetry (PIV) technique was employed using a single Canon G10 digital camera to capture the displacement field of the backfill around the pipe throughout the pull-out process. Tests No. 9a and 9b were conducted at full-scale in the field via vertical hydraulic jacking. Fine sand with properties very similar to that of Fraction E Sand was used. The measured moisture content of the soil was approximately 4.6% and the bulk unit weight 15 kN/m3 . The model pipe used had an external diameter of 0.2 m and a length of 1 m.

The principal objective of these tests is to understand how mobilization distance varies with H/D ratios in cohesionless soils. Another parallel objective is to understand the reliability of the shear component of the “true” uplift resistance in cohesionless soils at H/D ratios below 2, with a particular focus on H/D ratios below 1. Detailed results and conclusions on this aspect are explained in Wang et al., (2010).

4 APPARATUS AND TEST PROGRAM The test program for the entire test series is summarized in Table 2. Figure 3 illustrates the full-scale plane-strain test tank used for Tests No. 1 to 8. The container has internal dimensions of 1000 mm (L) × 76 mm © 2011 by Taylor & Francis Group, LLC

Full-scale test program for uplift resistance in sand.

841

5

RESULTS AND DISCUSSION

Representative uplift force-displacement raw data are illustrated in Figure 4. This data can be normalized using the DNV approach so as to be compared directly with the recommended tri-linear design curve. This is illustrated in Figure 5. Figure 5 illustrates that, if normalized with measured values for mobilization distance δf , the DNV upper and lower bound normalized plots for loose sand are reasonably representative of the experimental data. However, the measured δf values are very different from the DNV recommendations of the mobilization distances being between 0.5% H and 0.8% H. Comparison between the measured and the DNV recommended values for δf is summarized in Table 3.

Table 3. values.

Summary of actual and DNV recommended δf DNV Recommendation (mm)

Test No.

H (mm)

Measured δf (mm)

0.5 % H

0.8% H

1 2 3 4 5 6 7 8 9a 9b

10 40 50 100 200 350 129 258 1200 1600

2.5 3.0 2.5 3.3 4.6 10 10 8 110 215

0.05 0.2 0.25 0.5 1.0 1.75 0.65 1.29 6 8

0.08 0.32 0.4 0.8 1.6 2.8 1.0 2.1 9.6 13

Figure 4. Representative uplift force-displacement data.

Figure 6. Summary of normalized mobilization distance.

Figure 5. DNV characteristic response compared with representative:(a) plane-strain test data; and (b)field test data, normalised by measured δf instead of values suggested by DNV.

© 2011 by Taylor & Francis Group, LLC

It is apparent that the DNV code vastly underestimates the mobilization distance required to reach Rmax . In addition, DNV suggests that δf is only a function of H and independent of the H/D ratio, hence the diameter of the pipe should not affect δf if H remains constant. According to Table 3, this does not seem to hold: The cover heights in Tests No. 4 (D = 100 mm) and 7 (D = 258 mm) only differ by 29 mm, but δf values differ by 200%. Thusyanthan et al. (2010) suggests that a good correlation can be obtained by normalizing the mobilization distance with pipe external diameter and plotting this dimensionless mobilization, δf /D, against the H/D ratio on a log-linear scale. This correlation is shown in Figure 6, using experimental results from Tests No. 1 to 9 as well as other available uplift resistance data in loose sand from literature. A linear trend line is evident from Figure 6 which can be described by Equation (4). The latest full-scale uplift tests in rock exhibit similar linear trends with gradient apparently dependent upon grain size.

842

strain originate from both edges of the pipe crown, and start to propagate almost vertically towards the soil surface. Newborn compression fronts start to converge on and merge into the two macroscopic shear bands. Rmax is usually reached when this macroscopic shear band just reaches the soil surface. 3. At post-peak displacements (δ > 4.6 mm in this case), the existing mechanism reinforces itself, and the two macroscopic shear bands start to move sideways and widen in a very gradual manner. The widths of both the smaller compression fronts and the bigger macroscopic shear bands seem independent of the H/D ratio. At peak uplift resistance, the centre lines of the two shear bands coincide very well with the two shear planes specified in the Vertical Slip Surface Model. However, their finite but significant width suggests that this model is at most an approximation to the true uplift deformation mechanism in loose sand at these H/D ratios. PIV strain analysis also provides useful insight into the width of the influence zone above the soil surface during the uplift event. In loose sand, the width of this influence zone seems to be approximately 2.5 times the pipeline external diameter. Hence if additional downward force is to be provided by rock dump, the width of dumping should be of similar dimensions to ensure maximum efficiency. Figure 7. Evolution of total shear strain for Test No. 5.

7 6

CONCLUSIONS

DEFORMATION MECHANISM

For Tests No. 1 to 8, the soil displacement field during the uplift event was accurately measured at 5-second intervals, which corresponds to 0.025 mm of upward displacement by the model pipe. This was achieved using the non-contact digital image correlation technique of particle image velocimetry (PIV), described in detail by White et al. (2003). As an example, the evolution of soil shear strain with pipe upward displacement for Test No. 6 (H/D = 2) is illustrated in Figure 7. It is clearly visible that the uplift mechanism in loose sand can be divided into three phases: 1. At very small displacements (δ < 1 mm in this case), thin strands of compression fronts (interpreted as shear bands in total shear strain plots) originate from one side of the pipe crown, fanning out gradually to the other side and swiftly propagating through the backfill soil medium. Localized dilation shear zones start to appear underneath the pipe, which ultimately creates a wedge-shaped void. 2. At small pre-peak displacements (1 mm < δ < 4.6 mm in this case), propagation becomes slower and slower and almost comes to a standstill when these compression fronts have rotated by over 90◦ . Subsequent compression fronts start to superimpose on their predecessors. Two cumulated macroscopic shear bands of more than 5% total shear © 2011 by Taylor & Francis Group, LLC

This paper presented the results of full-scale tests on the upheaval buckling resistance of pipelines in loose saturated sand in an attempt to clarify the mechanics of the pipe-soil interaction. The results were compared with behaviour suggested by the DNV-RP-F110 design code which was shown to be non-conservative in estimating mobilization distance. On the other hand, Wang et al., (2010) shows that the DNV code is conservative in estimating the maximum available uplift resistance. Buckling is a stiffness dominated process, so the most important parameter in determining whether or not a strut or pipeline will buckle is the stiffness of the restraining “spring”. The mobilization distance predicted by the DNV code underestimates measured values by factors between 5 and 50, providing a non-conservative design approach. Equation 4 gives an expression for mobilization distance that is a function of both cover H and pipe diameter D.

ACKNOWLEDGEMENT The authors would like to thank all staff at the Schofield Centre, University of Cambridge for their help and advice throughout the testing program. The first author would also like to thank Trinity College, University of Cambridge, and KW Ltd. for

843

their generous financial support towards this research effort. REFERENCES Baumgard, A.J. (2000). Monotonic and cyclic soil response to upheaval buckling in offshore buried pipelines. PhD Thesis.University of Cambridge. Bransby, M.F., Newson, T.A., Brunning, P., and Davies, M.C.R. (2001). Numerical and centrifuge modelling of upheaval resistance ofburied pipelines. Proc. OMAE, Rio de Janeiro, June 2001. Cheuk, C.Y. (2005). Soil pipeline interaction at the seabed.PhD thesis.University of Cambridge. Croll, J.G.A. (1997). A simplified model of upheaval thermal buckling of subsea pipelines.Thin-Walled Structures 29(1-4): 59–78. Dickin, E.A. (1994). Uplift resistance of buried pipelines in sand. Soils and Foundations 34(2): 41–48. DNV-RP-F110, Global buckling of submarine pipelines – structural design due to high temperature/high pressure. Det Norske Veritas, Norway, 2007. Hobbs, R. (1984). In service buckling of heated pipelines.ASCE Journal of Transportation Engineering 110(2): 175–189. Moradi, M. and Craig, W.H. (1998). Observations of upheaval buckling of buried pipelines. Centrifuge 98, Kimura, Kusakaba&Takemura(eds), ISBN 90 5410 986 6 Palmer, A.C., Ellinas C.P., Richards, D.M., and Guijt, J. (1990).Design of submarine pipelines against upheaval buckling.Proc. Offshore Technology Conf., Houston, OTC 6335: 551–560. Palmer, A.C.White, D.J., Baumgard, A.J., Bolton, M.D., Barefoot, A.J.Finch, M., Powell, T., Faranski, A.S., Baldry, J.A.S. (2003). Uplift resistance of buried submarine pipelines: comparison between centrifuge modeling and full-scale tests. Géotechnique 53(10): 877–883.

© 2011 by Taylor & Francis Group, LLC

Pedersen, P.T. & Jensen, J.J. (1988).Upheaval creep of buried pipelines with initial imperfections. Marine Structures 1:11–22, 1988. Randolph, M. F., &Houlsby, G. T. (1984). The limiting pressure on a circular pile loaded laterally in cohesive soil. Géotechnique, 34(4): 613–623. Rowe, R.K., and Davis, E.A. (1982).The behaviour of anchor plates in sand.Géotechnique 32 (1): 25–41. Schaminée, P.E.L., Zorn, N.F., and Schotman, G.J.M. (1990). Soil response for pipeline upheaval buckling analysis: Full-scale laboratory tests and modelling. Offshore Technology Conference, Houston, OTC 6486 Thusyanthan, N.I., Mesmar, S., Wang J., and Haigh, S.K. (2010). Uplift resistance of buried pipelines and DNV-RPF110 guideline. Proc. Offshore Pipeline and Technology Conference.Feb 24–25, Amsterdam, Netherlands. Trautmann, C.H., O’Rourke, T.D., and Kulhawy, F.H. (1985). Uplift force-displacement response of buried pipe. ASCE Journal of Geotechnical Eng. Division 111(9): 1061–1075. Vesic, A.S. (1971). Breakout resistance of objects embedded in ocean bottom. ASCE Journal of the Soil Mechanics and Foundation Division.97 (9): 1183–1205. Wang, J., Ahmed, R., Haigh, S.K., Thusyanthan, N.I., and Mesmar, S. (2010). Uplift resistance of buried pipelines at low cover-diameter ratios. Proc. Offshore Technology Conference.May, 2010, Houston, USA. White, D.J., Barefoot, A.J., Bolton, M.D. (2001). Centrifuge modelling of upheaval buckling in sand. International Journal of Physical Modelling in Geotechnics, 2(1): 19–28. White, D. J., Take, W. A. & Bolton, M. D. (2003). Soil deformation measurement using particle image velocimetry (PIV) and photogrammetry.Géotechnique,53(7): 619–631.

844

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Theoretical, numerical and field studies of offshore pipeline sleeper crossings Z.J. Westgate, M.F. Randolph & D.J. White Centre for Offshore Foundation Systems, University of Western Australia, Perth

P. Brunning Acergy, Singapore

ABSTRACT: Offshore pipelines experience axial stresses due to internal pressure and thermal cycles during start-up and shut-down, leading to the formation of lateral buckles. Pipeline lay routes often include strategicallyspaced transverse sleepers, creating a vertical imperfection from which a controlled lateral buckle can be initiated. The as-laid pipeline embedment affects the pipe-soil interaction forces and therefore the buckle initiation response. The potential for increased embedment in the touchdown zones to either side of the sleeper exists, which can cause higher than expected lateral breakout forces. The magnitude of this embedment is difficult to quantify due to dynamic lay effects and changes in catenary forces as the pipe is laid over the sleeper. Theoretical solutions for the touchdown force at a sleeper crossing are presented for the case of an elastic seabed, and are compared to as-laid survey data in soft clay. Static and dynamic numerical analyses are also presented to illustrate the changes in pipe-soil contact force as a pipe is laid across a sleeper. This provides a rationale for asymmetry observed in as-laid embedment profiles and its influence from dynamic lay effects. General guidance is provided for pipeline designers to assist the assessment of pipe-soil interaction forces in the vicinity of sleepers.

1

INTRODUCTION

As offshore hydrocarbon developments have progressed into deeper waters, new pipeline design issues have arisen, such as lateral buckling, which occurs due to axial pipe stresses during thermal cycles of start-up and shut-down. This has led to development of lateral buckle mitigation techniques, such as sleepers. Sleepers allow controlled buckles to occur in predetermined locations along the route by creating a vertical imperfection on which the pipeline slides laterally (Sinclair et al. 2009). In deep water developments where soft fine-grained soils are prevalent, the lay process induces significant pipeline embedment due to dynamic lay effects (Randolph & White 2008).As-laid field surveys (Lund 2000, Westgate et al. 2010) show that pipeline embedment during normal lay conditions in the field can be up to an order of magnitude greater than the predicted static pipeline embedment based on the intact soil strength. Using the remoulded strength in this calculation has been shown to predict embedment closely matching field observations (Westgate et al. 2010). Pipeline embedment influences the lateral breakout resistance provided by the seabed soil. Together with dynamic lay effects, the variation in pipe-soil contact force across a sleeper influences the magnitude of pipeline embedment. This can lead to differences in the lateral breakout resistance along the touchdown © 2011 by Taylor & Francis Group, LLC

zones to either side of the sleeper, affecting the pipeline stresses along the buckle. This study illustrates the variation in pipe-soil contact force in the vicinity of sleepers, and the dependence of the resulting pipeline embedment on the asymmetric and dynamic nature of the lay process. The paper presents results of theoretical analyses used to calculate the pipe-soil contact force and embedment based on the standard catenary solution, and static and dynamic numerical analyses of the lay process. These are compared to as-laid field data for a pipeline installed at a soft clay site in deep water. 2 THEORETICAL ANALYSES Theoretical methods for calculating pipe-soil contact forces during pipe laying are well-established (Lenci & Callegari 2005, Palmer 2008). The governing equation for the response of a pipe being laid on an elastic seabed can be written as:

845

where E = elastic modulus, I = second moment of area, T0 = horizontal pipe tension, p = submerged pipe weight, k = seabed soil stiffness, w = pipe embedment, and x = the distance along the pipeline.

equation 1 to be solved. It is worth noting that for the extreme (though unrealistic) case of T0 = 0, the character of the solution to equation 1 changes from a catenary to a beam solution. Several variations of pipe lay over a sleeper were analysed, increasing in realism and computational effort (Table 1). Table 2 lists the pipeline properties and lay conditions for these analyses. These correspond to a pipeline for which the lay conditions and as-laid embedment is known (Case 8). 2.2

Figure 1. Idealised sleeper crossing (vertical scale exaggerated for clarity).

The horizontal pipe tension is calculated based on the standard catenary solution as:

where φ = the lay angle, defined as the inclination to the horizontal at the lay ramp departure point, and zw = the water depth. The horizontal component of tension is constant along the suspended pipeline. The pipe catenary creates a vertical force concentration at the touchdown zone expressed as the pipe-soil contact force V , normalised by the submerged pipe weight p. The ratio V /p is related to the characteristic length λ = (EI/T0 )0.5 , after Pesce et al. (1998). 2.1 The sleeper crossing problem The as-laid sleeper crossing problem can be idealised as a beam on an elastic seabed due to the small vertical deformation of the pipeline in the vicinity of the sleeper (Figure 1). For the case of a pipeline laid from a vessel, the maximum force concentration Vmax /p in the touchdown zone may be approximated as (Randolph and White 2008):

For the case of a pipeline lowered from a horizontal plane above a flat seabed under zero tension, i.e. ‘placed’, Vmax /p = 1 everywhere along the pipeline due to the absence of the catenary. The presence of a sleeper of height h above the seabed complicates this condition due to the additional boundary conditions imposed (Figure 1). Visual inspection shows that deflection, slope, shear and moment are continuous at x = 0, with deflection w = 0. Similarly, at x = L, w = − h and slope = 0, which allow © 2011 by Taylor & Francis Group, LLC

846

Static pipe lay and placement on linear seabed

Figure 2 shows the variation in V /p for the zero tension, pipe placement condition (Case 1) and the as-laid catenary tension of the real case (Case 2). For this geometry and tension there is little difference between the as-laid catenary case with tension and the zero tension placement case, with a maximum V /p equal to 1.38 and 1.5 respectively. The adopted stiffness of k = 20 kPa for the elastic solutions is consistent with the theoretical pipe penetration prediction of Merifield et al. (2009) for an intact shear strength gradient of su = 20 kPa/m (negligible mudline strength intercept) as applicable to the field case discussed later. This stiffness gave a nominal embedment due to the pipe weight p of w/D = 0.09 (i.e. w = 0.09 pipe diameters) and a maximum embedment due to the catenary contact force V of w/D = 0.13. Reducing the seabed stiffness in the theoretical solution represents a case where the seabed has softened due to dynamic lay effects. A reduced stiffness of 8 kPa represents a fourfold drop to remoulded conditions, based on a remoulded shear strength gradient of 5 kPa/m as applicable to the field case. This decreases the maximum force concentration by ∼10% so the nominal and maximum embedment values increase by less than fourfold to w/D = 0.22 and w/D = 0.27 respectively (Figure 3). These solutions represent symmetric sleeper crossing cases, with the same maximum contact force and embedment on each side of the sleeper. However, the lay process is asymmetric, as illustrated conceptually in Figure 4. As the pipe is laid from the lay vessel, it approaches the sleeper with a catenary configuration independent of the sleeper’s presence (dotted line). On the trailing side of the sleeper, the pipe lifts off of the seabed as it contacts and rotates over the sleeper (due to the pipe stiffness), causing the touchdown point to move away from the sleeper through the ‘uplift zone’ (dashed line to left of sleeper). On the leading (i.e. vessel) side of the sleeper, the additional weight of the suspended pipe causes the pipe to embed further compared to the nominal pipe catenary force (dashed line to right of sleeper). This load-unload mechanism results in a similar uplift zone on the leading side of the sleeper. The end result is an asymmetric embedment profile on a real (plastic) seabed (heavy solid line) which exceeds that of an idealised (elastic) seabed (light solid line).

Table 1.

List of analyses.

1 2 3 4 5 6 7 8

Theoretical Theoretical Theoretical Numerical Numerical Numerical Numerical Empirical

Table 2.

Pipeline properties and lay conditions.

Placement Static lay Static lay Static lay Static lay Static lay Dynamic lay Real pipe lay

Linear/intact Linear/intact Linear/remoulded Linear/intact Non-linear/intact Non-linear/remoulded Non-linear/intact Real soil

Parameter

Idealised

Field case

Outside diameter, D (m) Steel thickness (mm) Bending rigidity, EI (MNm2 ) Coating thickness (mm) Submerged pipe weight, p (kN/m) Water depth, zw (m) Lay angle, φ (deg) Horizontal pipe tension, T0 )kN) Sleeper height, h (m) Significant wave height, Hs (m)

0.32 19 44 0 0.57 1300 83 103 0.9 2

0.32 19 44 2.6 0.58 1240–1310 82.6 106–112 0.9–1.0 0.7–2.4

Zero As-laid As-laid As-laid As-laid As-laid As-laid As-laid

Increasing realism and computational effort

Figure 3. Theoretical analysis for remoulded seabed (Case 3).

For a horizontal tension of 103 kN, the catenaryinduced contact force is 1.73 and the final sleeper-induced contact force is 1.38, a reduction of 20%. For the final pipe-sleeper configuration, the maximum V /p reduces towards a value of 1.5 at zero tension. At this point, the maximum V /p for the catenary is infinite, but rapidly reduces for more realistic values of T0 /λp. At high normalised tension values, both conditions converge to V /p = 1. Figure 2. Theoretical analyses for intact seabed (Case 1 and 2).

An analysis of the lay process on a non-linear (plastic) seabed will show this asymmetry since the overloading history will force irreversible embedment. However, the elegant elastic cases allow trends related to the lay tension and seabed stiffness to be explored – in particular the variation in force concentration in the touchdown zones.

2.4

Figure 6 shows V /p as a function of the normalised seabed stiffness K = λ2 k/T0 . As the seabed stiffness increases, both the absolute V /p values and the ratio of V /p between the two cases increase. As the stiffness reduces, V /p converges to unity. Normalised stiffness values for soft clay seabeds are typically in the range of K = 100–1000, but for very weak remoulded soft clays K can be lower, and for stiffer clays K can exceed 10,000.

2.3 Effects of pipe tension

3

Figure 5 shows V /p as a function of the normalised pipe tension T0 /λp, for a seabed stiffness of k = 20 kPa. The temporary pipe-soil contact force from the pipe catenary is always greater than the final contact force after the pipeline has crossed the sleeper, with the difference increasing with reducing pipe tension. © 2011 by Taylor & Francis Group, LLC

Effects of seabed stiffness

FIELD STUDIES

Variations in pipeline embedment across sleepers were obtained from as-laid field survey data from a development in deep water with a soft clay seabed. The intact undrained shear strength gradient in the upper 0.5 m of the seabed is about 20 kPa/m, with a remoulded

847

Figure 7. As-laid field survey embedment profiles compared to as-laid theoretical solutions (Cases 2 and 3).

Figure 4. Sleeper crossing lay process (vertical scale exaggerated for clarity).

rotational motion, thus minimising effects of the vessel motions on the pipe-soil contact forces. Profiles of the as-laid normalised pipeline embedment w/D across the sleepers (Case 8) are shown in Figure 7, compared to the as-laid theoretical solution for the intact and remoulded strength gradients (Cases 2 and 3). The mean embedment for this pipeline away from the sleepers was 0.31D. At the sleeper crossings, the embedment ranged from about 0.2D to 0.5D. The leading side of the sleeper exhibited slightly deeper embedment consistent with the lay process discussed in Section 2. The as-laid embedment data showed that the height of the sleeper crown h above the seabed was between 0.9 and 1.0 m, which affects the length of the hanging spans L on each side of the sleeper (Figure 1). The theoretical solution for the intact seabed (Case 2) exhibits the correct shape of the embedment response, but significantly under predicts the magnitude of the embedment and over predicts the length of the hanging spans. If the remoulded strength is used in the theoretical solution (Case 3), the embedment and span length are closer to the field values, but the asymmetry is absent.

Figure 5. Influence of horizontal pipe tension on maximum pipe-soil contact forces.

4

Figure 6. Influence of seabed stiffness on maximum pipe-soil contact forces.

The dynamic riser analysis software, OrcaFlex (Orcina 2009), was used to explore the trends shown in the field data. The first numerical analysis (Case 4) comprised static pipe lay on a linear elastic seabed. Further analyses (Cases 5, 6, 7) show the influence of non-linear seabed stiffness, a reduced strength from remoulding and vessel motions respectively. The idealised pipe properties used in the numerical model are those used in the theoretical analyses (Table 2).

strength gradient of 5 kPa/m. The properties and lay conditions for the surveyed pipeline are summarised in Table 2. The pipeline was laid from a J-lay vessel with a lay ramp that permitted 3 degree-of-freedom © 2011 by Taylor & Francis Group, LLC

NUMERICAL ANALYSES

848

4.1

Static pipe lay on linear seabed

Figure 8 shows the variation in V /p and w/D for the linear seabed (Case 4). The solid line shows the final profile, which matches the theoretical solution

Figure 8. Variation in pipe-soil contact force and embedment for static pipe lay on linear intact seabed (Case 4).

from Case 2. The dashed line shows the maximum (temporary) profile exhibited during the lay process (Figure 4). As the pipe catenary approaches the sleeper, the force concentration factor is constant at V /p = 2.1, with a corresponding embedment w/D = 0.16. The final embedment reduces (elastically) to w/D = 0.09, corresponding to the submerged pipe weight. After making contact with the sleeper, V /p reduces as the pipe weight is redistributed to the sleeper and eventually to the seabed on the leading side of the sleeper. The increase in contact force on the leading side (due to the heavier catenary) is 20% higher than that on the trailing side, with a maximum V /p = 2.7. The pipe embedment for this force is w/D = 0.2. Once the sleeper crossing is completed, the maximum force concentration factor returns to the nominal value of V /p = 2.1, for which w/D = 0.16. For the elastic seabed, the increased V /p has no effect on the final embedment. Although an asymmetric final profile cannot be predicted by the elastic seabed model, the progressive laying simulation of Case 4 shows a maximum V /p profile that is consistent with the asymmetry in the field data of as-laid embedment (Figure 7).

Figure 9. Variation in pipe-soil contact force and embedment for static pipe lay on non-linear seabed (Cases 5 and 6).

uplift (transient tensile) zone is evident on both the trailing and leading sides of the sleeper, as discussed in Section 2. The final pipe embedment is close to the maximum embedment profile due to the plastic seabed deformation, and captures the asymmetry well. A second static non-linear case was carried out for the remoulded shear strength gradient of 5 kPa/m (Case 6). The final embedment profile for this case is also shown in Figure 9. The weaker penetration resistance due to the lower strength gradient results in deeper pipe embedment. This final ‘remoulded’ embedment is closer to the mean embedment profile observed in the field and captures the asymmetry.

4.2 Static pipe lay on non-linear seabed A non-linear seabed with the intact shear strength gradient of 20 kPa/m (mudline strength intercept of zero) was analysed for static pipe lay. The non-linear soil model in OrcaFlex has a stiff unloading response following penetration, thus capturing the effect of a previous overloading event (Randolph & Quiggin 2009). The model also captures increasing pipeline penetration with cycles of vertical loading, simulating the effects of soil softening. Default soil model parameters were used for the static analysis, which were based on intact strength profiles. Figure 9 shows the range in V /p (upper plot) and w/D (lower plot) for static pipe lay across the sleeper on a non-linear intact seabed (Case 5). Also shown is the final force concentration and embedment. An © 2011 by Taylor & Francis Group, LLC

4.3

Dynamic pipe lay on non-linear seabed

Vessel-induced pipe motions result in higher transient contact forces, which (for the non-linear plastic seabed) lead to greater penetration as well as incremental pipe embedment with each cycle. A dynamic analysis (Case 7) was carried out by adding a regular (Dean stream) significant wave height of Hs = 2 m with a wave period of 13 seconds (to match the field conditions) to the static non-linear Case 5.The vessel and pipe payout advanced at 0.1 m/s, i.e. a lay rate of 360 m/hr. This lay rate corresponds to the time period between welding operations, i.e. the minimum number of cyclic pipe motions in the touchdown zone. The intact soil strength gradient of 20 kPa/m was adopted as the non-linear soil model accounts for

849

zone, dictated by the lay rate, sea state, and vessel and pipeline dynamics. These analysis techniques can be used in pipeline design to determine the range of embedment likely to occur near sleepers, aiding assessment of the pipe-soil response during buckling. ACKNOWLEDGEMENTS

Figure 10. Variation in embedment for dynamic pipe lay on non-linear seabed (Case 7) compared to field survey data.

incremental penetration under cyclic motions. Inspection of individual nodes along the pipeline indicated that the force – penetration response reached near steady-state conditions within the 15 to 20 cycles of pipe motions to which it was subjected. The analysis is therefore not sensitive to the range of lay rate recorded during normal laying operations. This represents a touchdown zone length of about 23 m, which is realistic for this seabed, pipe and lay geometry. The non-linear soil model parameters were then optimised to match the shape of the field data profiles by reducing the non-dimensional maximum suction resistance factor and increasing the non-dimensional repenetration resistance depth factor (Randolph & Quiggin 2009). Figure 10 shows the final embedment profile from this analysis, which provides the best match to the field data. This example illustrates how soil non-linearity and the dynamic aspects of pipeline embedment can be simulated. 5

CONCLUSIONS

This study has presented a series of theoretical and numerical analyses showing the variation in pipe-soil contact force and the resulting pipeline embedment in the vicinity of sleeper crossings along offshore pipelines. The results have been compared to field survey embedment data for a pipeline installed at a soft clay site in deep water. The asymmetry within the pipe-soil contact force profiles across sleepers can only be captured realistically using non-linear seabed models that account for the loading history of the pipe-soil contact force during the sleeper crossing. Static pipe lay analysis using nonlinear seabed models and a remoulded shear strength to account for dynamic lay effects provided an embedment profile that generally matched the trends in the field data. A marginally closer match to field observations was obtained using an optimised dynamic pipe lay analysis that simulated incremental penetration due to cycles of vertical pipe movement in the touchdown

© 2011 by Taylor & Francis Group, LLC

This work forms part of the activities of the Centre for Offshore Foundation Systems, established under the Australian Research Council’s Research Centres Program and now supported by the State Government of Western Australia through the Centre of Excellence in Science and Innovation program. The work was carried out while the primary author was an Endeavour International Postgraduate Research Scholar. The second author is supported by an ARC Federation Fellowship (grant FF0561473). The third author is supported by an ARC Future Fellowship (grant FT0991816). This study is part of a COFS-Acergy research collaboration. Orcina Ltd, UK, assisted in developing a model to simulate pipe lay in OrcaFlex. REFERENCES Lenci, S. & Callegari, M. 2005. Simple analytical models for the J-lay problem, Acta Mechanica, Vol. 178: 23–39. Lund, K.M. 2000. Effect of Increase in Pipeline Soil Penetration from Installation, Proc. ETCE/OMAE2000 Joint Conf., Paper OMAE2000-PIPE5047. Merifield, R., White, D.J. & Randolph, M.F. 2009. The effect of soil heave on the response of partially-embedded pipelines in clay, ASCE J. of Geotechnical and Geoenvironmental Eng., 135(6): 819–829. Orcina. 2009. OrcaFlex User Manual, Version 9.2e, www.orcina.com.uk. Palmer, A. 2008. Touchdown indentation of the seabed, Applied Ocean Research, 30(3): 235–238. Pesce, C.P., Aranha, J.A.P. & Martins, C.A. 1998. The soil rigidity effect in the touchdown boundary layer of a catenary riser: Static problem, Proc. 8th Int. Offshore and Polar Eng. Conf., 207–213. Randolph, M.F. & Quiggin, P. 2009. Non-linear hysteretic seabed model for catenary pipeline contact, Proc. 28thInt. Conf. on Offshore Mech. and Arctic Eng., Paper OMAE2009-79259. Randolph, M.F. & White, D.J. 2008. Pipeline Embedment in Deep Water: Quantification and Processes, Proc. Offshore Tech. Conf., Paper OTC19128. Sinclair, F., Carr, M., Bruton, D. & Farrant, T. 2009. Design challenges and experience with controlled lateral buckle initiation methods, Proc. 28th Int. Conf. on Offshore Mech. and Arctic Eng., Paper OMAE2009-79434. Westgate, Z.J., White, D.J. & Randolph, M.F. 2010. Pipeline laying and embedment in soft fine-grained soils: field observations and numerical simulations, Proc. Offshore Tech. Conf., Paper OTC20407.

850

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE® D.J. White Advanced Geomechanics and Centre for Offshore Foundation Systems (COFS), Univ. of Western Australia

A.J. Hill BP Exploration, Sunbury-on-Thames, UK

Z.J. Westgate Advanced Geomechanics and Centre for Offshore Foundation Systems (COFS), Univ. of Western Australia

J-C. Ballard Fugro Engineers SA, Brussels, Belgium

ABSTRACT: The Fugro SMARTPIPE® is a new site investigation tool to measure pipe-soil interaction parameters in situ, at the seabed. It comprises a seabed frame with an instrumented model pipe that can be driven in the vertical, axial and lateral directions. This paper presents results from the first SMARTPIPE® campaign in deep water, focusing on axial pipe-soil interaction on soft clay. Pore pressure measurements on the surface of the test pipe provide data that allows the cyclic axial response to be interpreted in an effective stress framework. An effective stress failure criterion shows better agreement with the response across several axial sweeps than a total stress interpretation. In the example test presented here, transient positive excess pore pressure is generated when the pipe changes direction. This means that in this case, surprisingly, the full axial resistance is mobilised over a particular time rather than a particular distance.

1

INTRODUCTION

The Fugro SMARTPIPE® is a new site investigation tool that is designed to measure pipe-soil interaction parameters in situ, at the seabed. It comprises a seabed frame with an instrumented model pipe that can be driven in the vertical, axial and lateral directions whilst the corresponding loads are recorded. Descriptions of the original design of the SMARTPIPE® are given by Hill & Jacob (2008). A key motivation for the development of the SMARTPIPE® is the need to better quantify the interaction forces between on-bottom pipelines and the seabed in order to provide geotechnical input for the assessment of pipeline buckling and walking. The first offshore SMARTPIPE® field testing campaign was carried out at a deep water location during 2008. This paper describes some of the results from that campaign, focussing on axial pipe-soil interaction. Pipe-soil resistance forces depend on properties of both the soil (including the drainage conditions, and the relevant undrained or drained strength) and the pipe (including the embedment, diameter and surface coating). The SMARTPIPE® should be viewed as a model test. Through interpretation, this can yield information about the soil properties, and the response of a given pipe resting on that soil. © 2011 by Taylor & Francis Group, LLC

Just as the SMARTPIPE® is an evolving tool (the equipment has been enhanced since the campaign described here), so are the techniques to analyse, interpret and predict pipe-soil interaction. This paper provides insights into the best framework in which to interpret axial pipe-soil interaction. 2

DESCRIPTION OF THE SMARTPIPE®

The SMARTPIPE® can be deployed via a single lifting/communication cable, from vessels equipped with the equivalent of a stern-mounted 20 tonne capacity A-frame at least 5 m wide and 7 m tall. The maximum operational water depth is currently 2,500 m. Figure 1 shows a diagram of the SMARTPIPE®, as used in this deployment. The key dimensions of the model pipe are summarised in Table 1. The central measurement section of the SMARTPIPE® model pipe is attached to the main frame of the device by a pair of triaxial load cells, which indicate the vertical, axial and lateral loads applied to the measurement section (Figure 2). The measurement section is also equipped with a set of pore pressure transducers (PPTs) which record the pore water pressure at the surface of the pipe relative to a hydrostatic reference. There are five PPTs located along the invert of

851

Table 2.

Instrumentation and nomenclature.

Instrument

Measurement and symbol

Triaxial pipe load cells

Vertical pipe-soil load (/unit length) Lateral pipe-soil load (/unit length) Axial pipe-soil load Vertical pipe movement 1 Lateral pipe movement Axial pipe movement Excess pore pressure on pipe surface (at 5 locations on invert, and 4 locations at 30◦ from invert)

Displacement transducers Pore pressure transducer

V H F w uh ua u

1

Figure 1. Schematic SMARTPIPE® frame. Table 1.

view

from

This value is corrected for any frame settlement based on a displacement transducer linked to a settlement plate resting on the seabed.

beneath

Key dimensions and properties of SMARTPIPE®.

Dimension

Value

Model pipe diameter Overall length of model pipe Length of central measurement section Surface coating

225 mm 1100 mm 776 mm Polypropylene

Five types of test were carried out in the SMARTPIPE® campaign. These were (i) cyclic T-bar penetration tests, (ii) vertical pipe penetration tests, (iii) pore pressure dissipation tests, (iv) axial pipe tests, and (v) lateral pipe tests. This paper focuses on the results from one particular axial pipe test, which serves to illustrate several important phenomena that control the resistance that can be mobilised during axial pipe movement. 3

Figure 2. Cutaway view of instrumented pipe showing load cells and pore pressure transducers.

Figure 3. Pore pressure transducers on pipe surface.

the model pipe, and at each end of the measurement section there are PPTs located 30 degrees around the circumference on each side (Figure 3). The instrumentation and the resulting measurements are summarised in Table 2. A hydraulic system provides the actuation of the model pipe. The vertical axis can operate in displacement-rate control or in load-controlled modes, although the load-controlled mode currently involves some manual intervention. © 2011 by Taylor & Francis Group, LLC

SITE CHARACTERISATION

The seabed conditions at the site comprise very soft high plasticity marine clay, with a shear strength profile that increased monotonically with depth. The liquid and plastic limits were 183% and 77% respectively, giving a plastic index of 106%. The SMARTPIPE® features a miniature T-bar penetrometer, which is 12 mm in diameter (although the bar can be interchanged with other sizes, if required). This device is smaller than the conventional T-bar penetrometer, which has a diameter of 40 mm, to provide better strength resolution close to mudline. At this site, the T-bar results showed that a layer of slurry, 50– 100 mm in thickness was present, but had negligible strength. Below the slurry, the intact soil strength rose from approximately zero at the soil-slurry interface with a gradient of ∼ 10 kPa/m over the first 0.3–0.4 m depth (which is the range relevant to the pipe tests) before rising at a lower, more usual rate, at greater depth. The remoulded strength, derived from cyclic T-bar tests, rose at 3 kPa/m from zero at the soil-slurry interface. 4

EXCESS PORE PRESSURE DISSIPATION

The first stage of each SMARTPIPE® test involved vertical penetration of the pipe into the seabed, followed by a period of excess pore pressure dissipation. The vertical penetration response is not presented here. However, the resistance recorded during the faster penetrations was consistent with the calculated resistance based on the strength profiles

852

Table 3.

Cyclic axial test parameters.

Sweep

Length (mm)

Speed (mm/s)

1 2 3 4 5 6

+150 −200 +200 −200 +250 −250

0.04 mm/s

0.15 mm/s

Figure 4. Excess pore pressure dissipation after vertical penetration: comparison of SMARTPIPE® data with FE results.

derived from the T-bar tests, and the solutions presented by Randolph & White (2008), based on the analysis of Merifield et al. (2009). Slower rates of embedment led to reduced resistance, which can be attributed either to concurrent consolidation settlements or the effect of strain rate on soil strength. After penetration to a specified depth, the vertical load was held constant (or near-constant), typically for 4 hours, and the dissipation of excess pore pressure at the various PPT locations was monitored. The results from one particular test, which involved the highest and most steady vertical load level and therefore the largest and most consistent pore pressure response, have been compared with the numerical solutions presented by Krost et al. (2010) and Gourvenec & White (2010). These solutions are based on finite element (FE) analyses adopting an elastic soil model with coupled consolidation. Analyses with and without soil heave around the pipe were performed, showing minimal effect (Gourvenec & White 2010). The average excess pore pressure measured at the pipe invert shows excellent agreement with these solutions, where the numerical results are scaled according to a coefficient of consolidation of cv = 75 m2 /year (Figure 4). The good agreement throughout the dissipation period gives confidence in this back-analysis, which indicates a value of cv that is 200 times higher than values inferred from reconstituted samples of the same soil tested in the laboratory. This observation has implications for assessments of the drainage condition during pipe movements and other consolidation events at this site. One key benefit of the SMARTPIPE® tool is that the consolidation characteristics of the surficial soil can be measured in situ – in this case leading to a significant reassessment of this design soil property.

5

CYCLIC AXIAL PIPE TEST

5.1 Overall forces and displacements The cyclic axial pipe test report here was conducted at an embedment of 0.65D, over a period of 7 hours. The vertical load on the pipe was maintained in the range 1–1.2 kN/m throughout most of the test, which consisted of six axial sweeps with the lengths and rates of movement given in Table 3. © 2011 by Taylor & Francis Group, LLC

Figure 5. Time history during cyclic axial pipe test.

The time histories of imposed vertical load, imposed axial displacement and measured axial resistance are shown in Figure 5. The vertical load gradually decayed over the test, and a small cyclic component of vertical load (∼5% of the steady value) was also present, due to a slight misalignment in the drive spindle – which is evident in the load-displacement response (Figure 6). The axial response included a modest peak in resistance during the first sweep, and a ductile response during all subsequent sweeps (Figure 6). 5.2

Interpretation of effective stresses

The nine PPTs allow an effective stress interpretation of the pipe-soil response to be performed, although some assumptions must be made, in order to link the applied loads (which act around the entire pipe-soil contact surface) to the measured pore pressures (which are predominantly at the invert). We have adopted averaged quantities, using the inadvertent ‘wobble’ in the SMARTPIPE® spindle to calibrate between the invert and average pore pressures. The total load on the pipe-soil surface, N, is the applied vertical load, V, multiplied by a ‘wedging factor, taken here as ζ = 1.27 (White & Randolph 2007). The wedging effect arises because the sum of the normal forces on the curved pipe-soil interface exceeds the vertical force. The mean total normal stress is then found as σn = N/LP where P is the contact perimeter. A value of πD/2 is adopted for P: although the embedment is slightly greater than 0.5D, the soil was observed, via a camera mounted on the seabed

853

Figure 6. Load-displacement response in cyclic axial pipe test.

Figure 8. Effective stress interpretation: response during change of direction.

5.3 Total and effective stress failure criteria Figure 7. Mean measured pore pressure, total stress and effective stress response during cyclic axial pipe test.

frame, to stand clear of the pipe above the centreline. The mean total normal stress varies between approximately 4.5 and 3.5 kPa, showing a general decrease throughout the test. The mean measured excess pore pressure, uav,meas , based on 7 PPTs, of which 4 were located at the pipe invert, also decreased over the test period (2 PPTs were inoperative). Superimposed on this slow trend are sharp variations associated with changes of pipe direction (Figure 7). Since the mean measured pore pressure is based predominantly on data from the pipe invert, this value is likely to be an over-estimate of the mean pore pressure around the full pipe-soil contact width – which extends to the soil surface, where the excess pore pressure must be zero. The small cyclic load component proved to be useful in allowing an adjustment factor to be derived to account for this effect. It was found that if the mean measured pore pressure, uav,meas , is scaled by 80%, then the cyclic component is minimized within the inferred mean effective stress, σn
= σn – 0.8 uav,meas – as shown in the short period of data from 1–1.4 hours, close to a change in direction (Figure 8). This adjustment is based on an assumption that the small cyclic load is undrained and causes no changes in effective stress. A reversal of the axial movement results in a sharp rise in the mean pore pressure, and a corresponding reduction in the effective pipe-soil stress (Figure 8). In this case σn
falls from ∼3 kPa to ∼1 kPa, for about 0.15 hours (10 minutes) before steady conditions are re-established. This effect is linked to the mobilisation of axial resistance in Section 5.4. © 2011 by Taylor & Francis Group, LLC

Having inferred the variation mean total and effective stresses at the pipe soil surface, the data from all six sweeps can be compared via both total stress and effective stress failure criteria. The mean shear stress on the pipe surface is calculated as τav = H/(πD/2). In pipeline design, it is common to consider a friction factor defined based on the total forces: FF = H/V. The equivalent effective stress quantity is
FF
= τav /σn,av . The FF response is smoother than the FF
response. This is partly from the influence of the PPT data, which showed some small variations between sensors, indicating that the excess pore pressure was non-uniformly distributed (Figure 9). However, when these responses are plotted in the form of total and effective stress failure criteria, the effective stress interpretation is the most consistent. A total stress failure criterion for shearing on a prescribed interface takes the form of a limiting shear stress, independent of the imposed total normal stress. The data from the six sweeps shows that the steady limiting shear stress spanned the range 1–1.5 kPa, with cycle 2 showing a higher strength than cycles 1 and 3 (Figure 10). In an effective stress interpretation – σn
vs. τav – all three sweeps in each direction overlie each other (Figure 11). The first sweep shows a peak in resistance, to a stress ratio of τav /σn
= 0.81, suggesting an initial peak in the mobilized friction angle. The remaining data lie close to the effective stress failure criterion of White & Randolph (2007):

854

with fitting parameters of A = −0.4 and B = 0.6. This failure envelope and the axial test data are compared to results from a set of model pipe tests conducted at the Norwegian Geotechnical Institute

Figure 12. Comparison of effective stress failure response with data from other sources.

5.4

Figure 9. Friction factor response during cyclic axial pipe test.

Figure 10. Total stress interpretation of axial response.

Figure 11. Effective stress interpretation of axial response.

using the same clay (Figure 12). These tests were back-analysed in the same manner, using measured pore pressures and a wedging factor to derive effective stress quantities. Good agreement is evident, with the SMARTPIPE® data falling amongst the NGI data, towards the lower bound of the envelope. When the NGI results are interpreted in a total stress manner, the friction factor varies by a factor of 4 (Bruton et al. 2009). The effective stress approach gives closer bounds, and shows that a lower total friction factor is associated with high excess pore pressures around the pipe surface. Also shown on Figure 12 are values of friction angle derived from tilt table tests using the same clay. These also agree closely with the model test data. © 2011 by Taylor & Francis Group, LLC

Mobilisation: A function of distance or time?

The stress and pore pressure responses during the initial 60 mm of each axial sweep are shown in Figure 13. This figure also illustrates the extent of the ‘wobble’ of the spindle mentioned earlier, showing that it is evident as cyclic “noise” and does not mask the overall trend of the data. At every reversal there is an increase in pore pressure during the initial movement, resulting in a corresponding reduction in the effective stress. The pore pressure then recovers close to the initial value, as does the effective stress. The exception to this behaviour is the first sweep, in which the total stress reduces over the first 30 mm of movement, causing a corresponding reduction in the effective stress. In these tests the generation and recovery of excess pore pressure is strongly influenced by the rate of pipe movement. During the first two cycles, the pore pressure transient takes place over ∼20 mm of movement, whereas during the final cycle (which took place at a higher pipe velocity) the transient is spread over ∼60 mm. The same behaviour is evident in the shear response. Figure 14a shows the mobilisation of axial shear stress during the initial 60 mm of each sweep with the negative values of resistance retained for clarity. The general shape of each response is consistent, although in some cases there is an initial residual shear stress from the previous sweep. The exceptional case is the first sweep, which shows a brittle peak. The faster sweeps show a more compliant response, and a higher mobilisation distance. The shape of the mobilisation response in the different sweeps is better compared by considering the normalised shear resistance, S, defined as:

where τinit is the initial residual shear stress and τ60 is the fully-mobilised resistance at 60 mm displacement. When plotted in this manner, the responses at each speed overlie each other closely, with the exception of the brittle first sweep response. The effect of velocity is captured by relating the normalised shear stress to the time since the reversal point, rather than the distance. When plotted in this

855

transient observed over the same period of time after each reversal. This mechanism adds complexity to the selection of a mobilisation distance for design, if pipe movements occur at an undrained or partially-drained rate, creating excess pore pressure. 6

CONCLUSIONS

Results from the first field deployment of the SMARTPIPE® have been presented. The resulting data provides new insights into the mechanisms of pipe-soil interaction, with the benefit of having been gathered in situ, in deep water, on intact soil. A cyclic axial test has been back-analysed using an effective stress frictional failure criterion, which provides a more consistent interpretation than a total stress approach. However, the generation of excess pore pressure means that the pipeline friction factor for a given soil varies if movements occur at an undrained or partially-drained rate (or when lay-induced pore pressures remain). The importance of pore pressures is highlighted by evidence that, in this case, the full pipe-soil resistance is only mobilised when steady pore pressures are reached, which is after a fixed time, rather than a fixed displacement.

Figure 13. Detailed stress-pore pressure response during the initial 60 mm of each sweep.

ACKNOWLEDGEMENTS The authors wish to thank BP Exploration for permission to publish these data. The SMARTPIPE® was developed by Fugro with support from BP. REFERENCES

Figure 14. Shear stress mobilization during the initial 60 mm of each sweep.

Figure 15. Shear stress mobilization with elapsed time.

way, all of the mobilisation curves overlie each other, except for the first breakout (Figure 15). This remarkable agreement suggests that in this case the mobilisation of axial resistance is a time-related process, rather than being linked to the distance of shearing. This is consistent with the pore pressure © 2011 by Taylor & Francis Group, LLC

Bruton D., White D.J., Langford, T.L. & Hill A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments Proc. Offshore Technology Conference, Houston, USA. Paper OTC20096, 12pp. Gourvenec S.M. & White D.J. 2010. Elastic solutions for consolidation around seabed pipelines. Proc. Offshore Technology Conference, Houston. Paper 20554, 16pp. Hill, A.J. & Jacob, H. 2008. In-Situ Measurement of PipeSoil Interaction in Deep Water. Proc. Offshore Technology Conference, Houston, USA. Paper OTC 19528, 18pp. Krost K., Gourvenec S.M. & White D.J. 2009. Consolidation around partially-embedded submarine pipelines. Géotechnique, Accepted 19 January 2010, in press. Merifield, R., White, D.J. & Randolph, M.F. 2009. Effect of surface heave on response of partially embedded pipelines on clay. ASCE J Geotech..& Geoenv. Eng. 135(6):819– 829. Randolph, M.F. & White, D.J., 2008. Pipeline Embedment in Deep Water: Processes and Quantitative Assessment, Proc. Offshore Technology Conference, OTC 19128, 16 pp. White D.J. & Randolph M.F. 2007. Seabed characterisation and models for pipeline-soil interaction, Int. Journal of Offshore & Polar Engineering. 17(3), pp.193–204.

856

11 Trenching, ploughing, excavation and burial

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Influence of object geometry on penetration into the seabed A. Ivanovi´c, R.D. Neilson & G. Giuliani University of Aberdeen, Aberdeen, UK

M.F. Bransby University of Dundee, Dundee, UK

ABSTRACT: The seabed is disturbed by predominently horizontal cutting mechanisms during the installation of pipelines (e.g. by ploughs), by fishing gear components and by natural processes (e.g. iceberg scour). In each case the zone of soil disturbed by the device and/or the soil resistance force to the movement of the object is of interest to the geotechnical engineer. Consequently, this paper reports an experimental investigation of how different shaped objects penetrate the seabed. The results show that the penetration is influenced by both the front face angle and the weight of the object while the drag force is mostly influenced by the weight.

1

INTRODUCTION

Cutting processes that commonly disturb the seabed and have direct implications for the offshore industry are iceberg scour, anchor dragging, trawling and the cutting and ploughing processes associated with the installation of pipelines and cables. The first three processes are out of the control of an engineer and if contact of any of these is made with a pipeline this can cause damage. Pipeline burial to beneath the affected zone is the primary form of mitigation. Ploughing is a more controlled process, which is used during pipeline installation. Both iceberg scour and ploughing processes disturb the mechanical state of the soil as a rigid body penetrates horizontally through the seabed near the surface. These processes are examined in this paper by means of physical modelling. 1.1 Ice gouging The challenge that offshore pipeline engineers have been faced with in recent years is the estimation of the most economical burial depth for a pipeline, at which it remains safe and in a good condition. During the movement of an iceberg it has been reported by various studies (e.g. Palmer et al., 1990; Woodworth-Lynas et al., 1996) that the soil below a scouring ice keel will displace laterally, in the direction of ice movement, transverse to the ice movement and vertically. The scouring process will produce soil deformations, which can produce strains and stresses that can be transferred to any nearby buried pipeline, which in turn could be damaged. The seabed affected by an ice scour process is described generally through three zones (Palmer et al., 1990). Zone 1 is the top layer soil that is gouged by ice – there is direct contact with the ice and consequently, there is a high probability © 2011 by Taylor & Francis Group, LLC

859

Figure 1. The geometry of the penetrating object.

that any pipeline within this zone would be subjected to damage. Zone 2 is situated below the level of the base of the ice keel and is subjected to large plastic soil deformations. The magnitude of these soil deformations and therefore the depth of zone 2 are difficult to estimate because they are dependent on many factors such as the type of soil and the shape of the ice feature causing the ice scour. Finally, Zone 3 is where the soil deformation is limited to elastic (or at least small displacement) behaviour. Consequently, the safe burial depth is usually estimated to be within Zone 3. Since this may be very deep for large icebergs it may not be necessarily the most economical solution. The optimal burial depth therefore tends to be located in Zone 2 where the pipeline can perform as designed but without being damaged (Fig. 1). The pipeline does not have the same capability as the seabed to resist the ice contact forces. The challenge for the pipeline designer is to determine the safe burial depth of pipeline in order to avoid potential damage to the pipeline if the ice scour path coincides with the position of the pipeline. It is very difficult to define exactly ice scour scenarios because they depend on environmental forces, such as wind (in shallow water) and the current, as well as the seabed conditions. This

Figure 2. The geometry of the penetrating object.

explains why ice gouging poses a threat to submarine pipelines in arctic areas. The influence of iceberg scour on pipelines has been considered by Barrette & Timco (2008) where small scale laboratory experiments were undertaken in a 6 m long sand channel using real ice models. The research confirmed the presence of the three zones identified previously. 1.2

Figure 3. Sand channel used to contain the soil sample at the University of Aberdeen.

Ploughing

Pipelines are often protected by burial. One common method of burial is by ploughing, where a large (e.g. 20 m long) plough is towed along the seabed to form a trench (e.g. Palmer et al., 1979). The pipeline is placed in the open trench and the trench is then backfilled. One important aspect of the operation is the tow force from the support vessel required to progress the plough and so semi-empirical relationships have been developed to predict how this force varies with the weight of the plough, the depth of trench formed and the velocity (e.g. Reece & Grinsted, 1986; Cathie & Wintgens, 2001). Like an iceberg, the plough share (which performs the main ‘cut’) penetrates the seabed and moves horizontally during steady-state ploughing. However, unlike an iceberg, the face of the share is inclined backwards (i.e. with β < 90◦ in Fig. 2). 1.3 Aim of study This study of ice scouring and ploughing will focus on how the mechanical properties of the soil and the shape of the penetrating object affect the trenching depth, drag force and the subsurface deformations.The geometry considered is shown in Figure 2. The object moves horizontally through the seabed and cuts a trench of depth, D, measured from the original surface, requiring a force, F to move. The key characteristic of the geometry is the angle between the base and the front face of the object, β (Fig. 2). In terms of the soil resistance, F, to movement, the most applicable theory is that of retaining walls where the passive earth pressure may be considered to act in front of the object (and depend on D and the width of the object, L) with a base shear resistance (depending on the base roughness and self weight, W) also acting against the direction of movement. © 2011 by Taylor & Francis Group, LLC

2

EXPERIMENTAL METHOD

2.1 Test samples Laboratory experiments were conducted in which the penetration behaviour of objects of different shapes were examined. The objects were tested in a 4.8 m long and 50 cm wide channel in which the penetration of the object into the sediment, and the towing force were measured (Fig. 3). This channel has rails that support a trolley, to which the component to be tested is attached. The trolley allows the component to move freely in the vertical direction under a fixed vertically applied load, W, and includes a load cell which measures the drag force, F, and a LinearVariable DifferentialTransformer (LVDT) that measures the penetration of the component into the sediment, D, as the object was towed along the channel at constant speed by a winch mechanism. Potential rotational motions were restrained but reaction moments were not measured. Uniform grading, dry silica sand (rounded particles; D50 = 144 µm; φcrit = 32◦ etc.) was used in all tests to form the seabed sample. This was prepared by raking and leveling the sand between the tests to produce a sample with a unit weight of 15.6 kN/m3 (Dr = 25.6%). All samples were prepared dry to ensure fully drained conditions were generated during the cutting process for the preliminary experiments reported here. The test pieces were constructed using steel boxes with base dimensions, B = 150 mm and L = 150 mm where the angle of the front face, β, was varied from 60◦ to 150◦ as shown in Figure 4. An additional box with the front face at 90◦ and with a curved edge of

860

Figure 5. Drag force vs. penetration into the seabed for the 90◦ box towed at 0.1 m/s and 0.5 m/s (W = 50 N).

Figure 4. The steel boxes used as the penetrating objects.

15 mm radius was constructed to investigate how the curved edge affected its performance during seabed penetration. Each component was towed at a constant rate along the sediment placed in the tank and the drag force and the penetration measured. The tests show how the drag forces and soil displacements are affected by the shape and penetration depth. The test pieces were dragged at two different speeds, 0.1 and 0.5 m/s, and different masses were added to the original weight of the boxes themselves to give total weights of 50, 100 or 150 N during the tests.

3

Figure 6. Drag force vs. penetration t into the seabed for the 90◦ box towed at 0.1 m/s and 0.5 m/s (W = 100 N).

RESULTS

3.1 Effect of velocity on drag force The influence of the speed at which the boxes were towed (0.1 and 0.5 m/s) was examined first. Figures 5–7 show the force-penetration relationship for different weights for the box with the front face at 90◦ . In general it can be seen that the drag force increases with an increase in penetration which would be expected from retaining wall theory. For the three different weights taken together, there is no consistent difference in the F-D relationships for different velocities. For dry sand the drag force would not be expected to change with velocity for reasons of pore fluid movement, although differences might be expected at much higher rates because of dynamic inertial effects (i.e. local accelerations of soil around the fast-moving object as new soil is pushed away by the object). The variable forces shown for the lighter boxes towed at higher speeds are due to build up of a heap of sand in front of the test piece. The heap then spills to the side, reducing in height, and then the process repeats. This oscillation becomes less prominent as steady state conditions are achieved. © 2011 by Taylor & Francis Group, LLC

Figure 7. Drag force vs. penetration for box with β = 90◦ , towed at 0.1 and 0.5 m/s (W = 150 N).

3.2

Effect of geometry and weight on penetration and drag force

A series of tests were conducted in which each box shape (defined by angle β) was tested separately with total weights of 50, 100 or 150 N. Near steady-state conditions were generally achieved after towing the object between 2 and 3 m along the channel (see Figure 8). The results presented in Figures 9 and 10 show the penetrations and the drag forces for different weight and speed scenarios at steady-state conditions.

861

Figure 8. Vertical penetration against horizontal displacement along the seabed for the boxes with different front face angles (v = 0.1 m/s, W ≈ 150 N).

Figure 10. Steady-state drag force against front face angle for different object weights.

Figure 9. Steady-state penetration against front face angle for different object weights.

Figure 8 shows that for an approximately fixed W, the highest penetration is obtained when β = 60◦ which represents a shape applicable for ploughing applications. In contrast, the smallest penetration was noted for β = 150◦ , applicable to the ice scouring problem. The latter shape reaches the steady state penetration after a horizontal displacement of 0.4 m while for β = 60◦ , steady-state behaviour is obtained after a displacement of between 1.5 and 2 m. There is a clear trend of reduction in penetration with an increase in front face angle, β as shown in Figures 9a and 9b. Drag force however does not change as significantly with front face angle (Fig. 10), although there is a small reduction in drag force with increasing β. Given that penetration is reducing significantly, this suggests that the drag force for similar penetrations will be larger as β increases, and this will be investigated later. © 2011 by Taylor & Francis Group, LLC

Figure 11. Drag force vs. penetration relatiosnhip for 105◦ box towed along the seabed at 0.1 m/s with different weights.

There is a general trend of increasing penetration depth (Fig. 9) and drag force (Fig. 10) with increasing weight, W, for all β which would be expected. It is however noted that the results for β = 105◦ do not quite follow the general trend of the penetration data. Figures 11 and 12 show how the drag forces vary with object penetration during tests with β = 105◦ and 150◦ respectively. Each graph shows results for the three weight conditions. There is clearly a substantial drag component at zero penetration, which would be expected to be due to base shear. The interface friction coefficient, H/V = µ = tan δ was estimated by plotting the drag versus normal force at D = 0 and taking the gradient of the resulting plot. The drag force values for D = 0 used in this process were found by curve fitting the graph of drag force

862

Table 1. Passive coefficient, Kp values for different angles, β.

Figure 12. Drag force vs. penetration relatiosnhip for 150◦ box towed along the seabed at 0.1 m/s with different weights.

Figure 13. Drag force vs. weight relationship for estimation of interface friction coefficient.

against penetration and identifying the intercept at D = 0. This analysis revealed that the resistance at zero penetration is described by F ≈ 0.31 W
as shown in Figure 13. This implies a friction angle, δ = 17.2◦ for the smooth, rolled steel-sand interface. For larger penetrations (Fig. 11 and 12), the drag force, F, increases with depth of penetration. This drag term is likely to be due to the frontal resistance and side shear resistance, which are depth dependent. Similar results were reported by Phillips et al. (2005) for iceberg scouring and by Reece & Grinsted (1986) for ploughing. The drag force – penetration relationship might be expected to vary as:

where Kp = the passive earth pressure coefficient, D = penetration depth below the original ground surface and L = frontal width. Curve fits using Eq. 1 with δ = 17.2◦ , γ = 16 kN/m3 and Kp = 20 for β = 105◦ and Kp = 40 for β = 150◦ are shown in Figures 11 and 12 alongside the original test results. In order to achieve the best fit the value of Kp was adjusted for each front face angle β while keeping the measured interface friction angle and the soil unit weight constant. The selected values of Kp are shown in Table 1. The data show that Kp increases with front © 2011 by Taylor & Francis Group, LLC

Front face angle, β (degrees)

Kp

60 90 105 120 150

10 12 20 30 40

face angle β reflecting the increasing lateral and side pressures for blunter front face angles (larger β). The difference in Kp values to fit the data as front face angle changes, reflect the quite different flow patterns around the 60◦ and 150◦ shapes, which give the large difference. When β = 150◦ the object will push forward and compress vertically the sand in front causing a partial bearing capacity failure (e.g. see Palmer et al. 1990 for clay conditions). This will cause a large vertical reaction force which will prevent significant penetration into the seabed without large W and increase local effective stresses and normal contact forces and thereby reaction forces. In contrast the object with β = 60◦ has a face inclined with an upward normal component. The sand will flow over the front face (causing a small additional vertical downward load onto the base) and then slide to the sides with little vertical restraint. This may account for the reduced drag, despite the greater penetration and the variation of Kp with β. Finally, it should be noted that the Kp values backcalculated (Table 1) did not take into account the build up of sand in front of the object. Consideration of the spoil heaps would reduce the effective Kp values used in calculations, but require additional knowledge of the changing spoil heap size to calculate penetration resistance as penetration and translation varied. This will be studied in future experiments. An additional experiment was undertaken to examine soil deformation mechanisms by pulling a test piece with β = 150◦ adjacent to a Perspex box side with the same loading conditions as used in the main test series. Multiple digital images (e.g. Fig. 14) were captured of the process and analysed using geoPIV (White et al., 2003) to measure soil displacements during penetration. Figure 14 clearly shows the build up of sand in front of the object during steady-state (i.e. horizontal) penetration. Figure 15 shows the calculated instantaneous soil movements which occurred during penetration of the 150◦ test piece. The soil beneath shows downward components of movement, which generate the upward reaction force, which reduces penetration and increases drag force. 3.3

Influence of the curvature of front edge of the object

The data for 90c compared to 90 is interesting as there is a large difference in penetration (e.g. 55%

863

Figure 14. Digital image of object penetration at steady-state in apparatus at the University of Dundee (β = 150◦ ; W
/BL = 3.6 kPa).

The results show that the front face angle of the penetrating object and the vertical load applied to it significantly affect both the steady-state drag force and, to an even greater extent, the penetration depth through the soil. The more acute the angle between the base and the front face, the greater is the penetration. The practical implication of this finding is that if the geometry of a typical iceberg is known then it may be possible to estimate the required burial depth of pipelines. Furthermore, the results of this study indicate that by introducing a curved front edge, the penetration reduces but the drag force does not drop proportionally. This suggests that an object with a curved edge is likely to cause less damage to a buried pipeline than one with a sharp edge. ACKNOWLEDGEMENTS The authors would like to acknowledge the assistance of Steven Boertien (University of Dundee) who conducted the test shown in Figure 14. REFERENCES

Figure 15. Incremental soil displacements for object with β = 150◦ before steady-state penetration (D/B = 0.28; Scale factor = 3).

reduction for the condition shown in Fig. 8), which is not reflected in the drag force for higher speeds. The difference in penetration depth is due to the curved frontal edge which allows the sand to flow under the object rather than the edge acting as a cutting tool, reducing penetration depth for the object with the curved edge. A similar effect has been found while undertaking experiments on trawl otter doors. Whether such differences occur for the cases for β > 90◦ (for example for icebergs with abraded, rounded edges) requires further study. 4

CONCLUSIONS

The work presented in this paper focuses on the influence of iceberg and ploughing objects on the seabed. An experimental rig has been set up and boxes of different geometries tested.

© 2011 by Taylor & Francis Group, LLC

Barrette, P. & Timco G. (2008) Ice scouring in a large flume: Test set-up and preliminary observations ICETECH08133-RF. Cathie, D.N. & Wintgens, J-F. (2001). Pipeline trenching using plows: performance and geotechnical hazards. Proc. Offshore Technology Conference OTC 13145, Houston, May 2001. Palmer, A. C., Kenny, J. P., Perera, M. R. & Reece, A. R. (1979). “Design and operation of an underwater pipeline trenching plough” Géotechnique 29 (3): 305–322. Palmer, A.C., Konuk, I., Comforg, G. & Been, K. (1990). Ice gouging and the safety of marine pipelines. Proc. Offshore Technology Conference OTC 6371, Houston, May 1990. Phillips, R., Clark, J.I. & Kenny, S. (2005) PRISE studies on gouge forces and subgouge deformations. Procedeeings of the 18th International Conference on Port and Ocean Engineering under Artic Conditions (POAC), Potsdam, USA, Vol. 1 pp 75–84. Reece, A. R. & Grinsted, T. W. (1986). Soil Mechanics of Submarine Ploughs. Proc. Eighteenth Annual Offshore Technology Conf., Houston (5341) May 1986, 453–461. White, D. J.,Take, W.A. & Bolton, M.D. (2003). Soil deformation measurement using particle image velocimetry (PIV) and photogrammetry. Géotechnique 53(7): 619–631. Woodworth-Lynas, C.M.L., Nixon, J.D., Phillips, R. & Palmer, A.C. (1996) Subgouge deformations and the security of Arctic marine pipelines. Paper OTC8222 in Proceedings, Twenty-eighth Annual Offshore Technology Conference, Houston, 4 657.

864

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Investigation into the effect of forecutters on plough performance K.D. Lauder, M.J. Brown & M.F. Bransby University of Dundee, Dundee, UK

J. Pyrah CTC Marine Projects Ltd, Darlington, UK

ABSTRACT: Pipeline ploughing is a common method used to bury offshore pipelines for protection. During ploughing, rate effects which occur in fine grained granular soils increase the tow force required by the support vessel and may reduce achievable plough velocities, thereby increasing the duration of a contract. The share of the plough ‘cuts’ the trench with some ploughs featuring a second, smaller cutting tool known as a forecutter which sits in front of the share. The effectiveness of the forecutter in reducing tow forces during ploughing has been investigated by reduced-scale model testing and is outlined herein. The forecutter is shown to be beneficial in reducing the rate effect but has a negative impact on the ‘static’ component of tow force. Conditions where the forecutter should be of overall benefit to reducing the tow force during ploughing are described.

1

INTRODUCTION

1.1 Pipeline burial Offshore pipelines are commonly buried around 1.5 metres below the sea bed surface. Burial is used as a means of protection from fishing activities and hydrodynamic loading and if backfilled can prevent movements due to thermal expansion on commissioning of the pipeline. A pipeline plough, towed along the sea bed by a support vessel is one of the most common means by which pipelines are buried. Ploughs use a large share (see Fig. 1) to ‘cut’ a trench and often the pipeline runs through the plough and is laid directly into the trench as it is formed. It is important that a minimum cover to the pipeline is maintained, along with a relatively constant trench profile to maintain support, which can be particularly tricky when ploughing through geohazards such as sandwaves, (see Bransby et al. 2010a, b). For commercial reasons all these aims need to be achieved in the shortest possible time period.

Figure 1. Components of a typical pipeline plough.

The rate effect is therefore very important as it may control the velocity at which the plough can be towed. In addition to the share, a second smaller cutting blade, known as a forecutter (see Fig. 1) may be attached in front of the share and used as a means to reduce the rate effect. Although literature does exist on the merits of multiblade ploughing (e.g Hata, 1979), there is limited accessible general information available to make an informed decision as to whether a forecutter actually reduces rate effects during ploughing. The work outlined in this paper is a step towards a better appreciation of the forecutter’s effect on plough performance.

1.2 Rate effects Although ploughs are used widely by industry to bury pipelines, the soil mechanics occurring during ploughing are not wholly understood. When ploughing in sands and silts, the rate of ploughing is such that it causes partial drainage of the soil as it shears and in turn causes a rate effect whereby the tow force increases with plough velocity. The support vessels which provide the forward thrust to the plough have a limited capacity and so the velocity achieved may be limited by the tow force. © 2011 by Taylor & Francis Group, LLC

2 2.1

EXPERIMENTAL METHODS Introduction

Scale model testing was used to investigate the significance of a forecutter on plough performance. This approach allowed for sand conditions and the surface of each test bed to be prepared in a highly repeatable manner. A 50th scale pipeline plough was manufactured with a detachable forecutter. This allowed for

865

Table 1. depth.

Figure 2. Schematic of experimental apparatus.

Numbered skid settings and the resulting plough

Skid setting

Arm angle (deg)

Mean plough depth (mm)

1 2 3 4 5

67 50 44 30 17

19 26 30 36 41

a normal effective stress range of 3–26 kPa. Standard laboratory tests gave Gs = 2.63, ρmax = 19.48 kN/m3 and ρmin = 14.33 kN/m3 . 2.4

Sample preparation

The dry sand test beds were prepared by a two stage process of stirring the sand in the test tank to create a uniform density followed by removal of the uneven surface by scraping with a flat edge. The saturated test beds were prepared by first saturating the sand and then stirring and scraping the surface flat.The resulting densities were a loose state (Dr = 32%) in dry sand and medium dense state (Dr = 53%) in saturated sand. Figure 3. Comparison of the generation of steady state values.

3 a comparison of plough performance during ploughing tests both with and without a forecutter. Because all lengths scale by a factor of 50, the normal effective stresses during model testing are 50 times smaller than at prototype scale, with associated implications on dilation as described by Bolton (1986). 2.2 Apparatus and procedure The apparatus used to perform the ploughing tests is shown in Figure 2. In preparation for tests where the plough was to be used without a forecutter it was ballasted to keep the plough mass and its centre of gravity the same as when the forecutter was used. Before each test the plough was placed onto the sand at the end of the tank (left hand side in Fig. 2). The trolley was then moved into position above the plough and a load cell of 200 N capacity and a 250 mm stroke LVDT were attached to the plough. The tank is 0.5 m wide by 0.5 m deep and is 2 m long, which allowed the 250 mm long plough enough travel to find its steady state depth in the sand according to the skid settings and provide data at the desired trenching depth. For a description of how a plough maintains its depth by moment equilibrium, see Palmer (1979) and Lauder et al., (2008). 2.3 Material properties All of the tests described in this paper were conducted in poorly graded fine Congleton sand with D60 = 0.15 mm and D10 = 0.10 mm. Shear box tests revealed ϕcrit = 31◦ and for Dr = 53%, ϕmax = 37◦ over © 2011 by Taylor & Francis Group, LLC

866

3.1

RESULTS Establishing steady-state values

Figure 3 shows the plough sits on the surface of the sand (depth ≈ 0 mm) at the start of the test and by around 600 mm displacement it has achieved a constant depth known here as steady state. Steady state values are the value of tow force, plough depth and pitch when the plough is in dynamic equilibrium, i.e. it has reached its optimum depth according to its skid settings and from this point onwards the tow force, depth and pitch (defined by θ in Fig. 1) all show constant values with increasing plough displacement. Figure 3 compares results from two tests in loose dry sand: one with a forecutter and one without. Although the skid settings are the same for both tests the plough trenched slightly deeper without a forecutter than with, which was due the forecutter causing the plough to pitch forward slightly more than without. Comparing the tow force generated during the two tests it is clear that the forecutter causes a slight increase of tow force in dry sand even though it caused the plough to trench slightly shallower. 3.2 The forecutter’s influence on ploughing depth 50th scale ploughing tests were conducted in dry sand at a range of model depths from 19 mm–41 mm, which when scaled up by 50 times correspond to prototype scale depths of 0.95 m–2.05 m. Table 1 shows the skid settings used during the following tests to dictate the plough’s depth. The depth

Figure 4. Depiction of arm angle, defined as the angle of the arm relative to the heel of the share.

Figure 6. Relationship between plough depth and tow force in dry loose sand.

3.3 The forecutter’s influence on tow force in dry sand Figure 5. The change in plough pitch and depth due to the forecutter.

Cathie & Wintgens (2001) advanced earlier work by Reece and Grinsted (1986) and found that the ‘static’ tow force generated could be broken into two components, (see Eq. 1).

associated with each skid setting is the average plough depth expected for each skid setting and varies slightly from test to test. Arm angle is defined in Figure 4 as the angle between the heel of the share and the arm which supports the skids. Figure 4 shows two different arm angles of 17◦ and 44◦ respectively and highlights how arm angle influences the vertical distance between the skids and share. Figure 5 shows the effect of the forecutter on pitch and depth where aft pitching is described as being positive and forward pitching being negative. Each data point represents the average steady state value (as described in the previous section) of depth and pitch for a particular test. The data points that have been grouped together (circumscribed) are from tests carried out with the same skid settings, the exact setting (see Table 1 for reference) is indicated by the number which the oval also surrounds. The data in Figure 5 shows that the plough’s pitch will change significantly with depth and ranges from 0.15◦ at 19 mm depth to −1.2◦ at 41 mm depth. The difference between the tests with a forecutter and those without is only marginal but the forecutter does tend to make the plough pitch forward slightly more than it would without. This results in a small reduction in the plough depth for any particular skid setting. Although the difference between the depth of the plough when a forecutter is used and when not is only small, (around 2%) it should still be considered when examining the forecutter’s effect on the tow force generated during ploughing.

where Cw is a constant depending on interface friction between the plough and sand, W is the weight of the plough, z is the cutting depth, Cs is a constant dependent on relative density and γ is the unit weight of the sand. The first term in Eq. 1 is a base friction term which is the product of the plough’s self weight and the coefficient of friction between the plough and the surrounding sand. The weight of the 50th scale model plough is 16 N which is equivalent to 1/503 the weight of prototype weight. The interface angle between the plough and fine Congleton sand is 26◦ and the interface friction force is the product of the plough’s weight and the tangent of the interface angle and is 8 N. Figure 6 shows the tow force generated during the shallowest test conducted, plough depth, D = 19 mm model scale (or 50 × 19 mm = 0.95 m prototype scale) is only 9 N and therefore almost entirely due to interface friction. The deepest test (D = 41 mm) by comparison achieved a tow force of 20 N and therefore the base friction accounts for less than half of the total in this case. The second component of tow force during ploughing in dry sand is a function of the cube of the plough depth and is influenced by soil properties such as unit weight and angle of friction. Figure 6 shows the steady state tow force increasing with depth for two series of tests, one with and one without a forecutter. The force prediction lines in Figure 6 were generated by Eq. 1, where Cw = 0.48 and was the measured interface coefficient of friction. Cs is an arbitrary multiple of unit weight which best fits the data for each series and is

© 2011 by Taylor & Francis Group, LLC

867

Figure 8. Comparison between ploughing tests with and without a forecutter in saturated fine sand. Figure 7. Variation of tow force during saturated ploughing tests without a forecutter.

taken as 10.5 for tests without the forecutter and 12 for tests with the forecutter. Note that the value of Cw used to best fit the data is similar to Cathie & Wintgens (2001) published value of 0.4. The values of Cs used here are greater than the Cathie & Wintgens (2001) published value for loose sand of 5 and fit between their values of medium sand (Cs = 10) and dense sand (Cs = 15). The reason for the difference in Cs values could be due to the inaccuracies in deriving empirical parameters from field tests where soil conditions may not be known over large distances between discrete investigation points.Additionally the scale model results may produce higher than expected tow forces due to the low effective stresses resulting in increased dilation and causing the sand to act as if it were in a denser state. To give an idea of the engagement of the forecutter at each plough depth, the forecutter in the model was approximately 12 mm shallower than the share during the tests shown in Figure 6, however this value was sensitive to pitch. The two tests show that the forecutter causes an increase in tow force in dry sand of around 7%, which appears constant over the range of depths tested compared to the tests without the forecutter. It has not been investigated why this is the case but it is thought that the forecutter may increase the tow force by causing some sand to be sheared twice, firstly by the forecutter and secondly by the share. As rate effects are not present during ploughing in dry sand, the dry sand test results allow the static component of tow force to be examined in isolation. Figure 7 shows the force and displacement data recorded during two saturated ploughing tests, each performed without a forecutter and at different velocities of 43 m/h and 159 m/h respectively. The 43 m/h test shows the increase of tow force as the plough translates through transition and into steady state which is very similar to that of the transition and steady state data shown in Figure 3. © 2011 by Taylor & Francis Group, LLC

3.4 Tow force-displacement profiles of saturated ploughing tests The data recorded during the 159 m/h test is more irregular and is typical of saturated ploughing test data at velocities of around 120 m/h and upwards. The reason for this may be due to the unsteady nature of the resistance of shear planes. Strain dependant resistance is magnified by the increase in normal effective stresses due to reductions in pore water pressure during dilation, which then dissipate post peak. Comparison of the irregular data was made easier as the post transition data was averaged from 550 mm horizontal displacement to the end of the test, for all tests.

868

3.5 The effect of a forecutter on rate effects Saturated tests were carried out, both with and without a forecutter at depths of 37 mm and 38 mm respectively over a range of velocities from 28 m/h to 193 m/h in order to investigate how the tow force varied with plough velocity. Figure 8 shows how the steady state tow force varies with velocity for 19 saturated ploughing tests and allows comparison between tests where a forecutter was present and those where it was absent. On initial inspection, the results appear to contain less experimental error during low velocity tests than it does for tests where the velocity is greater than 120 m/h, as suggested by Figure 7. Figure 8 shows the tow force increases in an approximately linear manner with velocity for both test series and therefore linear trend lines were fitted to the data in using least squares regression. The rate effect is reduced for the test series where the forecutter was present and is highlighted by its shallower sloping trend line but increases the tow force at zero velocity. It has already been shown (see Figure 6) that the forecutter increases the tow force in dry sand and therefore it is reasonable to assume that when ploughing at speeds where the rate effect is relatively small the forecutter will increase the force required to tow the

plough. Consequently at velocities less than 65 m/hr the tow force is greatest for tests where a forecutter is present and for tests at velocities above 65m/h the tow force is greater when the forecutter is absent. The forecutter in effect creates a two stage ploughing process by cutting a shallow trench ahead of the share and in doing so reduces the distance between the share tip and the sands’ surface when the share makes its cut. This decreases the length of drainage path from the share and in turn reduces the rate effect by allowing a greater degree of dissipation during shearing. Although the rate effect is reduced in tests where the forecutter is present, the average plough depth was slightly less during tests where the forecutter was present and will have slightly skewed the results, making the forecutter appear additionally beneficial. It is felt however that this small difference in plough depth (3% less with a forecutter) will have little bearing on the results and the overall findings.

4

Figure 9. 50th scale plough data scaled to prototype scale.

IMPLICATIONS FOR PLOUGHING IN SANDS AND SILTS

test depth and Figure 9 therefore represents a plough trenching at 1.9 m in medium dense fine sand. Note that the actual tow forces may be increased slightly because of scale effects (higher φ and ψ because of low stress levels), but the type of effects observed are very likely to occur in full-scale ploughs. The equations of the best fit straight lines in Figure 9 show the tow force (F) as the sum of a constant (which is the result of the ‘static’ resistance to ploughing) and the rate effect, which is the product of velocity (v) and a multiplier. For the test series where the forecutter was used, the multiplier on the rate effect is only 60% of the tests without the forecutter and confirms that a forecutter is of use in reducing rate effects during ploughing. Consequently, it appears that its use is beneficial in soils where significant rate effects (and so high tow forces) are anticipated.

4.1 Limitations The results displayed in the previous section show clearly the effect of the forecutter on the performance of a 50th scale model pipeline plough. However for the results to be used directly by industry there are two main limitations that need to be addressed. Firstly all tests were performed in medium dense, fine Congleton sand, which was selected on the basis that fine sands and silty sands appear to cause the greatest problems to industry. This was noted by Reece and Grinsted, (1986) and is due to the fact that they are of relatively low permeability and readily dilate upon shearing. It would therefore be desirable to find out how sensitive the results are to permeability and also relative density. For example, ploughing in a slightly coarser sand may produce a force-velocity graph which shows ploughing with a forecutter will increase tow force at any velocity due to the rate dependant part of the overall tow force being small in comparison to the static component. The second limitation of this study is the reduced scale of testing and it is necessary to investigate how the rate effect might scale with plough size.

4.2 Use of a forecutter during ploughing Figure 9 shows the 50th scale plough data with the tow force scaled up to prototype scale. The mass of the model plough is 1/503 of prototype, the volume of sand affected by the model is assumed to be 1/503 of the prototype. Shear forces are assumed to be 1/503 the prototype assuming a Coulomb failure envelope, as depths and therefore effective stresses are 1/50 times those of the prototype and act on planes 1/502 the area of the prototype. This allows the assumption that tow force can simply be increased by a factor of 503 . The scaled depth of the tests is simply 50 times the model © 2011 by Taylor & Francis Group, LLC

5

CONCLUSIONS

A series of model scale pipeline plough tests have been conducted to investigate the effect of the forecutter on plough performance. The main findings were that: 1. The forecutter has a fairly minor impact on the pitch and overall stability of the plough. 2. Although the forecutter is relatively small in comparison to the share, its presence does cause a measureable (7%) increase in the ‘static’ component of tow force (see Fig. 5 and Fig. 7). 3. The forecutter is effective in reducing the rate effect during ploughing. 4. As the reduction in tow force achieved by its use is offset by an increase in the static component of tow force, it will only be beneficial for greater plough speeds. Further work is required to investigate how these speeds change with trench depth, plough scale and soil conditions.

869

ACKNOWLEDGEMENTS The first author is supported by an EPSRC DTA award with additional funding and technical assistance for this project provided by CTC Marine Projects Ltd, Darlington, England. REFERENCES Bolton, M. D. (1986). The strength and dilatancy of sands. Geotechnique 36, No. 1, 65–78. Bransby, M.F., Brown, M.J., Hatherley, A. & Lauder, K. (2010) Pipeline plough performance in sand waves. Part 1: Model testing. Canadian Geotechnical Journal. Vol. 47, No. 1. pp. 49–64. DOI: 10.1139/T09-077. ISSN 0008-3674. Bransby, M.F., Brown, M.J., Lauder, K. & Hatherley, A. (2010) Pipeline plough performance in sand waves. Part 2: Kinematic calculation method. Canadian Geotechnical Journal. Vol. 47, No. 1. pp. 65–77. DOI: 10.1139/T09091. ISSN 0008-3674.

© 2011 by Taylor & Francis Group, LLC

Cathie, D. N. & Wintgens, J. F. (2001) Pipeline Trenching Using Plows: Performance and Geotechnical Hazards. Proc. Thirty-third Annual Offshore Technology Conf., Houston (13145) April/May 2001, 1–14. Grinsted, T.W. (1985): Earthmoving in Submerged Sands, Unpublished Ph.D. thesis, University of Newcastle upon Tyne, UK. Hata, S. (1979). Submarine cable: multi-blade plough Geotechnique 29, No.1, 73–90. Lauder, K.D., Bransby, M.F., Brown, M.J., Pyrah, J., Steward, J. & Morgan, N. (2008) Experimental testing of the performance of pipeline ploughs, Proc. Eighteenth (2008) Int. Offshore and Polar Engineering Conf. Vancouver, Canada, July 6–11 2008. Palmer, A. C., Kenny, J. P., Perera, M. R. & Reece A. R. (1979). Design and operation of an underwater pipeline trenching ploughGeotechnique 29, No. 3, 305–322. Reece, A. R. and Grinsted, T. W. (1986). Soil Mechanics of Submarine Ploughs Proc. Eighteenth Annual Offshore Technology Conf., Houston (5341) May 1986, 453–461. Wood, D. M. (2004). Geotechnical Modeling. Oxfordshire: Spon Press. 246–258.

870

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

State-of-the-art jet trenching analysis in stiff clays J.B. Machin & P.A. Allan Geomarine Limited, Newcastle, UK

ABSTRACT: Apparently simple upon first inspection, following careful analysis the water jet excavation of stiff clay soils is found to be a complex process. Due to the early lack of technical data, engineers working in the field of soil jetting first used empirical approaches and rules-of-thumb to quantify and model the process. Over the last ten years or so more data has become available, both from research work and actual field operations. As a result, a more scientific approach to the problem is now available. A design methodology is described in this paper. It has recently been applied by the authors to the design of a number of jet trenching and excavation machines and in the assessment of a number of jet trenching and excavation projects in stiff clay soils.

1

INTRODUCTION

Water jetting is widely used in applications where underwater cutting and removal of a material is required. Typical applications include cutting of steel and concrete, rock cutting and excavation, surface cleaning, and soils dredging. Underwater jetting is rather less energy efficient than jetting in air because of the large amount of mainly useless viscous shear currents that are created by a turbulent jet prior to impact. However, in many cases this draw-back is more than countered by two major advantages: The first being the perception of relative safety and reduced likelihood of accidental damage associated with use of water jets. The second is the assumption that jetting is a fairly simple operation requiring equipment that uses only a small number of moving parts. Underwater soils jetting is becoming an increasingly widespread construction technique. Engineers involved, among them the authors, have found that a careful jet nozzle design that optimizes jet stream pressures and velocities, while minimizing energy requirements, dramatically improves system efficiency. The result is improved productivity as well as permitting new soil types and strength classifications, for example stiff to hard clays, to be excavated that have not previously been possible.

2

CHARACTERISTICS OF SUBMERGED WATER JETS

Figure 1 shows the idealized penetration of a submerged water jet impinging on a clay stratum. Standard references on fluid mechanics e.g. Blevins (1984) provide solutions for the propagation of such continuous non-cavitating submerged water jets. In an initial region, close to the jet nozzle, the velocity and © 2011 by Taylor & Francis Group, LLC

871

Figure 1. Idealized soil penetration of submerged water jet.

pressure in a “potential core” are approximately constant across the cross-section. This is followed by a transition region which is then followed by a region of fully developed flow. Here the jet entrains surrounding fluid via a viscous shear layer which leads to loss of jet energy. The jet diameter increases with distance from nozzle, and the jet velocity and pressure decay with distance from nozzle. It should be noted that these jets under analysis are generally high speed turbulent jets which possess a Reynolds Number greater than about 3,000. Figure 2 taken from Blevins (1984) shows experimental data of the decay in centre-line axial velocity along a submerged water jet versus is axial distance. Based on this and similar data which confirms the relationship, the authors have adopted a model of velocity ratio V/V◦ versus normalized distance d/D◦ from nozzle, where V is the jet velocity at distance d

where P is the pressure on the area of the small element, ρw is the mass density of the jet fluid, V is the axial velocity of the water jet, and f() is a function that expresses the velocity distribution of a submerged water jet. On the question of the shape of velocity distribution, as part of their research into the efficiency of jet mining of China Clay deposits in England, Davies and Jackson (1981) measured the impact pressure distribution underneath the jet impact zone and found it to resemble a “bell” shape. If the impact stagnation pressure is equated to the bearing capacity of the soil excavation face by q = Nc Su for a cohesive soil, then the threshold velocity required to cause bearing capacity failure follows: Figure 2. Decay in centre-line velocity, Blevins (1984).

where V is in m/s, Su is in kPa, and the approximation has been obtained by putting Nc = 6 and ρw = 1Te/m3 . An effect of a bearing capacity failure will be the development of some form of depression, hole or cavity in the soil. This will alter the flow characteristics and pressures. As the water turns and flows laterally over the soil, erosion due to viscous shear will occur. Equation (3) is of significance in that it determines the minimum required operating point or threshold for an underwater jetting system to begin to cut into a cohesive soil of known undrained shear strength. Equation (3) can be combined with Equations (1) and (2) to give us a value for the total depth of cut as a function of pressure:

Figure 3. Initial jet impact behaviour.

from a nozzle of diameter D◦ , and V◦ is the jet exit velocity at the nozzle. V is assumed constant up to six diameters from the nozzle, and then decreases with distance according to:

where d > 6 D◦ . Figure 1 shows a divergent fully developed jet angle forming a conical shaped cavity. The authors have found from experience that this angle is typically about 14 degrees. However the angle does seem to vary somewhat according to different authors although experience suggests that this typical value works well with the range of nozzle types normally used. 2.1

Where d is measured in metres and nozzle pressure P is measured in bar and a nozzle coefficient of discharge of 1.0 is assumed. Equation (4) is used to check that jetting systems have at least a minimum necessary ability to excavate clay of a given shear strength Su. For example, the equation suggests that a well designed nozzle of pressure 10 bar and diameter 20 mm can cut up to 15 cm into clay soil of shear strength 100 kPa. The equation can also be correlated with in-situ Cone Penetration Test (CPT) measurements of shear strength.

Initial jet impact behavior

When the high speed jet impacts on a perpendicular soil face the water velocity in the direction of travel is brought to an abrupt halt and even partially reflected backwards. A stagnation pressure must therefore arise at the point of impact. To again escape from the excavation face water will then also try to deflect sideways. Figure 3 illustrates this process, again after Blevins (1984). From classical fluid mechanics theory, the equation for the stagnation pressure function exerted on a small element of a plane perpendicular to the jet axis is:

© 2011 by Taylor & Francis Group, LLC

872

2.2

Correlation with test observations

Data was gathered and back-analyzed from four major jet impact testing research programs by Machin et al. (2001). These programs were carried out under controlled conditions in clay soils with strengths in the range 2.5 kPa to 120 kPa. Several hundred individual tests have been performed and back-analyzed. All of these tests to date have confirmed that upon impact the resulting soil cavity depth develops approximately as represented on Figures 4 and 5 which respectively illustrate the behavior of cavity formation both under a static nozzle and under a traversing nozzle of constant translational velocity.

Figure 6. Jetting arm tool nozzle array for V shaped trench.

to “hydro-fracturing”. It results in a rough-looking and uneven cylindrical cavity surrounded by a wide zone of disturbance. It is often not fully cleared of soil lumps.

Figure 4. Observed cavity formation with static nozzle.

Figure 5. Observed cavity formation with translating nozzle.

Some notable observations that were reported from these test results can be summarized as follows: 1. During the static nozzle tests the “Quasiinstantaneous” cavity depth was found to correlate very closely with the threshold bearing pressure theory of Equation (4). The cavity appeared to form in a very short time interval indeed (estimated as a fraction of a second). 2. During static tests a full cavity depth generally develops over several seconds. It is greater than the depth of the initial “Quasi-instantaneous” cavity and its depth varies with clay properties. 3. During the traversing nozzle tests it is observed that the “Quasi-instantaneous” cavity depth consistently occurs once translational speeds exceed about 0.1 to 0.5 m/s, depending on clay properties. For lower translation speeds deeper cavities are formed. 4. A clear cylindrical cavity is usually formed by the water jet. It is either vertical or bell-shaped and has fairly smooth sides. Minimum diameter or width is always greater than the nozzle diameter and is often recorded at up to about 3 times the nozzle diameter. 5. Occasionally during testing a different type of failure is observed. It resembles a soil upheaval due © 2011 by Taylor & Francis Group, LLC

Based on the above observations, the authors have concluded that for conservative jetting assessment purposes a key design approach should be as follows: In cases where cutting of the soil is required the jet must be demonstrated as able to form a sufficiently deep cavity according to Equation (4) which defines a “Quasi-instantaneous” depth of cut. It could be argued that this rule is over-conservative in that it does not account for the additional timedependant cavity propagation effect that is always observed. However, it is tentatively suggested that this is a soil softening and particle erosion effect with a magnitude that depends on the property of permeability – a property which is often difficult to predict with accuracy. Another issue which needs to be addressed is the use of the constant bearing capacity factor (Nc = 6) irrespective of depth. This contrasts with established factors for deep penetration bearing (e.g. Nc = 9) and for CPT’s (e.g. Nc = 12). Here the authors’ supposition is that the local erosive nature of the jet failure mechanism results in a shear failure process that is more representative of a shallow penetration than a deep one even at the base of the cavity. However this is clearly an area not yet fully understood. 2.3

Dis-aggregation in clay soils

The process described above illustrates how the excavation face of a clay soil is penetrated and cut up with water jets to form cavities. The linking up of these cavities can then act to form a “cookie-cutter” grid that results in the cohesive soil being continuously extruded into blocks. In order for mass excavation to occur it is vital that these blocks must then disaggregate (separate) into lumps that can be efficiently transported. Figure 6 illustrates a jetting arm tool with a nozzle array designed to cut the soil into blocks and form a “V” shaped pipeline trench. Therefore, the second part the jetting model is to establish that the soil will dis-aggregate into lumps that are suitable for removal. The authors propose that

873

there are in fact only two driving forces available for dis-aggregating the lumps of cohesive material and they are: 1. Self-weight/gravity. 2. Water flow that entrains the lumps with a viscous shear. On a vertical excavation face the clay extrusions will always tend to form blocks that cantilever outwards from the face. The depth and breadth section dimensions of the cantilever blocks will depend on the spacing between nozzles on the high pressure jet array. The effect of water flow around the clay bricks serves to provide an additional load that further contributes to their instability.

Table 1.

Clay Plasticity Classification

Plasticity Index (%)

Very Low Low Moderate High Very High Extremely High

<6 6–10 10–25 25–50 50–100 > 100

2.5

Hydro-fracture and fluidization behavior

When pressurized water finds its way into voids within the excavation face a different type of soil failure © 2011 by Taylor & Francis Group, LLC

Water Content Range for Good Workability (%) <5 5–10 10–20 20–30 > 30

known as “hydro-fracture” can occur. Given the right conditions this mechanism will open up incipient fractures within the soil, thus forming voids and cavities. Hydro-fracture is a different mechanism to the bearing failure approach. Its occurrence during testing is described earlier in this report. In certain rare cases (e.g., the over-consolidated clays around the Gullfaks site in the North Sea), the authors have anecdotally observed that natural preexisting fractures seem to be present within the soil to such an extent that hydro-fracture mechanisms apparently occurs with hardly any need for bearing failure type jetting. Unexpectedly high overall excavation productivity appears to occur. However, more typically the authors have found that hydro-fracturing effects provide a welcome, if often unexpected, supplement to conventional mechanisms. Hydraulic fracture in cohesive soil is theoretically not time-dependent. In practice, a hydraulic fracture needs to be initiated by water flow into a cavity or highly permeable layer and this obviously requires a finite duration in which to occur. The fracture may then develop provided that the water pressure is maintained at a value sufficient to continue to expand the flow of water into the fracture. Simple mathematical relationships have been developed to assess the initiation of hydro-fracturing such as Farmer (1983). The applicable equation for the pressure required is:

2.4 Tensile strength and plasticity effects From a structural analysis view-point, a rigorous solution to the dis-aggregation problem therefore requires knowledge of the propensity of these cantilevering blocks of soil to collapse. It therefore requires knowledge of the tensile strength property of the clay. Unfortunately information about clay tensile strength is rarely provided in traditional geotechnical data surveys. Furthermore, the parameter seems to be rather difficult to predict. Literature cites field values ranging from a theoretical maximum of twice undrained shear strength for a highly plastic clay, to one tenth of undrained shear strength (or even less) for clay close to or below its plastic limit. Tensile strength relies on the negative inter-particle pore pressures that remain in the clay blocks after the excavation face has been cut up by the high pressure jets. These negative pore pressures will be released as pore water migrates, resulting in softening at a rate dependent on permeability and lump dimensions. As a consequence tensile strength is also influenced by the presence of natural fracturing, “toughness” and “brittleness” of the soil, factors which are difficult to quantify but are reflected to a degree in its plasticity description. Heavily over-consolidated stiff to hard clays with undrained shear strengths of about 100 to 170 kPa or greater can really only exist at or below their plastic limit (Skempton & Northey1952). In many ways they behave like weak rock and tensile strengths in the range one half to one tenth of undrained shear strength could be tentatively predicted. Such readiness of the soil to break up into small lumps upon loading is also known by agricultural engineers as its “workability index”. The authors have found that referencing studies of this property have also proved very useful in assessments of disaggregation and predictions of dis-aggregated soil lump sizes formed. Rosa et al. (2003) have tabled the following model:

Soil Workability Index, Rosa et al. (2003).

Where P = applied pressure, σ = minor principal stress, T = tensile strength of material. In the case of jetting, the orientation of the minor principle stress will affect the direction along which the fissure opens. Depending on the in situ stresses in the soil, the orientation of the minor principle stress would normally be vertical in a stiff over-consolidated clay. A higher stress will be present in the horizontal directions. However, as the excavation face advances, the horizontal stress along its axis will be relieved and this will be become the minor principal stress. Hence jetting into this face will open a fissure in a plane perpendicular and the face of the excavation will fracture as clay blocks are broken. It is recommended that for each jetting project a calculation using Equation (5) should be made to check the limiting value of soil tensile strength below which a

874

to Hjulstrom (1935, 1939) and Shields (1936) previously cited, the authors use data provided by Wakefield (1997) and Wilson (1979). These papers include charts that give guidance on the effect of varying drag coefficient and shape factor for different soil particles in varying flow regimes.

4

HYDRAULIC FLOW EFFICIENCY

significant hydro-fracture event can be expected to initiate. This can be checked against the predicted value of Tensile Stress as discussed earlier. The pressure of the jetted water can also cause the soil to “fluidize”. Fluidization occurs when the soil pore pressures increase to equal the total external pressure, resulting in zero inter-locking shear forces between soil particles and complete dis-aggregation of the soil matrix. At this point the soil shear strength and tensile strength also reduce to zero. Fluidization may well occur during jetting in loose silts and silty sands, and soft clays near their liquid limit. The authors have observed fluidization during the jetting process at many different sites. A good example is the extremely “fat” seabed clay of the Green Canyon zone of the Mississippi Delta, in the deep water Gulf of Mexico.

The effect of nozzle geometry definitely makes a difference. Woodward (1985) has made an illuminating comparison of the pressure profiles of a number of commercially available water jetting nozzles. An efficient nozzle geometry is designed to minimize energy losses and give a discharge coefficient (Cd ) of typically 0.9 or better. The effect of straight section of the nozzle is also known to be important and both very short, or very long sections are likely to significantly increase energy loss. It is also of high importance that the hydraulic flow within the jetting arms and nozzles is well conditioned and streamlines are smooth and direct with no rapid changes of direction, eddy’s or other energy wasting diversions. The authors have found that the use of correctly designed flow straighteners is a highly effective way of improving the hydraulic flow efficiency of jetting arm designs. Flow straighteners are thin blades that are used to straighten hydraulic streamlines at the nozzle inlet.

3 TRANSPORTATION OF SOIL

5

After the phases of initial jet impact and soil disaggregation, the final stage that completes the analysis of the jetting process is an assessment of the transportation of the soil lumps and particles. A sketch of the overall jetting process is shown on Figure 7. Transportation of jetted soil particles and lumps is carried out by a turbid hydraulic flow. One source of this flow is from the water jets used to cut the cavities in the excavation face. Residual flow from these jets is deflected back from the excavation face and can then be harnessed for soil particle fluidization and lump erosion. At some distance away from the excavation face the turbid hydraulic flow will slow down to a quasiquiescent state. In this region soil particles and lumps are steadily re-deposited onto the seabed, at a rate controlled by the fall speed of the particles and lumps within the mixture. In between these two regions the fluidized layer always undergoes a steep and sudden shock-like transition as it changes from supercritical (high Froude number) to sub critical (low Froude number) flow. The transportation mechanism is well understood although, unfortunately, is rather complex to solve in closed form. However, there is a large amount of data and helpful charts in the literature. In addition

The authors have found that a soils jetting system can be systematically designed to provide a predictable and repeatable jet excavation performance in soil. A design procedure has been described. However, the authors have found that a great deal of uncertainty still remains in the detailed analysis of the soil jetting mechanism. Further work is definitely necessary if the industry wishes to further enhance the capability of soil jetting systems. In particular it is highlighted that soil softening and particle erosion effects, and their inter-relationship with permeability and plasticity, are currently not well understood. Also the role of the tensile strength parameter and its accurate determination is a matter of uncertainty. Additional testing programs and new types of in situ and laboratory testing are likely to be required.

Figure 7. Illustration of overall jet excavation process.

© 2011 by Taylor & Francis Group, LLC

CONCLUSIONS

REFERENCES Blevins, R.D., 1984, Applied Fluid Dynamics Handbook, Van Nostrand Reinhold Company. Davies, T.W., and Jackson, M.K., 1981, Optimum Conditions for the Hydraulic Mining of China Clay, Proceedings of the First USWaterjet Conference, Golden Colorado. Water Jet Technology Association.

875

Farmer, I., 1983, Engineering Behavior of Rocks. Chapman and Hall. Hjulstrom, F., 1935, Studies in the morphological activity of rivers illustrated by the River Fyris, Bulletin 25, Geology Institute of the University of Upsala, 221–258. Hjulstrom, F., 1939, Transportation of detritus by moving water”, Symposium on Recent Marine Sediments, ed.P.D.Trask, Specialist Publication of the Society of Economic Paleontology and Mineralogy, Tulsa USA, Vol. 4, 5–31. Machin, J.B., Messina F.D., Mangal J.K., Girard, J., Finch, M., 2001, Recent Research on Stiff Clay Jetting, Offshore Technology Conference, OTC13139, Houston, Texas. Rosa 2003, D. de la, Mayol, F., Diaz-Pereira, E., 2003, Aljarafe Model Soil Plasticity and Workability, Institute for Natural Resources and Agrobiology, Seville, Spain. Shields, A., 1936, Application of similarity principles and turbulence research to bed-Load movement, California

© 2011 by Taylor & Francis Group, LLC

Institute of Technology, W.M. Kekck lab of Hydraulics and Water Resources, Report No 167. Skempton A.W. and Northey R.D. (1952) The sensitivity of clays. Geotechnique, Vol 3 30–53. Wakefield, A.W., 1997, The Complete Model of the Solidshandling jet Pump, 9th International Conference on Transport and Sedimentation of Solid Particles, Cracow, Proceedings Volume 2. Wilson, K.C., 1979, Deposition-Limit Nomograms for Particles of Various Densities in Pipeline Flow, Proc. Hydrotransport 6, BHRA Fluid Engineering, Cranfield, UK. Woodward, M.J., 1985, An Experimental Comparison of Commercially Available Steady Straight-Pattern Water Jetting Nozzles, Proceedings of the Third US Waterjet Conference, Pittsburgh, Pennsylvania. Water Jet Technology Association.

876

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Numerical modelling of soil around offshore pipeline plough shares W. Peng & M.F. Bransby University of Dundee, Dundee, UK

ABSTRACT: Plough share performance has been investigated by conducting a series of finite element analyses in which the process was simplified to that of a plane strain problem. Fully drained analyses have confirmed the relationship between tow force and trench depth and weight as observed previously during experiments. Analyses conducted to investigate rate effects confirmed that partially drained conditions were achieved and that the force-velocity relationships are unlikely to be linear over a wide range of velocities. 1

INTRODUCTION

Pipeline protection is often achieved through burial in the seabed sediment. One common way that this is achieved is by ploughing (e.g. Finch et al., 2000). A large (e.g. approximately 20 m long) plough (Fig. 1a) is dragged through the seabed, and the plough forms a ‘V’-shaped trench into which the pipeline is placed before usually being backfilled. A large support vessel is needed to conduct pipeline ploughing because as well as significant support infrastructure a large bollard pull is required to move the plough through the seabed. Significant tow force is required firstly because of the size and consequent weight of the plough needed to find a suitably deep trench (typically up to 1.8 m) and secondly because of rate effects. The latter mean that tow forces increase with plough velocity and will limit the achievable plough velocity for a given bollard pull. Accurate prediction of pipeline or cable plough performance, particularly velocity, is critical for offshore contractors as it affects the tender price and the choice of installation method and trenching tool. The variation of tow force with plough weight and trench depth is reasonably well known (e.g. Reece & Grinsted, 1986) and is partly analogous to the lateral capacity of a gravity retaining wall with a base friction and passive lateral pressure term. However, the variation of tow force with plough velocity (the plough rate effect) is less well understood, being believed to depend on the dilation potential of the soil, its permeability as well as the trench depth (Palmer et al. 1999; Cathie & Wintgens, 2001). Currently, methodologies exist which are based on, or calibrated from, backcalculation of previous trenching (e.g. Cathie & Wintgens, 2001). These have limited accuracy because of the uncertainty about and variability of soil conditions in the field. To examine the above issues, a series of finite element (FE) analyses have been carried out. These investigated directly how tow force varies with plough weight, trench depth and velocity. However, because of © 2011 by Taylor & Francis Group, LLC

Figure 1. Plough shape and 2D simplification.

the complexity of three dimensional shape of a pipeline plough, the problem was simplified here for the case of a plane strain (i.e. infinitely wide) share (Fig. 1b). Despite the simplification, the analyses were expected to capture the key aspects of soil behaviour around plough shares during ploughing which provides most of the tow resistance. The paper will explain the methodology used before presenting results for both ‘static’ (drained) and ‘dynamic’ (partially drained) plough speeds. 2 2.1

NUMERICAL METHODOLOGY Introduction

The plough share geometry was idealized to an infinitely wide wedge as shown in Figure 1b. The share

877

is assumed to be cutting a trench of depth D and the share blade angle was denoted as β. The external forces W
and F act on the blade where W
is the vertical component of the plough weight (minus the vertical component of the tow force and load taken by the skids for a full plough) and F is the horizontal component of tow force taken by the share. FE analyses were carried out using the Plaxis program. The share blade was modeled as a stiff, linear elastic material with a rough interface (δ = φ) to model a rigid object, whereas the soil used effective stress constitutive models to describe its behaviour. A range of different blade angles were investigated starting with β = 90◦ to allow benchmarking. 2.2 Soil characterisation Soil was described using a linear elastic, perfectly plastic constitutive law using the Mohr-Coulomb failure criterion. In general, for fully drained analyses strength was modeled using a friction angle, φ and a separate specified dilation angle, ψ, with a small apparent cohesion, c
given to prevent numerical errors at small effective stress. After using this constitutive model for both fully drained and undrained analyses, it was found to be less suitable for partially drained analyses. This was because there was no dilation cut-off so that soil continued to dilate after peak shear stress conditions were achieved. Consequently, later analyses used the ‘Hardening Soil’ model in Plaxis which has the ability to stop dilating after a critical state voids ratio has been reached. The type of calculation method used in Plaxis was ‘plastic’ for analyses of static resistance force both in drained and undrained condition, but ‘consolidation’ for the dynamic force analysis where the speed of model movement was controlled and partial drainage of pore water was modelled. 2.3

Consolidation analyses were required for the partially drained ‘dynamic’ loading conditions. Hence for these conditions sufficient time (one-year) was required after application of W
to ensure full soil consolidation under the plough weight. The final calculation step involved lateral movement of the plough share through the soil. This was done using displacement control whereby only the horizontal displacement component of the vertical, left edge of the share was displaced. The method allowed possible vertical displacements of the share under the fixed vertical load but prevented rotation (which is unlikely to occur in steady-state ploughing conditions). The reaction forces were calculated for the incrementally applied displacement in order to obtain the soil resistance (F) – displacement (u) curve for the share. These confirmed that a steady state condition was achieved for all analyses. For the partially drained analyses, combinations of displacement magnitude and analysis time were selected to specify displacement velocities ranging from 0.0008 m/hr to 600 m/hr. 3 3.1

RESULTS: ‘STATIC’ ANALYSES Introduction

A series of analyses were first conducted for drained conditions. These were design to investigate the reaction forces on the simplified plough share under ‘static’ conditions. Following validation (with β = 90◦ ), the effects of varying the vertical reaction force (or plough weight), W
and the embedment depth were investigated for one share geometry and soil condition before examining the effects of changing share blade angle and soil properties. For static analyses, F is expected to depend on a modified version Cathie & Wingens (2001) equation:

Geometry selection, mesh discretisation and loading

The geometry modeled is shown approximately in Figure 1b. Different analyses investigated the effect of varying both D and β. Both vertical and horizontal soil displacements were prevented at the base and lateral boundaries which were placed suitably remotely as not to affect the plough capacity. The mesh density was chosen following a validation exercise where results were benchmarked against retaining wall solutions. The mesh chosen used 15node elements which were concentrated in the area near the share tip and on the base. Share loading was achieved in three calculation stages. In the first stage, in situ soil stresses were fixed with no displacements occurring. In the second stage, the weight of the plough, W
was applied. As this occurred the plough settled fractionally to reach an equilibrium position with realistic contact stresses between the plough and the share. For the case of fully drained analyses, this occurred instantaneously. © 2011 by Taylor & Francis Group, LLC

where Cw = base friction coefficient (Cw = tan δ, where δ is the interface friction angle); Cs = passive soil resistance (similar to Kp /2); and γ is the effective unit weight of soil. Note that the second term contains D2 rather than the D3 suggested for real ploughs because of the plane strain analysis conducted here. 3.2

Example problem

Results are shown first for a typical drained analysis where the soil displacements and reaction forces acting on the share are investigated in detail. In the analyses a blade of weight, W
= 268 kN/m (27.3 tonnes per m width) with angle, β = 35◦ cuts a trench of depth, D = 1.5 m though soil with unit weight, γ = 17 kN/m3 , angle of friction, φ = 35◦ and dilation angle ψ = 10◦ . Note that this corresponds approximately to realistic conditions except that the buoyant unit weight would

878

Figure 2. Displacement magnitudes at failure (γ = 17 kN/m3 ; φ = 35◦ ; β = 35◦ ; ψ = 10◦ ; D = 1.5 m).

be closer to 9–10 kN/m3 and the share would not be fully rough (i.e. δ = φ here). Contours of calculated incremental soil displacements are shown in Figure 2 at the point of soil failure (when peak soil resistance is achieved). As the share translates to the right (with no rotation because of the displacement fixity), a wedge of soil above and to the right of the blade is pushed upwards. A shear plane inclined at approximately 40◦ to the horizontal forms from the tip of the share blade to reach the soil surface. This agrees reasonably well with the angle = 90–φ/2– β/2 (=90–35/2–35/2 = 45◦ shown dotted in Fig. 2) predicted by Lewis & Coyne (1999), for an inclined blade. Both the normal and tangential reaction stresses acting on the inclined face of the share (Fig. 3a) and the share base (Fig. 3b) at failure were extracted from the output data. For the inclined face the two stress components were obtained from the horizontal, vertical and shear stresses in the plane using Mohr’s circle. These stresses were also integrated numerically to obtain the normal and tangential forces acting on the share base (Nb and Tb ) and share face (Ns , Ts ) as defined in Figure 1b. The vertical and horizontal components of these forces were added and compared to the applied W
and measured total force F respectively with good agreement confirming that the method was accurate. Figure 3a reveals a triangular distribution of both normal and shear stresses acted on the inclined share face at failure. This is as expected as there is no normal stress at the soil surface because there in no vertical effective stress, but the vertical effective stress increases with length along the face as the depth in creases. The dotted line on Figure 3a gives an indication of the limiting shear stress acting on that plane calculated from the normal stress and the angle of friction (τn = σn
tan φ). Soil is clearly shearing on the rough face with limiting shear stress conditions as the soil ‘slides’ past the blade. Figure 3b shows that the distribution of stresses on the base of the share is complex. There is a nonuniform distribution of normal stress and, surprisingly, the soil is not at limiting shear stress conditions on the share base. Note that the weight, W
of the plough alone generates an average σn
= 44.8 kPa. The additional normal stress is because the soil reaction forces on the inclined face (Ns and Ts in Fig. ab) give a net downwards reaction thereby increasing the vertical © 2011 by Taylor & Francis Group, LLC

Figure 3. Distribution of calculated normal and shear forces.

Figure 4. Contours of horizontal stresses σyy (γ = 17 kN/m3 ; φ = 35◦ ; β = 35◦ ; ψ = 10◦ ; D = 1.5 m).

force (W
+ Ns cos β – Ts sin β) applied to the share base. Finally, Figure 4 shows the contours of horizontal stresses in the soil when peak soil resistance is mobilized and confirms that the largest soil stresses occur near the tip of the share.

3.3

Effect of plough weight

A series of analyses were carried out in which the deadweight of the plough was varied for a fixed trench depth, share blade shape and soil parameters. A typical result of resistance force against plough weight is shown in Figure 5 for a plough with blade angle, β = 56◦ in soil with φ = δ = 35◦ and ψ = 10◦ . For an analysis with no plough weight, the reaction forces on the share face ensure that there is a tow force required (like a passive pressure on a retaining wall) and the

879

Figure 7. Tow force against trench depth squared (γ = 17 kN/m3 ; φ = 35◦ ; ψ = 10◦ ; β = 56◦ ).

Figure 5. Tow force against plough weight (γ = 17 kN/m3 ; φ = 35◦ ; ψ = 10◦ ; β = 6◦ ).

Figure 6. Force-displacement curves for different trench depths (γ = 17 kN/m3 ; φ = 35◦ ; ψ = 10◦ ; β = 56◦ ; W
= 600 kN/m).

total reaction force increases linearly as the plough weight increases, i.e. dF/dW
= Cw (from Eq. 1). The analyses confirmed that Cw ≈ tan φ = tan δ for the rough plough shares considered for a range of drained conditions. Clearly real ploughs have smoother shares as Cathie & Wintgens (2001) suggested that Cw = 0.4 (or δ = 21.8◦ ). 3.4 Plough share depth Trench depth (or share depth) were investigated directly in a series of analyses where D was varied from 0 to 3 m. The tow force – share displacement curves (Fig. 6) show: (i) that tow force increases with trench depth, and; (ii) the distance required to mobilize peak load increases with D. Note however, that the true displacement to mobilize peak resistance for a plough will depend on the plough kinematics and the generation of steady-state spoil heaps around the share, neither of which is modeled here. Tow force appears to increase with trench depth squared (Fig. 7) as expected for plane strain conditions. The intercept of the trend line at D = 0 corresponds to the base shear condition (F = Cw W
) and the gradient of the line dF/dD2 = Cs γ (where Cs = 41.758/17 = 2.456 in Fig. 7). Note that the tow force result for D = 0 m (Fig. 6) is lower than expected © 2011 by Taylor & Francis Group, LLC

Figure 8. K2p and Kp tan φ plotted against deduced Cs for different angles of friction (β = 56◦ ).

and does not follow the trend shown in Fig. 7. This requires further investigation, but is not important for realistic plough geometries where D > 0. Additional analyses in which the unit weight of the soil was varied for a given D confirmed equation 1 and allowed separate calculation of Cs for given soil conditions with excellent agreement. 3.5

Change in angle of friction

A suite of finite element analyses were conducted in which Cw and Cs were deduced for set of FE analyses each with different angles of friction. Cw was confirmed to vary with the interface friction angle (Cw = tan δ). The deduced values of Cs are plotted against K2p and Kp tanφ in Figure 8, where the passive lateral earth pressure coefficient, Kp = (1 + sinφ)/(1 − sinφ). The linearity of both relationships suggests that these terms could be incorporated in expressions for Cs . 3.6

Change in share blade angle

Analyses were repeated for one angle of friction (φ = 35◦ ; ‘medium dense sand’) with different blade angles to determine how Cs varied with blade angle (Fig. 9). As expected, the tow force (and so Cs ) increases as the share blade become less sharp (i.e. β increases). The exact form of the relationship for

880

Figure 9. Variation of Cs with blade angle.

different soil conditions requires further study (for example the large increase in Cs from β = 56.5 to 61.5◦ ), but Cs appears relatively insensitive to blade angle at the low blade angles used by ploughs. 3.7

Summary

In summary, drained analyses confirmed Grinsted’s (1985) and Cathie & Wintgens’ (2001) equation for static ploughing (Eq. 1). The base shear resistance term appeared to give dF/dW
= tan δ as expected, but the Cs term was a more complex function of φ. From the study of one blade angle it appeared that Cs linearly depended on Kp tan φ or K2p with further investigation required to ascertain the generality of this finding. Finally, the effect of the blade angle, β is unclear and it not accounted for in Cathie & Wintgens’ (2001) model where it is implicitly assumed that this angle is constant for all ploughs. 4

RESULTS: RATE EFFECTS

Following the drained calculations, additional analyses were conducted to investigate how tow force changes with velocity and soil permeability. For these analyses the trench depth, D, plough weight, W
and blade angle, β were kept constant together with the soil properties. Soil was modeled using the Hardening soil model with φ = 35◦ and a dilation cut-off was used. Consolidation analyses were conducted to allow calculation of excess pore pressure migration. A wide range of velocities from 0.0008 m/hr to 600 m/hr were investigated for soil with three different permeabilities, k = 0.42 m/hr, 0.042 m/hr and 0.0042 m/hr. These permeabilities represent soils ranging from fine sand to silt. Cathie & Wingens (2001) extended Eq. 1 to include rate effects:

where Cd = dynamic rate term; and v is the plough rate. For plane strain conditions, it is not clear whether the second term should contain D or D2 (the D2 term suggested is an empirical fit to data.) © 2011 by Taylor & Francis Group, LLC

Figure 10. Tow force against velocity for different soil permeabilities.

Figure 10 plots steady-state tow force against velocity as calculated using FEA. When velocity is plotted linearly (Fig. 10a) there is no clear similarity between the curves. In contrast to equation 2, all three are nonlinear, although each may be considered to contain an almost linear section at relatively low velocities. However, when plotted with a logarithmic velocity scale (Fig. 10b), the similarity of each curve becomes clearer. At low velocity F hardly varies with increased velocity (dF/dv = 0) and has the same value for each permeability: this is fully drained behaviour and the drag forces will be governed by the ‘static’ part of the tow force equation (Eq. 1). At very high velocity, the tow force approaches another, much higher plateau force where dF/dv = 0: this is fully undrained behaviour. In the section between the two extremes, partial drainage occurs. The form of the curve is similar to that observed for shallow foundations (Finnie, 1993), penetrometers (Silva et al., 2006) or pipelines (Bransby & Ireland, 2009) displaced at different, constant velocities through soil, where velocity was normalized by the coefficient of consolidation, cv , divided by the drainage path length, D viz. vD/cv . For the analysis with identical geometries (so D is constant) and soil stiffness (so Eo is constant), the dimensionless group vD/cv will vary directly with v/k. Alternatively, Palmer (1999) suggested that rate effects in ploughs depend on vS/cv , where S is the dilation potential. Again, this suggests that for otherwise identical soils apart from their permeability, the rate effect will depend on v/k. Consequently, Figure 11 re-plots the data with the velocity normalized by the permeability. The normalization is successful as the data reduce to one line.

881

effects predicted (far larger than that observed experimentally or in the field) which may be due either to the soil constitutive model used or the two-dimensional nature of the analysis. Furthermore, additional analyses are required to investigate the effect of varying the trench depth, D. 5

CONCLUSIONS

A series of finite element analyses have been conducted to investigate the behaviour of plough shares ‘cutting’ soil during ploughing. Although the analyses have considered a simplified wide (plane strain) plough share only, the results have allowed investigation of soil behaviour during both ‘static’(drained) and ‘dynamic’ (partially drained) ploughing. The results show that the base friction component of loading is approximated well in existing solutions (e.g. dF/dW
= tan δ is appropriate), but that the depth dependent term (Cs γD2 ), and particularly the prediction of the rate component of drag resistance may need improvement. REFERENCES Figure 11. Tow force against normalised velocity for different soil permeabilities.

The force-normalised velocity relationship shown in Figure 11a could be approximated by a linear relationship for v/k < 25 (see dashed line on Fig. 11a). However, this is a very low velocity range. For example for a soil with d10 = 0.1 mm (a value close to the mean of the field data shown by Cathie & Wintgens, 2001), v/k = 25 corresponds to a velocity, v ≈ 9 m/hr compared to likely real ploughing rates of several hundreds of metres per hour. Likely plough progress 100 m/hr < v < 1000 m/hr corresponds to 140 < v/k < 2800 for the example soil given above. This might suggest that the linear forcevelocity section fitted by field data may trend towards the almost undrained section of the line (dotted in Fig. 10a) and produce an over-estimate of the static component as indicated. The above findings should be treated with some caution. There appears to be very large dynamic rate

© 2011 by Taylor & Francis Group, LLC

Bransby, M.F. & Ireland, J. 2009. Rate effects during pipeline upheaval buckling in sand. ICE Geotechnical Engineering 162(5): 247–256. Cathie, D.N. & Wintgens, J-F. 2001. Pipeline trenching using plows: performance and geotechnical hazards. Proc. Offshore Technology Conference OTC 13145, Houston, May 2001. Finch, M., Fisher, R., Palmer, A.C. & Baumgard, A. 2000. An integrated approach to pipeline burial in the 21st century. Deep Offshore Technology 2000. Finnie, I.M.S. 1993. The performance of shallow foundations in calcareous soil, PhD thesis, University of Western Australia. Grinstead, T.W. 1985. Earthmoving in submerged sands. PhD Dissertation, University of Newcastle-upon-Tyne, UK. Coyne, J.C. & Lewis, G.W. 1999. Analysis of plowing forces for a finite-width blade in dense, ocean bottom sand. Palmer, A.C. 1999. Speed effects in cutting and ploughing. Géotechnique 49(3): 285–294. Reese, A.R. & Grinstead, T.W. 1986. Soil mechanics of submarine ploughs, OTC 5341 (May 1986) Silva M.F., White D.J. & Bolton M.D. 2006 An analytical study of the effect of penetration rate on PCPTs in clay. Int. J. Num. Anal. Meth. Geomechanics 30: 501–527.

882

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Anchor–chain–rockfill–soil interaction: Evolution of design methods H. van Lottum & H.J. Luger Deltares, Delft, The Netherlands

ABSTRACT: Deltares (former GeoDelft and Delft Hydraulics) has carried out different kinds of research in the area of anchor – rockfill – soil interaction. The paper gives an overview of the research that was carried out during the last decade by Deltares on the encounters of anchors and rock fill berms and the employment and embedment of standard shipping anchors as well as high holding power offshore anchors in the soil. The paper presents a view on future research that is needed to improve the understanding of dragging anchors and pipeline protection systems 1

INTRODUCTION

Many offshore pipelines and cables are located in areas where they are exposed to damage by dragging or dropping anchors. To prevent pipelines and cables to fail in such events different kinds of protection measures are designed. For better understanding the mechanisms that occur in these protection measures and how anchors and chains approach the pipeline or cable, both physical modelling and analytical research is required. Such research leads to improvement of present protection measures and new protection techniques may be developed. In the last decade, Deltares carried out several research projects on anchor–chain–rockfil–soil interaction. Within this research, Deltares developed custom made design tools and gained new insights on mechanisms involved in this interaction. These new insights gave us our view on developments needed to improve description on behaviour of offshore anchors and design of pipeline protection covers. At first recent (model) research will be described, resulting in the authors perspective on the development of future design methods. 2

RECENT RESEARCH

2.1 Lamma pipeline cover verification For the Lamma gas-pipeline near Hong Kong, a verification of the designed pipeline protection was requested. To this end, Deltares developed the calculation model CoverCalc that aims to predict the protection level provided by an arbitrary rockfill cover over a pipeline. Calculations were accompanied by model tests carried out in a large test flume. 2.1.1 CoverCalc CoverCalc was developed in 2004 to enable the comparison of rockfill cover shapes and volumes. © 2011 by Taylor & Francis Group, LLC

CoverCalc calculates the cutting energy of the anchor chain into the moment up to the moment that the anchor arrives at the berm or the chain comes too close to the pipeline. This energy is expressed in the Integrated Protection Coefficient (IPC). Two effects are included in the IPC: the strength of the cover and the position of the area relative to the pipe (follows from the berm geometry). The cutting of the anchor through the cover is conservatively not included in the IPC. CoverCalc relates the IPC (the cutting energy) to the increase of the potential energy (the lifting) of the anchor. The model has been calibrated to experimental data of results of anchors through a rockfill berm in a pre-dredged berm. The maximum IPC, which can be mobilised by an anchor approaching at a certain depth, is determined. This is used to assess whether a given berm will provide enough protection when an anchor approaches. 2.1.2 Experimental research To investigate the behaviour of the anchor during the encounter with the rockfill berm, physical model test were carried out in the Dreding Flume of Deltares. In the Dredging Flume a test section with a length of 16.5 m and a width of 2.4 m was created. In this section, a clay layer was laid with a depth of 1.4 m. The anchors were pulled by the wagon present above the flume. The dragging speed during the tests was 0.44 m/s. Figure 1 shows a photo of a cover after testing in the dredging flume. These tests proved the adequacy of the cover that was designed for the Lamma pipeline, since the anchors passed over the pipeline without causing harm. By varying the berm geometry the tests were also used to assess a possible optimisation of berm width and height to reduce the total volume of required rockfill. As far as the damage mechanism was concerned they showed that most berm damage resulted when the chain and the anchor removed stones from

883

Figure 2. Furrow caused by the moving chain.

Figure 1. Rockfill cover after testing.

the cover and created a furrow across the berm. It also became clear that for the investigated anchor (US Navy Stockless) the inherent instability of the anchor is an important parameter for the protection mechanism: it is not just the chain that is lifted and causes the anchor to pass over the pipeline. A case was recorded where the anchor made a 180 degree roll and surfaced on its own accord, not by the lifting of the chain. We believe that this instability of the anchor is more likely to occur when the berm is approached at an angle such that one of the flukes contacts the berm before the other or when the non-homogeneous character of the berm causes uneven forces on both flukes. At this stage numerical models to take these effects into account are not available.

2.2

Centrifuge test on interaction chain–rockfill

In 2005 Deltares carried out centrifuge tests to obtain both qualitative and quantitative insight in the processes which govern the cutting of a chain into a rockfill berm and the transport of rockfill created by the moving chain. In order to be able to investigate these processes in a representative and controlled manner a system was required which pulls an anchor chain with a controlled speed over and through a rockfill berm. A special designed test setup was designed to carry out these tests (Van Lottum et al. 2007). While moving the chain over the rockfill berm, the chain cut into the berm and was partially covered by berm material. The cutting action of the chain caused a significant amount of stone transport in the rockfill cover, which gradually led to the formation of a kind of trench across the cover and deposition of stones at the location where the chain exited the rockfill berm, see Figure 2. The interaction between chain and rockfill in this transport phenomenon depended on the size of the shackles in relation to the grain size distribution of the rockfill. Also the size of the shackles relative to the size of the berm plays a role as the capacity for transport is proportional with the size of the shackles. If the transported volume is negligible in comparison © 2011 by Taylor & Francis Group, LLC

to the volume of the cover, the transport phenomenon is not playing a significant role in the process and may be ignored. The tests showed that the chain behaves as a special case of long strip foundation. If one looks into the parallel movement (in the pulling direction) and the perpendicular movement (downwards, cutting into the berm) one can assume that per unit length of chain each ratio between these two displacement components will have a unique set of soil response forces. This concept is illustrated in Figure 3. The arrows indicate the ratio between parallel and perpendicular movement. Envelope A could represent the situation that the chain is lying on the berm, since no interaction force between berm and chain develops when the chain moves upwards (is lifted from the berm) and a relatively small force is needed for a pure perpendicular downward movement into the berm. Envelope B represents a situation where the chain is at some depth in the berm, since a breakout force is needed for an upward movement and in general higher forces are required for any of the movement directions, which is the result of the higher effective stress level around the buried chain. These tests showed also that the chain – rockfill size ratio is an important parameter in the penetration of the chain in the berm.The bigger the rockfill size the larger the resistance against penetration into the berm. The penetration of the chain is also characterised by the chain-saw like cutting and material transport process that disintegrates the berm. The correct modeling of material transport by the chain is therefore another essential part of any model, which is to predict movement of the chain trough the berm and to analyse the effect of a rockfill berm as a protective measure. Chain penetration rate is directly dependent on the curvature of the chain over the berm and the chain tension. The development of a model that predicts the performance of a rockfill berm as protective measure requires a proper model for the anchor forces, since these will determine the chain tension and therefore to a large extend the penetration speed.

884

Figure 4. Shear planes in front of embedded anchor.

Figure 3. Concept of yield envelopes for a chain.

2.3 Anchor dragging tests in sand and clay in the centrifuge Deltares designed and constructed a model setup to be able to perform anchor pulling tests at model scale in a geotechnical centrifuge under controlled laboratory conditions. With this test setup anchor pulling tests can be performed up to 1.8 m in a predefined soil body of sand or clay. For a typical test at 100g this represents a dragging distance of 180 m in the prototype situation. During such tests, the soil phenomena and the behavior of the anchor can be monitored by video and various kinds of transducers. In 2008 two preliminary anchor dragging tests with a Vryhof Stevpris anchor were carried out to show the possibilities of the test setup. One test was carried out in sand and one in clay overlying a sand layer. These tests are described in detail byVan Lottum et al. (2010). In both tests, the anchor was placed on top of the sand and was dragged from this initial situation with a speed of 100 mm/min into the soil. The test in sand showed that the anchor slowly digs into the sand creating a pattern of shear planes in front of the flukes see Figure 4. The development of the pulling force shows small drops in the pulling force which correlates with the observed shear planes. The values measured in these test were in agreement with the values mentioned in the Vryhof Anchor Manual (2010). After the dragging test in clay (overlying a sand layer), the clay surface showed a small trough with some heave above the location of the anchor. After exposing the anchor it became visible that anchor had © 2011 by Taylor & Francis Group, LLC

885

Figure 5. Scratch marks of the flukes on the sand layer after removing the clay layer.

not penetrated the sand layer under the clay. Probably the downward force exerted by clay on the flukes is too low to penetrate into the stiff sand. The scratch marks of the flukes showed that the anchor was dragged with the pointed ends of the flukes through the top of the sand, see Figure 5. The test in clay showed that the boundary between sand and the clay limited the penetration depth and therefore the anchor reached an equilibrium drag force at a reduced depth. This illustrates the need for an anchor-soil interaction model that takes the effect of strength variation between subsequent layers into account. 2.4 Anchor dragging test with SEA-anchor In 2008 model tests were performed on suction embedded anchors (SEA’s) in a stiff clay. The (steel) wire that was attached to the SEA cut through the clay and created a plane over which the clay could be split after the test. The strands of the wire left marks in the clay as shown in Figure 6. The path that an anchor follows depends, among other factors, on the interaction between the anchor chain and the soil. The available numerical models

the soil. This shape is derived by an analysis of the equilibrium of forces that act on the chain. The tension force in the chain may vary along the chain when the weight of the chain and the friction between the chain and the soil in the direction parallel to the chain are accounted for. In this type of model, when the soil is homogeneous in the horizontal direction, it can be assumed that the movement of the chain always has a component perpendicular to the chain and a component parallel to the chain. Consequently, it is often assumed that both the full perpendicular resistance and the full parallel friction are mobilized by the chain during the penetration and dragging process. The typical description of the anchor is based on a theoretical or experimental determined capacity that depends on the depth below the surface and the soil properties that are found at that depth. The orientation of the anchor follows from an equilibrium of the weight of the anchor, the chain force exerted at the shackle and the soil resistance.

Figure 6. Travel marks in the clay.

known to date assume that the chain always mobilises the full lateral resistance. The marks in the clay in Figure 6 show that during a phase that the force at the shackle is temporarily reduced, the direction of wire displacement changes to more parallel (slip) and less perpendicular penetration of the wire. In accordance with the concept of a yield-envelope, where parallel and perpendicular reaction forces vary and depend on the direction of displacement, the transitions from I → II → III → IV in Figure 6 coincides with the positions on the yield envelopes 1 – 2 – 2 – 1 in Figure 3.

3

EVOLUTION OF DESIGN METHODS

For the design of mooring systems on the one hand and protection systems for pipelines and cables on the other hand a good understanding of the relevant mechanisms is required. While significant progress has been made on the analytical and numerical descriptions of anchor installation, embedment and holding capacity (Neubecker & Randolph 1996, DNV 2007) physical testing remains very important (Ozmutlu, 2009) as the intricate geometries of the anchor systems do not lend themselves easily for analytical or numerical analyses. The authors feel that the way forward lies in the proper blend and use of numerical and experimental testing tools. Below a description of potential numerical tools and the supplemental experimental activities that provide an interesting perspective for the development of future design method is given.

3.1 “Simple” 2-D models The basic models that are available now describe the path of the anchor chain and the anchor during embedment in a laterally homogeneous soil. In these models, the interaction between chain and soil is characterized by the lateral (perpendicular) resistance that the anchor-chain experiences in the soil, which together with the weight of the chain and the tension in the chain determines the inverse catinary shape of the chain in © 2011 by Taylor & Francis Group, LLC

3.2

Improved chain modelling

Laterally non-homogeneous seabed geometries are needed to model a cover over a pipeline or cable.These, in turn, require an improved chain model: If the chain comes into contact with a stronger part of the subsoil (or a rock cover) the perpendicular resistance that is mobilized by the chain may prove to be well below the maximum resistance that might be mobilized in that soil. Simpler models that always assume the maximum perpendicular resistance to be mobilised will not suffice in that situation. Interaction between chain and soil (or chain and rockfill) needs to be modeled in such a way that not always the maximum perpendicular resistance is mobilized, but that also slip dominated movement can be described. Only then a chain that does not penetrate into the rockfill can be correctly modeled. Also situations where the resistance of the anchor is (temporarily) reduced are characterized by a (temporary) slip dominated chain movement. Reference is made to Figures 3 & 6 for illustration of this concept. This involves the implementation of a “failure envelope” that describes the interaction between the parallel and perpendicular movement of the chain in the soil. The full perpendicular capacity is only mobilized if the chain movement is purely perpendicular and the full parallel friction is only mobilized if the chain movement is purely parallel. Yield envelopes that describe these interactions will show that the soil reaction force on the moving chain is not in line with the direction of the chain movement, but that the ratio of perpendicular to parallel force tends to be larger than the ratio of the perpendicular to parallel movements. While quite experimental work is available that focuses on the lateral resistance of chains in the subsoil (Degenkamp & Dutta, 1989), more work is required to determine the complete yield envelopes of chains

886

and wires in soils and rockfill and build proper yield envelope models. A feature that is more difficult to implement in this type of model is the effect of layer transitions in the direct vicinity of the chain on the yield-envelope. While for chains the over-all effect of such layer transitions may be relatively small, In view of the relative small diameter of the chain or wire the distance over which layer transitions have an influence is small and this effect may be neglected. For anchors (which are much larger) this is not the case. 3.3 Anchor models

3.4 Transport phenomena The effect of rockfill transport from the cover by the chain that is passing over or through the cover was illustrated by centrifuge tests (Van Lottum et al. 2007). If a chain is dragged over the berm for a long enough period, the berm deforms and may further and further be penetrated by the chain. Any structure under the berm, that needs protection, like a pipeline or a cable, would have to be placed in a trench, well below the ground surface. As the change in cover geometry will affect the yield envelope for the chain, the chain-induced transport needs to be accounted for in the models.

Many basic anchor models assume a direction of movement in the direction parallel to the flukes, combined with a rotation that enables the anchor weight and soil reaction forces on the anchor to make equilibrium with the chain force that is exerted on the anchor. The reaction forces may be derived in a theoretical manner, by analyzing the separate contributions parallel and perpendicular to the flukes, crown and shank. When considering the special shapes of some of the existing (and popular) anchors it will be clear that such theoretical analyses may have severe shortcomings. By comparison of actual anchor penetration and dragging tests with prediction of the models the description of the overall anchor-soil interaction may be optimized for a single anchor and a single soil type. Similar to the description of the interaction between chain and soil with a failure envelope a much more versatile modeling of the anchor-soil interaction is possible with a yield-envelope approach. Unlike the chain-soil model, which (in 2D models) only needs two degrees of freedom to describe the relation between lateral and parallel movements and forces the anchor yield-envelope model needs three degrees of freedom, since it also has to describe the interaction with the rotation of the anchor and the resisting moment against this rotation. As an alternative to theoretical derivations of the yield-envelope of the anchor model tests one could use to determine and construct such yield envelopes for arbitrary anchors and ideally create a database with tested and verified yield envelopes for various anchor types. A complication that the anchor yield envelope has is that, much more than the chain-soil models, the anchor “feels” the effect strength gradients, layer transitions and the seabed surface. One approach to deal with this is to “subdivide” the contributing parts of the anchor (e.g. the flukes and shank), into individual parts and assign these parts resistances which reflect the soil that the particular anchor piece is in contact with. An approach along these lines was described by Ruijnen (2005) for a clay that had a shear strength that increased with depth. That there is a need to deal with the effect of layer transitions on the anchor behavior is demonstrated by the anchor pulling experiments described earlier in this paper (Figure 5). © 2011 by Taylor & Francis Group, LLC

887

3.5 The “ideal” 2-dimensional model From the descriptions above the ideal 2-D model for anchor-chain-soil-rockfill interaction can be defined. It contains provisions for: • An arbitrary subsoil geometry. • Chain-soil interaction that is based on a 2 D.O.F.

yield-envelope approach. • Anchor-soil interaction that is based on a 3 D.O.F.

yield-envelope approach and accounts for layer transitions and surface effects. • A way to assess the changes of the resistance in a rockfill berm due to the loss of stones that are being removed from the berm by the chain itself. 3.6

Extension to 3D and discrete element models

Most anchors that are being pulled and penetrate into the seabed will not do this in the perfect vertical plane that forms the space in which the 2-D models are described. Out of plane pulling and rotation or destabilization of the anchor is to be expected in practice. Clearly 2D models are inadequate to describe this. When considering a rockfill protection over a pipeline or cable a perpendicular approach by the anchor will be exception rather than rule. When cutting across a pipeline cover at an angle (not perpendicular) the effective width of the cover increases and the chances that the anchor damages the pipeline reduce. Moreover, apart from the actual resistance that the cover mobilizes against the chain one of the mechanisms that prevents the anchor from penetrating the berm and damaging the pipeline is the destabilisation of the anchor. This can be triggered by unequal forces acting on the flukes of the anchor, which would happen when one of the flukes reaches the cover before the other. Another cause for differential forces lies in the fact that the cover, consisting of relatively large stones, is not acting as a homogeneous continuum but has locally stronger and weaker spots. While the first effect might be handled by a true 3-D anchor model (which requires 6 degrees of freedom) and a true 3-D chain-soil model, for the latter effect the way forward would lie in the employment of Discrete Element Models (D.E.M.). Such models also have potential for a realistic analysis of the rockfill interaction and transport phenomena, which are believed

to have large influence on the protective properties of the cover (Gaudin et al. 2007). 3.7 The “ideal” 3-dimensional model The ideal 3-D model would enable the analysis of anchor break-out, which is typically effectuated by backwards pulling of the chain when done on purpose or sideways pulling when in severe loading conditions one or more mooring lines of e.g. a SBM-system have failed. It would also enable the analysis of a vessel riding on a single anchor in situations where the anchor loading direction changes due to variations in wind or flow direction. The main extension of the 2D model lies in the more complex description of the yield envelopes and the generalized numerical analysis of the anchor chain and anchor in three dimensions. 4

CONCLUSION

A number of improvements are possible in the numerical anchor-chain-soil-rockfill interaction models. These improvements require experimental input and verification, in particular to define proper yield envelopes for anchors and chain, both in soil and rockfill, and for the understanding of the transport phenomena that have been observed. Despite improvements in analytical techniques, verification of actual designs by means of physical modeling will be required in the time to come. There is an obvious conservatism in the assumption that all anchor–pipeline incidents occur at perpendicular tracks. Designers should try to explore this conservatism and verify the extra margin experimentally to arrive at more economic cover designs.

© 2011 by Taylor & Francis Group, LLC

REFERENCES Degenkamp, G &Dutta, A. 1989, Soil Resistances to Embedded Anchor Chain in Soft Clay, Journal of Geotechnical Engineering, (115)10, 1420–1438. DNV 2007, “DIGIN User Manual”, Technical Report 20070882, Joint Industry Project, Det Norske Veritas, Hovik, Norway. Gaudin, C., Vlahos, G., Randolph M.F., 2007, Investigation in centrifuge of anchor-pipeline interaction. International journal of offshore and polar engineering , 17 (1) 67–73. Neubecker, S.R. & Randolph, M.F. 1996, The Performance of Drag Anchor and Chain Systems in Cohesive Soil. Marine Georesources and Geotechnology, 14: 77–96. Ozmutlu, S. 2009. The Value of Model Testing in Understanding the Behaviour of Offshore Anchors: Towards New Generation Anchors. Offshore Technology Conference, 4–7 May 2009. OTC-20035. Ruijnen, R.M. 2005, Influence of Anchor Geometry and Soil Properties on Numerical Modeling of Drag Anchor Behavior in Soft Clay. 1st ISFOG, Perth. Van Lottum H., Luger H.J., Bezuijen A., 2007, Interaction between anchor chains and rockfill tested in a centrifuge model. The Proceedings of the 17th International offshore and polar engineering conference, Lisbon, Portugal, July 1–6, 2007. Van Lottum, H., Luger H.J., Bezuijen, A., 2010. Centrifuge anchor dragging tests in sand and clay. Proceedings of the International Conference on Physical Modelling in Geotechnics 2010. Vryhof Anchors B.V. 2010, Anchor Manual 2010, Capelle a/d Yssel, The Netherlands.

888

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Development of a jet trenching model in sand J-F. Vanden Berghe Fugro Engineers SA, Belgium

J. Pyrah & S. Gooding CTC Marine Projects, UK

H. Capart National University of Taiwan, Taiwan

ABSTRACT: This paper presents a jet trenching model applicable to realistic jetting configurations that has been developed based on earlier research focussing on idealised jetting configurations (Vanden Berghe, 2008). Specifically, oblique and upright swords featuring multiple jets of different orientations are considered here. The swords are modelled after the jetting swords of the jet trenchers operated by CTC Marine Projects for cable and pipeline burial tasks. The jet trenching model has been validated and calibrated on the basis of a set of laboratory small scale tests. The tests were conducted at scale 1:30 with jetting tools modelled after the swords of CTC jet trencher. The experiment analysed the parameters influencing the trenching performance. It included the total jetting power, the progress rate of the trencher and the sand bed properties. Comparison between laboratory and model results were presented and discussed. The results of this research make it possible to better understand the mechanisms controlling the performances of pipeline burial operations by jet trenching. The proposed model helps the engineers in assessing the performance of the jet trenching machines. 1

INTRODUCTION

The main issue of a trenching process is to choose the right trencher and the right trencher configuration in order to optimise the trenching process duration, as well as to use a tool suited for given soil conditions. Usually the aim is to lower the product to a given target depth in the minimum time. The entire trench process duration is function of many parameters, which include: the number of passes required to lower the product, and the trencher speed of advance. This paper presents a jet trenching model assessing the burial depth achieved after each pass of the trencher taking into account the trencher performances (available jetting power and swords configuration), the soil condition (grain size and density) and the product characteristics (weight and stiffness). 2

GENERAL DESCRIPTION OF THE JET TRENCHING MODEL

3

The jet trenching model is based on the combination of the 2 following models: The first model assesses the shape of the trench created by the jet trencher. This model is called the multiple jets trenching model. The assumptions and formulation of the model are described in Section 3 together with the laboratory test results used to validate and calibrate the model. © 2011 by Taylor & Francis Group, LLC

The second model computes the likely shape adopted by the pipeline when the soil supporting it has been removed. This model is called the pipeline model and is described in section 4. The 2 models are combined in order to simulate the trenching process. The interaction between the two models allows the assessment of an appropriate trencher speed of advance and number of passes, for given soil, pipeline and trencher properties. The main assumption of the model presented above is the independence of the trenching model and the pipeline model. The trenching model ignores the presence of the pipeline and the pipeline model does not consider the hydrodynamic forces induced by the jetting process in the computation of the pipeline deflection. This assumption holds for small diameter or heavy products but may not for very large diameter or light products.

3.1

MULTIPLE JETS TRENCHING MODEL Model basis

Several laboratory observations (Su, 2007, Vanden Berghe, 2008) indicate that the interactions between the jet-induced current, ambient seawater and sand seabed can be characterised into five different processes: entrainment, erosion, deposition, breaching, and overspill (as illustrated in Figure 1).

889

Figure 2. Experimental setup (Vanden Berghe, 2008).

Figure 1. Key physical process during jet trenching.

In the longitudinal direction, the dynamics of the jet-induced turbulent current are assumed to be controlled by entrainment of ambient water into the current, erosion of seabed material (water + sediment) into the current, and deposition of material due to the gravitational settling of sand grains back to the bed. From their initial erosion to their eventual deposition, suspended sand grains cause the current density to exceed the density of seawater, and the corresponding density stratification tends to damp turbulent mixing. Weakened eddies then lose their ability to entrain ambient water and erode bed material. The model needs to capture this mechanism. In the lateral direction, the trench cross-sectional shape is affected by the breaching of sidewalls. Breaching is a type of underwater slumping driven by reduced specific gravity and paced by the rate of infiltration of seawater required for sand bed dilatancy to take place. When the sediment-laden current rises above the trench walls, moreover, overspill of suspended sediment out of the trench occurs. This is important in practice because it causes a loss of sand cover over the cable or pipeline, decreasing the level of protection achieved even if a deep trench has been incised. The model development proceeded in the following steps. Based on prior experience with 2D horizontal plane jets (Perng, 2006; Perng and Capart, 2008), a model for a single 3D round jet travelling along the seabed was developed. This model incorporated all five processes (entrainment, erosion, deposition, breaching and overspill) and took stratification into account. This model was validated by comparison with a series of small-scale lab experiments involving point jets originating from a travelling needle (Su et al., 2007; Su, 2008). This model was then extended to multiple jets, injected at various locations and orientations, and coalescing into different currents inside the evolving trench. This in turn requires the model to be able to handle trench cross-sections of arbitrary shape.

the mechanism involved in the trenching process and to guide the development of the theoretical model that would represent these mechanisms as much as possible. The apparatus used for the experiments is illustrated in 3.2.2. A wheeled carriage travelling above the tank at a given speed supports an articulated arm ending with a jetting device. Several jetting devices were used going from a thin syringe needle to swords modelled after the jetting swords of the jet trenchers operated by CTC Marine Projects. More than 100 tests were conducted investigating mainly the influence of the jet configuration (sword inclination, jet orientation, rear jet,…), jetting power, travelling speed, grain size of the sediments and seabed density. All experiments took place under submerged conditions. The conversion from the laboratory scale to the prototype scale assumes a geometrical scale factor of 30. The non-geometrical variables (velocities, discharge and power) are scaled according to Froude similarity, which preserves the relative influence of inertial, density and gravity effects (see Vanden Berghe; 2008). In model tests governed by Froude similarity, problems may be encountered due to the effects of surface tension and/or viscosity. In the present case, these two effects are not expected to cause significant difficulties. The exception regards the possible influence of seabed permeability on the erosion and breaching response of the sand grains. In small-scale tests, it is not possible to simultaneously reproduce permeability and Froude number effects. For this reason, the experiments were conducted with 2 sands, allowing the possible effects of permeability to be estimated and corrected for in the numerical model. Two sand materials were used for the tests: very fine sand (d50 = 0.08 mm) and fine sand (d50 = 0.17 mm). Scaled to prototype size based on the fall velocity of the grains, the very fine sand is equivalent to medium sand (d50 = 0.17 mm) and the fine sand to coarse sand (d50 = 0.99 mm). Additional information on the laboratory tests are provided in Su (2008), Su (2007) and Vanden Berghe (2008).

3.2 Laboratory experiments 3.2.1 Experimental setup (Vanden Berghe, 2008) While much can be learned from trencher operation data, visualisation of the sediment motion is difficult in field conditions. This motivates the use of smallscale laboratory experiments. The focus is to examine © 2011 by Taylor & Francis Group, LLC

890

3.2.2 Experimental results Many tests were performed with different testing parameters.An example of results is shown in Figure 4. For convenience and an easier comparison with actual operational performances, the results are presented at prototype scale.

Figure 5. Shallow layer jet-induced suspension flow model. • The length of open trench varies approximately lin-

early with the total flow rates supplied to the jetting tools. For a given jetting configuration, an increase in flow rate of about 25% associated with doubling the jetting power leads to an increase in trench length of about 20 to 25%.

Figure 3. Picture of the laboratory jetting tool.

3.3

Figure 4. Influence of progress speed and grain size on trenching performance (jetting power = 1.5 MW).

The results of Figure 4 are for inclined jetting swords with vertical downward facing jets. The jetting power is 1.5 MW. The figure compares the results for 2 grain sizes (d50 = 0.17 mm and 0.99 mm) and several progress speed varying between 200m/hr and 1,200 m/hr. The main findings of the experimental campaign can be summarised as follows: • The trench lengths achieved by a given set of jetting

tools depend greatly on sand size and to a lesser degree on the density of the seabed. Finer sands lead to longer trenches by slowing the pace of resedimentation. Denser beds lead to longer trenches by reducing the pace of sidewall breaching and slumping. • Regardless of conditions, giving a downward orientation to the forward jets leads to increased maximum trench depths. However, this increase in depth is limited to the immediate vicinity of the front jetting sword. An increase in localised trench depth does not translate into an overall increase in trench length. • For loose beds, the orientation of the forward jets appears to exert little to no influence on the ultimate length of open trench maintained behind the sword. Various configurations of the front sword lead to identical backfill profiles to the rear of the trencher. © 2011 by Taylor & Francis Group, LLC

Numerical model

3.3.1 Main assumptions The theory envisions a slender current of water and suspended sand flowing along a sand bottom of variable topography submerged in deep water ambient. The following vertical flow structure is assumed (see Figure 5). Motion is confined to a fully turbulent layer of thickness ht . The overlying ambient water is assumed quiescent, and the underlying solid-like sediment bed is assumed static. Embedded within the turbulent layer is a fluidised layer of sand and water of thickness hs , inside which the weight of the dilute sand phase is entirely supported by turbulent eddies. It is assumed that the interfaces between the layers are sharply defined and that the distribution of flow properties (longitudinal flow velocity u, turbulent kinetic energy k and sediment concentration c) is uniform within a given layer. Transfers of mass and momentum between the layers and across the interfaces are however allowed to take place. It is assumed that vertical accelerations are negligible, and that the vertical velocities are cinematically constrained by the thickness variations of the different layers. Vertical velocities are thus allowed, but not explicitly included in the description. The theory thus involves the following five variables, or degrees of freedom (see Figure 5): • Thickness of the turbulent bottom current (hs ); • Thickness of the fluidised layer in which sand is

entrained by the turbulent water (ht );

• Elevation of the loose, static sediment bed (zs ); • Sediment concentration inside the fluidised layer

(Ct );

• Longitudinal velocity of the water and sand inside

the turbulent layer (ut ). 3.3.2 Governing equations All five variables are function of both longitudinal coordinate and time, and their coupled evolution must be described by a set of five governing equations.

891

These equations are obtained by applying mass conservation and momentum balance to the various layers. The different layers defined hereabove do interact. This interaction is characterised by the 3 following exchange fluxes: • The flux ew of the entrainment of quiescent water

into the turbulent current is computed on the basis of common model in hydrodynamics, i.e. proportionally to the current velocity:

where E is the entrainment coefficient, b is the effective width of the turbulent layer flowing along the trench bottom and u is the velocity of the turbulent layer • The flux et between the 2 turbulent layers only depends on the settling velocity of the grains:

Figure 6. Processes influencing the transversal section (a) vertical and lateral erosion of the bottom current (b) deposition of sediments, (c) breaching of the trench wall.

The evolution rules used to update the trench shape under the influence of the various processes are illustrated in Figure 6. Three processes can be identified:

where ωs is the effective fall speed of the sediment grains • The flux es between the seabed and the turbulent current is governed by the following equation

The first term is a turbulent erosion rate, formulated by analogy with turbulent entrainment (i.e. proportional to the current speed), and the second term is a deposition rate due to settling of sediment grains back to the seabed. Dimensionless coefficient ε (0 ≤ ε ≤ 1) depends on the size of the sediment grains and expresses the relative ease with which they can be eroded by the current. Upon erosion, the trenched seabed transfers sand to the current (es > 0), whereas the opposite occurs upon deposition (es < 0). The parameter θ denotes the angle below horizontal formed by the tangent to the bottom profile of the trench Turbulence, which controls both entrainment and erosion, is strongly damped by density stratification. In other words, the erosion flux and the entrainment flux decrease and could possibly be equal to zero when the density and the thickness of the turbulent layer increase. This is taken into account by expressing the entrainment coefficient E as a function of the Richardson number, which accounts for the increase of the potential energy relating to the kinetic energy. 3.3.3 Lateral variation of the trench The last component of the model is the procedure used to evolve the cross-sectional shape of the trench. Because multiple processes are considered, it was found that the trench shape can not be limited to a predefined geometry with limited degrees of freedom (e.g. a trapeze with variable top and bottom widths). Instead, the trench shape should be able to evolve freely. For these reasons, it was decided to characterise the trench transversal profile by a polygonal line allowing, for example, the trench to be wider at its bottom than at its surface, as observed in some experiments. © 2011 by Taylor & Francis Group, LLC

892

• The first process that influences the shape of the

transversal section of the trench is the lateral erosion of the bottom current. As the turbulent current erodes the bottom of the trench, it also erodes the side walls. The lateral erosion is slower than the vertical one and it can reasonably be assumed that it varies proportionally with the trench depth as illustrated in Figure 6a. • The second process is the redeposition of the sediments that is assumed to be horizontal and uniform on the trench width (see Figure 6b) • The last process is the breaching i.e. the collapse of the side wall towards the natural equilibrium slope of the sand. This mechanism acts to widen the trench independently of jetting action and depends on the speed of retreat of a vertical trench wall λ0 , the natural angle of repose of the sand and the slope of the trench wall. Parameter λ0 is dependent on the properties of the seabed material (void ratio and permeability). The approach is similar to the breaching speed formulated in Mastbergen and Van den Berg (2003).

3.4

Comparison of model with experimental data

Model predictions and experimental results were compared. An example of comparison is shown in Figure 7. Both quantitatively and qualitatively, a good level of agreement is recorded between computation and experiment. Trench shapes and dimensions are well reproduced, and their responses to parametric variations are well captured by the model. The same model parameters are used for all jetting configurations, without case-by-case tuning. For the two different sands, only three parameters must be set differently: the single grain fall velocity ωs , the erodibility ε, and the breaching speed λ0 . Provided these values are set adequately, the contrast between the 2 sands analysed is well modelled. For the coarser sand, the trench is shorter, can be trenched at higher progress rates U and exhibits less sand loss out of the

Figure 9. Influence of the lay tension on the pipeline deflection.

Figure 7. Comparison between experimental and model results (jetting power = 1.5 MW – progress rate = 200 m/hr).

resolved analytically and the following equation of the pipeline deflection can be obtained:

where y = Pipeline Deflection (positive downwards) (m), q = Pipeline Submerged Unit Weight (N/m), T = Lay Tension (N), L = Beam Length (m), E = Elasticity modulus (N/m2 ), I = Moment of Inertia of the Beam Cross-section (m4 ) and x = Distance from beginning of Trench (m). As illustrated in Figure 9, taking the lay tension into account is very important to assess the likely shape of the product behind the trencher.

Figure 8. Equivalent pipeline model.

trench due to overspill. The model accurately captures these influences of sand size on the trenching response.

4

PIPELINE DEFLECTION MODEL

In order to assess the burial depth during trenching operations, it is required to compute the likely shape adopted by the pipeline when the soil supporting it has been removed by the trencher. The assessment of the pipeline deflection into the trench is based on an elastic beam model of the pipeline. The pipeline sinking into the trench is assumed equivalent to a hyperstatic cantilever beam uniformly loaded, in which the left-hand end is completely fixed and the right-hand end is restrained in rotation (as illustrated in Figure 8). The left-hand end simulates the pipeline lying on the seabed whereas the right-hand end represents the pipeline touchdown point in the trench. A lay tension is also applied at the right-hand position. After touching the re-sedimentation front, the pipeline displacements are assumed to be restrained. No pipeline settling is assumed in the re-deposited sand that still may be liquefied. The pipeline model described above can be resolved analytically. Because the lay tension in the pipeline applied a bending moment that depends on the pipeline deflection, the formulation required the used of a differential equation. This differential equation can be © 2011 by Taylor & Francis Group, LLC

893

5

CASE STUDY

In an engineering environment the model can be used to either anticipate the performance of a given trencher in a variety of conditions or configure a trencher for optimal use in the soil conditions anticipated at the trenching location. The following case study is presented as an example of the use and effectiveness of the model in configuring a jet trencher for operations on a pipeline trenching project. The pipeline was located in the North Sea, the soil conditions were anticipated to consist of a layer (up to 0.85m thick) of very loose to loose fine silty sand overlying medium dense fine silty sand. The product was a 10” rigid steel pipe, which the client required to lower to 1.0 m below seabed. Several iterations of the model were run to identify the optimum jet leg configuration and the optimum trencher progress rate. The model was run with upper and lower bound soil conditions to account for the vertical and lateral variability of the soils. Figure 10 illustrates the output from the model at the optimal conditions. A depth of lowering of 1.0 m below seabed with at least 0.8 m cover was predicted at a forward progress rate of 275 m/hr. During operations at sea the trencher was operated at a progress rate of as close as possible to 275 m/hr and the product lowered to between 1.0 m to 1.2 m

Based on these observations, a model using the gravity- and jet-driven turbidity currents theory has been developed. This model is able to simulate the experiments quite well and highlights the importance of the different mechanisms. It has also been successfully used for several trenching projects in the North Sea and can be used to assess the best jetting configuration as illustrated in Figure 12. REFERENCES

Figure 10. Illustration of model use for burial assessment.

Figure 11. Depth of lowering achieved.

Figure 12. Illustration of model capability to investigate the benefit of rear jets.

below seabed with a cover thickness of between 1.0 m and 1.2 m. Hence, the model was found to provide a reasonably close but slightly conservative prediction for lowering in this instance (Figure 11). 6

Ahrens, J.P. (2000) A Fall-Velocity Equation. Journal of Waterway, Port, Coastal, and Ocean Engineering, Vol. 126, No. 2, pp. 99–102. García, M.H., and Parker, G. (1991) Entrainment of bed sediment into suspension. Journal of Hydraulic Engineering, Vol. 117, No. 4, pp. 414–435 Hughes, S.A. (1993) Physical Models and Laboratory Techniques in Coastal Engineering. World Scientific. Mastbergen, D.R., and Van den Berg, J.H. (2003) Breaching in fine sands and the generation of sustained turbidity currents in submarine canyons. Sedimentology, Vol. 50, No. 4, pp. 625–637. Perng, A.T.H. (2006) Trenching of underwater sand beds by steadily moving jets. PhD thesis, Graduate Institute of Civil Engineering, National Taiwan University. Perng, A.T.H., and H. Capart (2008) Underwater sand bed erosion and internal jump formation by travelling plane jets. Journal of Fluid Mechanics, Vol. 595, pp. 1–43. Spence, K.J., and Guymer, I. (1997) Small-scale laboratory flowslides. Géotechnique, Vol. 47, No. 5, pp. 915–932 Su, J.C.C., A.T.H. Perng, and H. Capart (2007) Underwater trench incision and turbid overspill due to moving point jets. Proc. XXXII Congress IAHR, Venice, Italy, July 2007. Su, J.C.C. (2008) Seabed trenching by moving point jets: experiments and theory. MSc thesis, Graduate Institute of Civil Engineering, National Taiwan University. Vanden Berghe J-F., Capart H. and Su J.C.C., (2008), Induced Trenching Operations: Mechanisms Involved, OTC-19441, proceeding of 2008 Offshore Technology Conference held in Houston, Texas, U.S.A., 5–8 May 2008, Van den Berg, J.H., Van Gelder, A., and Mastbergen, D.R. (2002) The importance of breaching as a mechanism of subaqueous slope failure in fine sand. Sedimentology, Vol. 49, No. 1, pp. 81–95.

CONCLUSIONS

The laboratory experiments showed that the jetting performance of a jet trenching tool depends mainly on five mechanisms: entrainment, erosion, deposition, breaching, and overspill. These mechanisms control both the trench depth and the trench length.

© 2011 by Taylor & Francis Group, LLC

894

12 Design and risk

© 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Structural factors affecting the system capacity of jacket pile foundations J.Y. Chen & R.B. Gilbert The University of Texas at Austin

J.D. Murff Consultant

A.G. Young Geoscience Earth & Marine Services

F.J. Puskar Energo Engineering

ABSTRACT: This paper presents an analysis of various structural factors affecting the system capacity of jacket pile foundations. We use a simplified foundation model based on the upper-bound plasticity theory and a three-dimensional finite element model of the platform structure to analyze these factors. Well conductors can contribute significantly to the shear capacity of a foundation system. Conductors can also contribute to the overturning capacity when their moment capacities are large relative to the piles. The foundation capacity in shear is generally more sensitive to the strength of the steel in the piles than the strength of the soil. Jacket leg stubs penetrating below the seafloor can increase the shear capacity of the foundation system. The rigidity and strength of the jacket can limit the full mobilization of foundation system capacity. Foundation system redundancy depends on the number of piles, the geometry of the pile system, the direction of the load, and the ratio of shear force to overturning moment in the load. Collectively, these structural factors have a significant impact on foundation performance. Piles supporting the platform and the platform structure itself need to be considered as a complete system to achieve an effective and consistent level of foundation performance. 1

INTRODUCTION

Table 1.

This paper presents an analysis of various structural factors affecting the system capacity of jacket pile foundations. Jacket platforms are fixed base offshore structures used in relatively shallow water (less than 450 m) to produce oil and gas worldwide. These platforms are generally supported by open-ended steel pipe piles driven through the legs of the jacket and connected to the jacket above the sea level. Occasionally, the annuli between the jacket legs and piles are grouted to enhance their connections. These platforms are often equipped with well conductors. Geotechnical engineers often pay more attention to the axial and lateral capacities of a single pile than the overall behavior of the pile system. Furthermore, structural factors, which are as important as geotechnical factors, are sometimes overlooked in determining the capacity of the foundation system. The effects of well conductors, jacket leg stubs, yield stress for piles, structural rigidity and system redundancy on the capacity of the foundation system are investigated in this paper. This study is part of a research project undertaken to compare the predicted and observed performance of jacket pile foundations in recent hurricanes in the Gulf of Mexico (OTRC 2009 and Gilbert © 2011 by Taylor & Francis Group, LLC

Key parameters for referenced platforms.

Platform number

Water depth (m)

Number of piles

Number of conductors

Foundation plan geometry

9 10 22 30 31

19 109 34 47 30

4 3 4 6 8

1 1 None 12 None

Square Triangular Square Rectangular Rectangular

et al. 2010). Platform numbering used in this paper is kept the same as that in OTRC (2009) to facilitate reference to the report. Key parameters for the platforms referenced in this paper are presented in Table 1.

2

MODELS FOR ANALYSIS

In this study we used a simplified foundation model based on the upper-bound plasticity theory (Murff & Wesselink 1986, Murff 1987, Tang & Gilbert 1992, Murff 1999, and OTRC 2009) to analyze various structural factors affecting foundation system capacity. We

897

Figure 1. Schematic of the simplified upper-bound plasticity model for the foundation system.

also used a three-dimensional (3-D) structural model, SACS™, based on the finite element method to perform pushover analysis to estimate the capacity of the structural system including the foundation. The 3-D model was also used to infer the hurricane load on the foundation from the hindcast wave, wind and current conditions in the hurricane. Figure 1 presents a schematic of the simplified foundation model. This model assumes a plastic collapse mechanism, where all elements of resistance are characterized as rigid and perfectly plastic. This model incorporates both the structural capacities of piles and conductors and the axial and lateral capacities of soils. The piles and wells in the system collapse when one hinge forms at the head, where it is constrained by the bottom of the jacket leg for a pile or the bottom row of conductor guide framing for a well, and a second hinge forms at some depth below the head. The collapse of the entire system occurs when two hinges form in each of the piles and wells in the system due to the translation and rotation of the platform base (Fig. 1). The structure supported by the piles is assumed to be perfectly rigid and infinitely strong so that it will not fail and can distribute the load as necessary to develop the full foundation collapse mechanism. Well conductors are modeled as piles that are connected to the structure with rollers so that they can only be loaded horizontally. The solution provides an upper-bound approximation to the system capacity because it does not explicitly satisfy force and moment equilibrium. The best upper-bound solution, hereinafter designated as the solution, is the mechanism that incorporates a combination of base translation and rotation that gives the minimum system capacity. Figure 2 is an interaction diagram showing the upper-bound foundation system capacity of Platform 22 in the end-on direction. This platform has four legs, each supported by a 1067-mm (42-inch) diameter pile battered in two directions. Its foundation system capacity is represented by a base shear versus overturning moment interaction curve. Comparisons between the upper-bound solutions and results from more rigorous 3-D pushover analyses of jacket structures indicate that the upper-bound model overestimates the base shear and overturning moment causing foundation failure by approximately 10% (Murff & Wesselink 1986 and OTRC 2009). As such, the upper-bound © 2011 by Taylor & Francis Group, LLC

Figure 2. Upper-bound and expected foundation system capacities of Platform 22 in the end-on direction as compared to the hurricane load and pushover capacity according to the 3-D model.

capacity was reduced by 10% to obtain the expected foundation system capacity. The pushover failure of this platform is dominated by piles plunging and pulling out. As expected, the pushover failure load matches up very well with the expected foundation system capacity (Fig. 2). The potential foundation failure can be dominated by shear, where the lateral capacities of the piles and conductors contribute mostly to the foundation system capacity, or by overturning, where the axial capacities of the piles contribute mostly to the system capacity. Both structural and soil capacities work together to provide axial and lateral capacities for the piles and conductors in the system in these failure modes. The failure can also be dominated by a combination of shear and overturning. These distinct regions of foundation behavior are shown along the interaction curves (Fig. 2). In the shear failure mode, the base shear increases with increasing overturning moment due to pile batters. The potential failure mode of the foundation of Platform 22 is a combination of shear and overturning, which also agrees with the results from 3-D pushover analysis.

3 WELL CONDUCTORS Well conductors can contribute significantly to the shear capacity of a foundation system. The structural capacity of the outer casing for the well conductor is modeled herein to represent the minimum contribution of the conductor to foundation system capacity. Figure 3 presents the foundation system capacity interaction curves of Platform 30 showing the effect of the twelve 610-mm (24-inch) diameter conductors on the foundation system capacity. This platform has six legs, each supported by a 1219-mm (48-inch) diameter pile. The shear capacity of the foundation system (i.e. the base shear at a low overturning moment) including the conductors is about 30% higher than that excluding

898

Figure 3. Expected foundation system capacities of Platform 30 in the diagonal direction.

Figure 5. Increase in foundation system capacity of Platform 22 in the end-on direction due to increasing steel yield stress.

including the only conductor is more than 2.5 times the shear capacity excluding the conductor (Fig. 4). Furthermore, the conductor increases the overturning capacity of the foundation system by about 20% due to its high moment capacity. The maximum load in Hurricane Katrina on the foundation is higher than the foundation system capacity excluding the conductor. Failures of the foundation system could have occurred but no sign of foundation failures was found, suggesting that the conductor probably contributed to the survival of this foundation. The potential failure mode of this foundation excluding the conductor is dominated by combined shear and overturning, while it is dominated by overturning when the large conductor is included.

Figure 4. Expected foundation system capacities of Platform 9 in the end-on direction.

the conductors (Fig. 3). In contrast, the conductors increase the overturning capacity of the foundation system minimally because they are not constrained to contribute their axial capacities and can only contribute their relatively low moment capacities. The maximum load in Hurricane Katrina on the foundation is about the same as the foundation system capacity excluding the conductors. The potential failure mode of this foundation is dominated by shear. The foundation system could have shown significant distress since it was nearly loaded to its capacity. However, no sign of distress was found on the foundation during underwater inspections, suggesting that the conductors probably contributed to the shear capacity of this foundation. Figure 4 presents the foundation system capacity interaction curves of Platform 9 showing the effect of the 1829-mm (72-inch) diameter conductor on the foundation system capacity. This platform has four legs, each supported by a 762-mm (30-inch) diameter pile. The conductor is more than two times larger in diameter than the piles, which are battered in two directions.The shear capacity of the foundation system © 2011 by Taylor & Francis Group, LLC

4

STEEL YIELD STRESS

The yield stress for steel tubular members used in offshore construction is higher than its nominal value on average (Energo 2009). The higher yield stress leads to higher shear and overturning capacities of the foundation system. Its effect is generally greater on shear capacity than on overturning capacity. Figure 5 presents the foundation system capacity interaction curves of Platform 22. The shear capacity of the foundation system assuming a yield stress of 345 MPa for the piles is about 30% higher than that assuming 248 MPa (Fig. 5). The increase in shear capacity is nearly proportional to the increase in steel yield stress. In contrast, the increase in steel yield stress only increases the overturning capacity of the foundation system minimally, because the overturning capacity is dominated by the axial capacities of piles in tension and compression rather than their moment capacity. The effect of steel yield stress is more significant in shear dominated failure mode when stronger soils are present near the mudline.

899

Figure 7. Comparison of pushover failure load and foundation system capacity of Platform 31 in the broadside direction.

Figure 6. Increase in foundation system capacity of Platform 31 in the diagonal direction due to higher lateral soil resistance, increasing steel yield stress and modeling jacket leg stubs.

6 5

JACKET LEG STUBS

Jacket leg stubs penetrating below the seafloor can increase the shear capacity of the foundation system by constraining the first plastic hinges to form at the bottom of the leg stubs and pushing the second plastic hinges deeper below the mudline. These leg stubs allow the foundation system to mobilize more lateral soil resistance. Figure 6 presents a series of capacity interaction curves for Platform 31 showing the effects of pile yield stress, lateral soil resistance and jacket leg stubs on the system capacity. This platform consists of two structurally connected 4-leg jacket structures. The 914-mm (36-inch) and 1067-mm (42-inch) diameter piles supporting these jackets are battered in two directions. The base case foundation system capacity interaction curve was developed using a nominal yield stress of 248 MPa and cyclic p-y curves (Fig. 6). The maximum load in Hurricane Ike on the foundation is higher than the base case foundation system capacity and yet failures of the foundation did not occur. Rather, a few structural members of this platform above the mudline suffered damage. When piles are loaded laterally to failure under extreme loading conditions, such as those in a hurricane, they are pushed into undisturbed soils at large lateral displacements. Static p-y curves, which represent higher lateral resistance than cyclic p-y curves, are most suitable to model the lateral soil resistance (e.g. Jeanjean 2009). The intermediate capacity interaction curve developed using static p-y curves and a yield stress of 285 MPa, which is 15% higher than the nominal value, nearly surpasses the maximum load in Hurricane Ike. Furthermore, if the jacket leg stubs are modeled by constraining the first plastic hinges to form at the bottom of the leg stubs, the foundation capacity increases further (Fig. 6) and the survival of this foundation can well be explained. The effect of jacket leg stubs increases as the strength of the soil near the mudline increases. © 2011 by Taylor & Francis Group, LLC

RIGIDITY OF THE STRUCTURAL SYSTEM

The rigidity and strength of the jacket can limit the full mobilization of foundation system capacity. The simplified foundation model assumes a perfectly rigid and infinitely strong structure. Therefore, the foundation system capacity determined from this model represents the highest capacity that a structure can mobilize. Figure 7 presents the foundation system capacity interaction curve of Platform 31 and compares it with the pushover failure load in the broadside direction. Several mudline framing members were failed in the pushover analysis, resulting in a pushover capacity that is dominated by structural failures and lower than the foundation capacity. The rigidity and strength of connections between well conductors and the jacket can particularly affect the contribution of conductors to the overall capacity of the foundation system in shear (OTRC 2009).

900

7

SYSTEM REDUNDANCY

Foundation system redundancy depends strongly on the number of piles, the geometry of the pile system, the direction of the load, and the ratio of shear force to overturning moment in the load. The sensitivity of foundation system capacity to the variation in the capacity of a single pile in the system reflects qualitatively the redundancy of the system. The more sensitive the system capacity is to the capacity of a single pile, the less redundant it is. Sensitivity analyses were performed by increasing and decreasing the axial and lateral capacities of one pile by 30% for Platform 10. The ±30% variation reflects possible variations in soil properties, driving conditions and structural properties between piles; it corresponds roughly to the coefficient of variation between measured and predicted capacities of individual piles (e.g. Tang and Gilbert 1992). Platform 10 is a 3-leg jacket structure supported by a system of

Figure 8. Foundation plan of Platform 10. Figure 11. Effect of Pile C on the system capacity of Platform 10 in the hurricane loading direction.

Figure 9. Effect of Pile A on the system capacity of Platform 10 in the hurricane loading direction. Figure 12. Possible range of foundation system capacity of Platform 10 due to increasing or decreasing the axial and lateral capacities of one pile by 30% in the hurricane loading direction.

Figure 10. Effect of Pile B on the system capacity of Platform 10 in the hurricane loading direction.

three 1219-mm (48-inch) diameter piles with a 508mm (20-inch) diameter well conductor (Fig. 8). The length of Pile A is 80.8 m (265 feet) and the length of Piles B and C is 67.1 m (220 feet). Figures 9–11 present the capacity interaction curves of this foundation system due to increasing and decreasing the axial and lateral capacities of Piles A, B © 2011 by Taylor & Francis Group, LLC

901

and C, respectively. The base case foundation system capacity interaction curve is also presented in these figures to provide a reference for comparison. Pile C is the most influential pile in this foundation system as shown by the large variation in foundation system capacity (Fig. 11) and Pile A is the least influential pile (Fig. 9). These results are because Pile C is further away from the axis of rotation than Piles A and B and the direction of batter for Pile C is more effective in resisting overturning moment in the hurricane loading direction. Figure 12 presents the bounds of all capacity interaction curves shown in Figures 9–11. It shows the possible range of foundation system capacity of this platform due to ±30% variation in the capacities of each pile. The overturning capacity of the foundation system is much more sensitive to the axial capacity of a single pile than the shear capacity of the system is to the lateral capacity of a single pile as shown by the wider range of variation (Fig. 12). Similar sensitivity analyses were performed for the foundation of Platform 30 (Fig. 13). Figure 14 presents the bounds of all capacity interaction curves from the

8

CONCLUSIONS

Structural factors are significant in estimating the system capacity of jacket pile foundations. Well conductors, jacket leg stubs and steel yield stress are important to the shear capacity of the foundation system while foundation system redundancy is critical to the overturning capacity of the foundation system. The best practice in offshore platform design and assessment is for structural and geotechnical engineers to work closely together. Piles supporting the platform and the platform structure itself need to be considered as a complete system in order to achieve an effective and consistent level of performance for the foundation. ACKNOWLEDGMENTS We acknowledge Minerals Management Service and American Petroleum Institute for providing financial support through Offshore Technology Research Center for the research projects upon which this paper is based. We also acknowledge Engineering Dynamics, Inc. for providing use of their Structural Analysis Computer Software (SACS™). Britain Materek, Justin Carpenter, Sean Verret, Aditya Hariharan and Matthew Garcia contributed significantly to the research projects. The views and opinions presented herein are ours alone and do not necessarily reflect those of our sponsors.

Figure 13. Foundation plan of Platform 30.

REFERENCES

Figure 14. Possible range of foundation system capacity of Platform 30 due to increasing or decreasing the axial and lateral capacities of one pile by 30% in the diagonal direction.

sensitivity analyses. PilesA2 and C1 are the most influential piles of this foundation system in the diagonal loading direction because they are further away from the axis of rotation and their batters are more effective. The overturning capacity of the foundation system is still much more sensitive to the axial capacity of a single pile than the shear capacity of the system is to the lateral capacity of a single pile (Fig. 14). However, comparing Figures 12 and 14, the variation in the capacity of the 3-pile foundation system is greater than the variation of the 6-pile foundation system because the 3-pile system is less redundant than the 6-pile system. The 3-pile foundation system was failed in overturning in Hurricane Ike. The 6-pile foundation system survived the loading in Hurricane Katrina even though it was nearly loaded to its shear capacity. The effect of redundancy on foundation system capacity likely contributed to the actual performance of these platforms when they were loaded to their capacities. © 2011 by Taylor & Francis Group, LLC

Energo. 2009. Evaluation ofYield Stress for Steel Members in Gulf of Mexico Fixed Platforms. Final Report to American Petroleum Institute. Gilbert, R.B., Chen, J.Y., Materek, B., Puskar, F., Verret, S., Carpenter, J., Young, A. and Murff, J.D. 2010. Comparison of observed and predicted performance for jacket pile foundations in hurricanes. Offshore Technology Conference. OTC 20861. Jeanjean, P. 2009. Re-assessment of p-y curves for soft clays from centrifuge testing and finite element modeling. Offshore Technology Conference. OTC 20158. Murff, J.D. 1987. Plastic collapse of long piles under inclined loading. International Journal for Numerical and Analytical Methods in Geomechanics 11: 185–192. Murff, J.D. 1999. The mechanics of pile foundation collapse. In Jose Roesset (ed.), Analysis, Design, Construction and Testing of Deep Foundations; Proceedings of the OTRC ’99 Conference, Austin, Texas, 29–30 April 1999. Murff, J.D. & Wesselink, B.D. 1986. Collapse analysis of pile foundations. 3rd International Conference on Numerical Methods in Offshore Piling, Nantes, France: 445–459. OTRC. 2009. Analysis of Potential Conservatism in Foundation Design for Offshore Platform Assessment. Final Project Report Prepared for the Minerals Management Service. MMS Project Number 612. Tang, W.H. & Gilbert, R.B. 1992. Offshore Pile System Reliability. Final Report to American Petroleum Institute.

902

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

The new API Recommended Practice for Geotechnical Engineering: RP 2GEO P. Jeanjean BP America Inc., Houston, TX, USA

P.G. Watson Advanced Geomechanics, Perth, Australia

H.J. Kolk Fugro Engineers, B.V., The Netherlands

S. Lacasse Norwegian Geotechnical Institute, Oslo, Norway

ABSTRACT: The American Petroleum Institute (API), through the sub-committee on offshore structures SubCommittee 2 (SC2) and its Resource Group 7 (RG7) on geotechnical engineering, will soon publish a new recommended practice (RP) on geotechnical engineering, entitled RP 2GEO. This new RP covers both shallow foundations and pile foundations and is aligned to the greatest extent possible with the current ISO 19901-4 (for shallow foundations) and ISO 19902 (for pile foundations for fixed steel structures). This paper describes the key features of the new code and how its compares both with existing ISO guidance and with previous guidance in API RP 2A WSD 21st Ed. The philosophy behind the technical changes in RP 2GEO will also be explained. The authors of the papers are all members of RG7 and its ISO equivalent committee, Working Group 10 (TC7/SC7/WG10). 1

INTRODUCTION

Recommendations on geotechnical practice in API codes pertaining to offshore structures are prepared and written by the geotechnical Resource Group 7 (RG7) (formally known as RG5). In the International Standard Organization (ISO), the group equivalent to RG7 is Work Group 10 (WG10) on Foundations, formally referred to as TC67/SC7/WG10. Recognizing the need for seamless integration of geotechnical practice around the world, RG7 and WG10 have functioned as a joint committee preparing and jointly approving API and ISO foundation guidance. RP 2GEO is aligned with ISO 19901-4 (2003) and ISO 19902 (2007). At the time of writing this paper, the recommended practice “RP 2GEO (1st Edition)” is pending approval for issue by API. Its title will be “ANSI/API Recommended Practice 2GEO Geotechnical and Foundation Design Considerations”. This paper describes the key features of RP 2GEO and how its compares with existing guidance in ISO 19901-4 (2003) for shallow foundations and ISO 19902 (2007) for pile foundations, and with previous guidance in API RP 2A 21st Edition. The philosophy behind the technical changes in this new RP will also be explained. The “lumped” safety factor approach in use in API RP 2A WSD 21st Ed was retained in RP 2GEO. Key © 2011 by Taylor & Francis Group, LLC

changes were implemented in the shallow foundation section including adopting the Brinch Hansen method instead of the Vesic approach currently used in API RP 2A. In the deep foundation section, recommendations on the use of cone penetration test based methods for pile design in sands are now provided, while the soil structure interaction curves, so called t-z curves, for sands have also been modified. Because ISO 19901-4 uses a load and material factor approach, and RP 2GEO a lumped safety factor, the two codes do not compare exactly. Where possible, care was however taken so that the two codes provide foundations of comparable sizes for given load and soil profile. Because of space limitations, this paper presents an abbreviated discussion and a more detailed description of RP 2GEO can be found in Jeanjean et al. (2010). 2

DESIGN OF SHALLOW FOUNDATIONS

2.1 API 21st edition (2000) The framework proposed by Vesic (1975) was included in the RP 2A 10th Edition (1979) and was unchanged through the 21st edition (2000). In Vesic (1975), the bearing capacity factors Nc and Nq were the same as those recommended by Meyerhof (1951). The Nγ factor was approximated with

903

an empirical relationship (Nγ = 2(Nq + 1) tanφ
) that closely matched the results of Caquot and Kerisel (1953).

friction angle, the ultimate load envelopes are significantly smaller when using RP 2GEO compared to ISO 19901-4. No effort was made to harmonize these envelopes. • However, due to differences in the way safety factors in RP 2GEO and load/material factors in ISO 19901-4 are applied, envelopes of allowable loads are similar for both codes.

2.2 The ISO 19901-4 (2003) approach The Brinch Hansen (1970) method included bearing capacity factors Nc and Nq that were the same as those recommended by Meyerhof (1951) and Vesic (1975). The Nγ factor was approximated by an empirical relationship Nγ = 1.5 (Nq − 1) tanφ
. Values of φ
used for the derivation of Nc and Nq are based on plane strain

conditions, φPS (rather than triaxial conditions, φTX ),
and it was suggested that φPS may be (conservatively)
. taken as 10% higher than φTX The Brinch Hansen (1970) approach and the load and material factor framework was adopted for the ISO 19901-4 (2003) standard on shallow foundations, with modified depth and shape factors, as summarized in Table 1.

Overall, the calculations performed with RP 2GEO and ISO 19901-4 for foundations on soils behaving as drained will result in comparable foundation size to resist a given set of allowance (unfactored) load. Going forward, the greatest challenge to align the RP 2GEO and ISO 19901-4 texts lies in how to reconcile the lumped safety factor, working stress design (WSD) API approach with the partial load factor and material factor design ISO approach. 3

CALCULATION OF PILE CAPACITY

2.3 The RP 2GEO approach

3.1

The RG7/WG10 group decided that the differences between ISO 19901-4 (2003) and API RP 2A 21st Edition (2000), while not entirely reconcilable, would be minimized in RP 2GEO. The key differences between the adopted recommendations of RP 2GEO and those of RP 2A 21st Edition, and ISO 19901-4 (2003) are summarized in Table 1.

Design practices for the axial capacity of piles in sand have been and still are mostly aligned around the world, except for the choice of the value for K, the coefficient of lateral earth pressure. Although a K value of 0.8 for compression and tension is used in the Gulf of Mexico, K values of 0.7 in compression and 0.5 in tension are still often applied for North Sea projects and is recommended in DNV (1992) and Lloyd’s Register (1989, 2008). These values can be upgraded for North Sea projects for dense sands, if high quality CPT data is available (Jeanjean et al., 2010).

2.4 Comparison of RP 2GEO and ISO 19901-4 for jacket mudmats In the example of Fig. 1, failure envelopes are developed in V-H space. The curves depict the combinations of allowable (i.e. unfactored) vertical loads, V, and horizontal loads, H, that satisfy the requirements of the RP 2A 21st Ed., RP 2GEO, and ISO practices. Failure envelopes are developed for a surface (no skirts) 10 m square mudmat founded on soil with constant su = 10 kPa. The material factor was taken as 1.25 in the ISO 19901-4 (2003) calculations. Note that the two allowable envelopes generated using ISO 19901-4 (2003) represent different (assumed) combinations of dead and environmental load: a. All vertical and lateral loads are assumed to be environmental and the load factor is 1.35. b. 50% of the vertical load is assumed to be dead load, with the remainder (and all lateral load) assumed to be environmental. The load factor is therefore taken as 0.5(1.1 + 1.35) = 1.225 for the vertical load and as 1.35 for the horizontal load. It can be seen that comparable envelopes of ultimate capacity are generated using ISO 19901-4 (2003) and RP 2GEO. However, differences in factoring between WSD and load/material factor design approaches leads to differences in allowable load. In comparing drained envelopes from ISO 19901-4 (2003) and RP 2GEO (Fig. 2), it is typically seen that: • Because RP 2GEO uses a smaller triaxial friction

angle and ISO 19901-4 uses a larger plane strain © 2011 by Taylor & Francis Group, LLC

3.2

Gulf of Mexico and North Sea practices

Key studies on pile capacity in sands 1993–2008

Key studies between 1993 and 2008 that influenced the recommendations in RP 2GEO include the following (a summary of the findings of each study can be found in Jeanjean et al., 2010): Offshore Technology Research Center (OTRC) – Fugro McClelland (1990–1994) study; Hossain and Briaud (1992) and (1993); Fugro McClelland (1994); The EURIPIDES pile load tests (1995); The MTD Method: Jardine and Chow (1996) and the ICP method: Jardine et al. (2005); The Ras Tanajib II pile load tests (1996–1999); APIsponsored Fugro Engineers Pile Study (2003–2004); 2004 seminar on “Piles in Sand” – Houston TX; Studies at the Norwegian Geotechnical Institute (NGI): 1995 to 2005; The University of Western Australia (UWA) study - Lehane et al. (2005). Detailed references to these studies can be found in Jeanjean et al (2010). 3.3 The RP 2GEO approach for pile design The modifications to the RP 2A 21st Edition (2000), as implemented in RP 2GEO, are now summarized. These modifications were initially published as part of the RP 2A Addendum and Errata #3 (2007) in October 2007. Unfortunately, several key typographical

904

Table 1.

Key differences between RP 2A 21st Ed., RP 2GEO, and ISO 19901-4 for shallow foundations.

Topic Framework Factors

RP 2A 21st Ed. (2000)

RP 2GEO

ISO 19901-4

Lumped safety factors, independent of type of loads (i.e. gravity vs environmental)

Lumped safety factors, independent of types of load (i.e. gravity vs environmental)

Load and material factors, with load factors varying for gravity loads and environmental loads.

Not included

Included with unfactored soil effective unit weight Horizontal load is transferred to foundation base to calculate inclination factors and safety factor against sliding. Vertical load is transferred to foundation base.

Undrained bearing capacity Overburden term in Included with soil bearing capacity total unit weight equation Horizontal load No recommendations transfer to foundation base Vertical load transfer to foundation base

No recommendations

Bearing capacity factors Inclination factors

Nc = Meyerhof (1951) = Prandtl (1920) = 5.14 Vesic (1975)

Shape factors

Vesic (1975)

Depth factors Base inclination and seafloor slope factors Linearly increasing shear strength profiles

Vesic (1975) Vesic (1975)

Horizontal load is transferred to foundation base to calculate inclination factors and safety factor against sliding. Vertical load is transferred to foundation base only to calculate load eccentricity and effective area (if applicable), not to calculate bearing safety factors. Nc = Meyerhof (1951) = Prandtl (1920) = 5.14 Brinch Hansen (1970) with unfactored horizontal load (where sliding outside the effective area used to reduce inclination factor, this is to be factored) DNV (1974) and Salençon & Matar (1983) with the shape factor for circular foundations and pure vertical loads changed from 0.2 to 0.18 for uniform soil conditions DNV (1992) = DNV (1977) Brinch Hansen (1970)

No recommendations

Davis and Booker (1973)

Davis and Booker (1973)

Included with soil total unit weight

Not included

No recommendations

Horizontal load is transferred to foundation base to calculate inclination factors. Not included Nq = Meyerhof (1951) = Reissner (1924) Nγ = Brinch Hansen (1970)

Included with unfactored *soil effective unit weight Horizontal load is transferred to foundation base to calculate inclination factors. Vertical load is transferred to foundation base. Nq = Meyerhof (1951) = Reissner (1924) Nγ = Brinch Hansen (1970)

Brinch Hansen (1970) with unfactored loads Brinch Hansen (1970) DNV (1977)

Brinch Hansen (1970) with factored loads Brinch Hansen (1970) DNV (1977)

Brinch Hansen (1970)

No recommendations

Not included in bearing capacity equation Triaxial friction angle

Included in bearing capacity equation Plane strain friction angle (10% greater than triaxial)

Drained bearing capacity Overburden term in bearing capacity equation Horizontal load transfer to foundation base Vertical load transfer to foundation base Bearing capacity factors

Inclination factors Shape factors Depth factors Base inclination and seafloor slope factors Effective cohesion c’ Choice of drained friction angle for sand

No recommendations Nq = Meyerhof (1951) = Reissner (1924) Nγ = Vesic (1975) (approximated from Caquot and Kerisel (1953)) Vesic (1975) Vesic (1975) Vesic (1975) = Brinch Hansen (1970) Vesic (1975) Included in bearing capacity equation No recommendations

© 2011 by Taylor & Francis Group, LLC

905

Nc = Meyerhof (1951) = 4 Prandtl (1920) = 5.1 Brinch Hansen (1970) with factored loads

DNV (1974) and Salençon & Matar (1983)

DNV (1992) = DNV (1977) No recommendations

Figure 3. Ratio of Calculated-to-Measured pile compression capacity, according to the API main text method, as a function of sand relative density.

Figure 1. V-H envelope for 10 m square mudmats founded on soil with su = 10 kPa (0.21 ksf).

Figure 4. Ratio of Calculated-to-Measured pile compression capacity, according to the API main text method, as a function of pile Length-to-Diameter ratio.

Figure 2. V-H envelope for 10 m square mudmats founded on soil with triaxial φ
= 32◦ .

mistakes occurred in the Commentary and this document is not to be used in design. The same modifications are also included in ISO 19902 (2007), but without typographical errors. The bias of the API (1993) method, as documented by some of the above studies, is illustrated in Figures 3 to 6, as a function of the relative density of sand and the pile-length diameter ratio, both for compression and tension loading. A key question is how the RP 2A calculated pile capacities in sand should be viewed. Are they (1) an accurate and unbiased estimate, (2) a conservative © 2011 by Taylor & Francis Group, LLC

906

Figure 5. Ratio of Calculated-to-Measured pile tension capacity, according to the API main text method, as a function of sand relative density.

estimate or (3) an estimate for prudent design based on the (best) available experience? Murff (2005) stated that the API method is a design method and not a prediction method. The track record of the method has been good. However, there is probably only a fraction of the API-designed foundations that have been subjected to their design loads. It is the opinion of the four authors that the pile capacities calculated with the API approach provide

Figure 6. Ratio of Calculated-to-Measured pile tension capacity, according to the API main text method, as a function of pile Length-to-Diameter ratio. Figure 7. Measured cone profile in West Delta area.

good, appropriately conservative estimates for the design of piles for offshore installation in medium to dense sands. Its application is simple, therefore unambiguous, and should minimize user errors. 3.4 Modifications to the main text method The main text method was modified as follows: • The symbols K and δ were removed and replaced

by the product “β = K·tan (δ)” in the pile design method to reduce confusion and lower the risk of errors (see Jeanjean et al., 2010, for more details). • To remove unconservatism, the API main text method is no longer recommended for very loose and loose sands, loose sand-silts, and medium dense silts, and dense silts. The CPT-based methods in the Annex are recommended instead. The designer should be aware that areas of potential unconservatism still remain in the RP 2GEO main text method. In comparison to the CPT-based methods, the method may not be conservative in tension for thick deposits of medium dense sands (Fig. 5). The description of where the main text method is conservative (short piles less than 45 m (150 ft) in compression in dense to very send sands) and where the method may be unconservative (all other cases) has been changed from previous RP 2A text. 3.5 Addition of CPT-based methods in the Annex The RG7/WG10 committee decided after long deliberations to include four CPT-based methods in the Commentary/Annex section of RP2GEO. No agreement was reached on which method was the preferable one to include in the main text. The four methods in the annex are: the Simplified-ICP, Fugro-05, NGI-05, and Offshore UWA-05 methods. All methods arediscussed in Lehane et al. (2005). 3.6 Example of pile capacity calculations. To illustrate the use of the CPT methods, the capacity of a driven pile is calculated for a site in the Gulf of Mexico West Delta area (Fig. 7). Pile capacities © 2011 by Taylor & Francis Group, LLC

907

Figure 8. Comparison of friction capacity in tension, as a function of pile penetration depth.

were calculated by using the API main text and the CPT methods with the cone resistance profile shown in red for sand layers. For clays layers, the API main text method was used with the shear strength profile shown in green. The capacity curves, calculated for a 1.07 m (42in) pile with 38.1 mm (1.5 in.) wall in Fig. 8 and 9 reflect the biases of Fig. 3 to 6 and illustrate the dilemma of RG7/WG10 noted in section 4.5. The example herein is only one example of calculations with the four CPT and API main text methods. They are for a slender pile in a highly layered profile. Other soil profiles can give other relative pile capacity profiles for each of the different analysis methods. For clarity of comparison, the calculated end bearing is not averaged over a few pile diameters to account for soil layering. This averaging of the end bearing when the pile tip approaches a sand-clay layer interface should be included in an actual design.

Brinch-Hansen, J. 1970: “A Revised and Extended Formula for Bearing Capacity”, Geoteknisk Institut, Bulletin No. 28, p. 5–11, Copenhagen. Caquot A., and Kerisel, J. 1953: “ Sur le Terme de Surface dans le Calcul des Fondations en Milieu Pulvérulent” Proc. of the 3rd Conference on Soil Mechanics and Foundation Engineering, Vol. 1, Zürich. Davis, E.H. and Booker, J.R. 1973: “The Effect of Increasing Strength with Depth on the Bearing Capacity of Clays”, Géotechnique, 23 (4), 551–563. DNV 1974: “Rules for the Design, Construction and Inspection of Fixed Offshore Structures”, DetNorskeVeritas, Oslo. DNV 1977: “Rules for the Design, Construction and Inspection of Offshore Structures – Appendix F – Foundations”, Det Norske Veritas, Oslo. DNV 1992: “Foundations”, Classification notes No. 30.4, Det Norske Veritas, February Fugro Engineers 2004: “Axial Pile Capacity Design Method for Offshore Driven Piles In Sand”, Fugro Report No.: P1003 by A. Baaijens and H.J. Kolk to the American Petroleum Institute, August 5. Fugro McClelland 1994: “Axial Capacity of Piles in Sand”, Report No. 0201-1485 by T. Hamilton and J.S. Templeton Figure 9. Comparison of total capacity in compression, as to the American Petroleum Institute, August. a function of pile penetration depth. Hossain, M. K., and Briaud, J.L. 1993: “Improved Soil Characterization for Pipe Piles in Sand in API RP2A”,Proceedings, Offshore Technology Conference, 4 CONCLUSION Houston,TX,Paper 7193. ISO 19901-4 2003: Petroleum and natural gas industries — The API and ISO committees working to develop Specific requirements for offshore structures — Part 4: guidelines for offshore foundations recently introGeotechnical and foundation design considerations, 1st duced RP 2GEO. Except for its working stress design Edition ISO 19902 2007: Petroleum and natural gas industries — (WSD) framework, RP 2GEO incorporates pile design Fixed Steel Offshore Structures, 1st Edition recommendations similar to those of ISO 19902 and Jardine, R., Chow, F. 1996: “New Design Methods for Offrecommendations on shallow foundations that are shore Piles”, MTD Publication 96/103 by The Marine largely aligned with those of ISO 19901-4, thereby Technology Directorate Ltd, constituting a great step in harmonizing offshore Jeanjean, P., Watson, P.G., Kolk, H., and Lacasse, S. 2010: geotechnical practices around the world. “RP 2GEO: The new API Recommended Practice for Geotechnical Engineering.”, Proceedings, Offshore Technology Conference, Houston, TX, Paper 20631 Lehane, B., Schneider, J.A., and Xu, X. 2005: “A Review ACKNOWLEDGEMENT of Design Methods of Offshore Driven Piles in Siliceous Sands”, University of Western Australia, Report GEO The members of theAPI RG7 and ISOTC67/SC7/WG10 05358, September committees are acknowledged for their many passionLloyd’s Register 1989: “Rules and Regulations for the Classiate discussions on geotechnical matters over the years fication of Fixed Offshore Installations’, Part 3, Chapter 2. and their contributions to the writing and editing of RP Lloyd’s Register 2008: “Rules and Regulations for the Clas2GEO. sification of a floating offshore installation at a fixed location”, Part 3, Chapter 12., April The authors are grateful to their respective compaMeyerhof, G.G. 1951: “The Ultimate Bearing Capacity of nies for the permission to publish. Foundations”, Géotechnique, Vol. 2, 301–332 Murff, J.D. 2005: Personal communication to P. Jeanjean and the RG7/WG10 resource groups, June. REFERENCES Salençon, J., and Matar, M. 1983: “Bearing Capacity of Circular Shallow Foundations in Foundation Engineering, ANSI/API Recommended Practice 2GEO “Geotechnical and Soil Properties”, Foundation Design and Construction, Foundation Design Considerations”, 1st Edition, (under Vol. 1, pp.159–168, Presses de l’Ecole Nationale des Ponts development). et Chausées. API RP 2A 1st Edition (1969) to 21st Edition (2000): “RecVesic, A.S. 1975: “Chapter3 – Bearing Capacity of Shalommended Practice for Planning, Designing and Conlow Foundations”, Foundation Engineering Handbook, structing Fixed Offshore Platforms – Working Stresses Edited by Winterkorn, H.F. and Fang, .H.Y., Von Nostrand Design”. Reinhold Company, NewYork, pp.121–147. API RP 2A-WSD Errata 2007: “Errata and Supplement 3 Wisch, D., and Mangiavacchi, A. 2010: “Strategy and Structo API Recommended Practice 2A-WSD, Recommended ture of the API 2 Series Standards, 2010 and Beyond”, Practice for Planning, Designing and Constructing Fixed Proceedings, Offshore Technology Conference, Houston, Offshore Platforms – Working Stresses Design”, 21st TX, Paper 30831. Edition, December 2000.”, October 2007. © 2011 by Taylor & Francis Group, LLC

908

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Comparison of ISO 19901-2 and API RP 2A seismic design criteria for a site in the Caspian Sea, Turkmenistan Z.A. Lubkowski Arup, London, UK

J.E. Alarcon AIR, London, UK

Z.A. Razak Petronas, Kuala Lumpur, Malaysia

ABSTRACT: Due to unavailability of zoning data, as per clause 2.3.3c of the API RP-2A WSD 22nd Edition, a probabilistic seismic hazard assessment was carried out for the proposed offshore facilities by the Operator of Block 1, Turkmenistan, Caspian Sea. The purpose of the study was to derive strength level event (SLE) and ductility level event (DLE) seismic design criteria for the facilities. These seismic design criteria are compared to those derived using the potentially more rigorous approach presented in ISO 19901-2, where the concepts of “consequence of failure” and “exposure level” are directly accounted for. ISO 19901-2 allows the designer to consider the risks and uncertainties in making decisions to come up with the abnormal level earthquake (ALE) and the extreme level earthquake (ELE) ground motion values for design. This paper presents a comparison of these two seismic design criteria and also examines the influence of different seismic hazard models on the resulting design criteria.

1

BACKGROUND

The site, known as Block 1, is located in the Turkmenistan sector of the Caspian Sea, approximately 70 km offshore in water depth ranging from 30 m to 90 m. The design code selected for the proposed gravity based platform is API RP-2A (2000). However, since API does not provide seismic criteria outside of the USA the operator of Block 1 embarked on a study to determine values for the design of their facilities. The principal findings from that study were peak ground acceleration (PGA) values of 0.2 g and 0.4 g for return periods of 200 years SLE and 2000 years DLE criteria respectively. Subsequently a more detailed probabilistic seismic hazard assessment (PSHA) was carried out (Fugro, 2006) based on the latest available data at that time. This more detailed study found that the PGA values to be considerably higher, ranging from 0.34 g to 0.73 g for the 200 years SLE and 2000 years DLE respectively. This magnitude of increase in the seismic criteria had significant implications for the seismic design of the platform. The operator therefore decided to implement performance based seismic analysis methods together with revised seismic criteria based on the newly released ISO 19901-2 (2004) code, where the concept of “consequence of failure” and “exposure level” together with platform type and function are © 2011 by Taylor & Francis Group, LLC

directly accounted for in defining appropriate seismic design criteria. This paper presents a comparison of these two seismic design criteria for this project and also examines the influence of different seismic hazard models on the resulting design criteria.

2

OFFSHORE DESIGN GUIDELINES

API RP-2A (2000) defines two levels of earthquake ground motion for design. Firstly, those which have a reasonable likelihood of not being exceeded at the site during the platform life span, and secondly those ground motions from a rare intense earthquake (RIE). The first level of ground motions is named the strength level earthquake (SLE) while the latter is named the ductility level earthquake (DLE). API requires, from the structural and geotechnical design, that the offshore platform should have no significant damage for the SLE and that collapse is to be prevented during the DLE. API also presents exposure categories for selection of design levels; these exposure categories depend on a life-safety level (i.e. is the structure manned and non-evacuated, manned and evacuated or unmanned) and on a consequences of failure level (i.e., high, medium or low).

909

However, API does not specify the actual level of ground motion to be used and thus, the selection of return periods for the SLE and the DLE are left at the owner’s descretion. Furthermore, there is a potential inconsistency in the use of the API guidelines, which is related to the widespread selection of the DLE ground motions as twice of the SLE level (i.e., DLE = 2.0 × SLE). API mentions in Section 2.3.6.d that “No ductility analysis of conventional jacket-type structures with 8 or more legs is required if the structure is to be located in an area where the intensity ratio of [DLE] to [SLE] is 2 or less…and the following conditions are adhered to in configuring the structure and proportioning members:…” after which the “DLE = 2.0 × SLE” rule has developed. To the author’s best knowledge, API does not state this as a general requirement for defining the DLE ground motion. ISO 19901-2 (2004) also proposes two levels of earthquake ground motion for design. The extreme level earthquake (ELE) is that at which the structure should be able to sustain no major damage while the abnormal level earthquake (ALE) is that at which the structure should not suffer complete loss of integrity. Three levels of exposure L1, L2 and L3 (where L1 represents facilities that require higher standards of design) are considered in the definition of ground motion design levels (see Section 6.4 in ISO 19901-2 for further details). The potential advantage of ISO over API lies in the fact that the recommended return periods for design are clearly determined based on the seismicity characteristics of the region surrounding each particular site. The ALE and ELE return periods are defined by following a step-by-step procedure that makes use of the part of the hazard curve corresponding to a certain response period. 3

SEISMIC HAZARD MODEL

The hazard curves produced from a PSHA are the primary inputs for defining the design levels in both the API RP-2A and ISO 19901-2 approaches. At the site, two independent PSHAs were carried out, the first one was by Fugro in 2006 and the second by Arup in 2007. Both studies implemented state-of-practice methods and thus the results can be considered to be mutually complementary. PSHAs combine seismic source zoning, earthquake recurrence and ground-motion predictive (attenuation) equations to produce hazard curves that represent the level of ground motion, and an associated annual frequency of being exceeded. These key elements are explained briefly in the subsequent sections, and include the description on how these elements were implemented in the Fugro and Arup models. 3.1 Tectonic setting The Caspian Sea region is located in the northern most collision boundary between the Arabian and Eurasian © 2011 by Taylor & Francis Group, LLC

plates. The predominant northward movement of the Arabian plate creates a compressive environment in the Caucasus, which induces predominantly westward movement of the Anatolian (Turkish) block and northeastward movement of the Iranian block. The detailed tectonic setting in the project area is complex and describing its details lies beyond the scope of this paper. However, it is important to highlight that the denser oceanic crust is being subducted beneath the lighter central Caspian Sea continental crust (Knapp et al., 2004; Jackson et al., 2002). The deformation along this complex plate margin is believed to be partitioned onto both strike-slip faults and fold and thrust structures including the Apsheron sill or Apsheron anticline. This is a linear uplifted structure of faults and folds, the axis which extends from Baku in Azerbaijan to Cheleken in Turkmenistan.

3.2

Seismic source models

The geographical variation of earthquake activity is represented in a seismic source model. They are based on the distribution of observed seismic activity together with geological and tectonic understanding of the region. The zones represent areas where the seismicity is assumed to be homogenous; i.e. an equal chance that a given earthquake will occur at any point. Fugro and Arup used a similar database of historic and instrumental recordings in their respective analyses. Replicated events were removed from the compiled database considering: the quality of information (some were given a higher confidence level due to better documentation), the completeness of information and the magnitude scale used (a uniform magnitude scale is required for the composite catalogue, so preference was given to entries that minimise adjustments). Foreshocks and aftershocks were removed from the catalogue following the Gardner and Knopoff (1974) methodology. Figure 1 shows the idealised seismic source model developed by Arup (2007). A similar model was developed by Fugro (2006). The seismic source models included shallow crustal sources (10 in the Fugro model, 20 in the Arup case) and subduction zone sources of the South Caspian Sea. The model also includes the Apsheron sill or Apsheron anticline, modelled as a 72 km long planar source located about 10 km from the site. Differences in the models arise largely from the range of expert opinions embodied in the PSHA source zonation process. However, a detailed discussion is beyond the scope of this paper.

3.3

Recurrence relationships

Recurrence relationships in the form of the “Gutenberg-Richter” equation (Log10 N = a − bM) were assigned to each of the Arup and Fugro seismic source zones. Table 1 presents the values for the Arup model, as a detailed comparison and discussion is beyond the scope of this paper.

910

Table 1. Mean magnitude-recurrence parameters and maximum magnitude per zone for the Arup seismic source model.

Figure 1. Arup seismic source models, showing shallow crustal (upper) and deep source zones (lower).

Source zone ID

b

Activity rate, a (N > 4.0/ year)

Maximum magnitude Mw

AZB1 AZB2 AZB3 AZB4S AZB5S AZB6S AZB7S AZB8 AZB9 AZB10 AZB11 AZB12S AZB13S AZB14 AZB15 AZB16 AZB17 AZB18 AZB19 AZB20 AZB4D AZB5D AZB6D AZB7D AZB12D AZB13D

0.74 1.40 1.00 0.83 1.84 1.30 1.00 0.93 1.40 1.20 0.75 0.83 0.83 0.75 0.75 1.20 1.00 0.83 1.00 1.30 0.75 0.93 0.93 1.20 0.93 0.83

0.095 4.837 0.083 1.196 7.652 3.755 0.486 0.458 1.326 0.231 0.349 3.783 0.543 0.45 0.528 0.19 1.852 1.445 1.717 2.695 0.691 4.183 1.598 0.86 2.246 0.816

7.5 7.0 8.0 8.0 7.2 7.2 8.0 8.3 7.0 6.5 8.0 8.0 8.0 8.0 8.0 6.5 8.0 8.0 8.0 7.2 8.0 8.3 8.3 6.5 8.3 8.0

Table 2. PSHAs.

Ground-motion predictive equations used in the

3.4 Ground motion predictive equations

Tectonic regime Fugro

A set of ground-motion predictive (attenuation) equations (GMPEs) is required to define what ground motion could be expected at location A due to an earthquake of known magnitude occurring at location B. These equations are generally derived from observations from past earthquakes, and play a significant role in a PSHA. The GMPEs used in the Fugro and Arup studies are summarised in Table 2 The large number of shallow crustal zones with respect to the subduction ones in the Arup model is the fundamental reason for the selection of two equations applicable to this kind of region. Epistemic uncertainties were considered in both studies using a logic tree approach.The logic trees were applied on parameters such as the maximum magnitude, recurrence parameters and GMPEs. The different branches of the logic trees and their relative weightings (i.e., the values given that represent the degree of confidence in each branch) are not presented herein.

Shallow Crustal Abrahamson & Silva (1997)

4

SEISMIC HAZARD RESULTS

The basic methodology adopted in the PSHA is based on that originally proposed by Cornell (1968) from © 2011 by Taylor & Francis Group, LLC

Subduction

Arup

Ambraseys et al. (1996) Sadigh et al. (1997) Youngs et al. (1997) Youngs et al. (1997) Crouse (1991)

the work of Cornell (1964) and Esteva (1967). The analyses presented here calculate the peak ground acceleration (PGA) and the uniform hazard response spectra (UHRS) for given return periods. For the purpose of comparing ISO and API methodologies using two different, but very robust seismic models, the Fugro model was re-run using the Arup program SISMIC. A comparison of the UHRS from the Fugro and Arup models is presented in Figure 2 for return periods of 200 and 2000 years. The results presented in Figure 2 show a good agreement for the response spectra with a return period of 200 years, whilst significant differences for the response spectra with a return period of 2000 years. The observed differences in the Fugro and Arup models do not come as a surprise since two different seismic hazard models are not expected to

911

Figure 2. Comparison of UHRS for 200 and 2000 years return period from the Fugro and Arup models. Table 3. Comparison of Arup and Fugro PGA values for different return periods. The ratios between Arup and Fugro values are given in the last row. Return period (years)

SLE in (g)

DLE in (g)

100

150

200

1000

1500

2000

Arup Fugro Ratio A/F

0.18 0.30 0.60

0.22 0.34 0.65

0.27 0.38 0.71

0.65 0.66 0.98

0.79 0.75 1.05

0.88 0.81 1.09

From the data presented in Table 3 it is observed that the SLE ground motion accelerations are highly dependent on the seismic model employed, with the DLE ground motions showing slightly more stable results. Since the SLE requires the structure to sustain a low level of damage, a difference of around 30% in the structural demand may impose significant construction costs and large variability in the safety levels evaluated. If the widely implemented practice of scaling DLE from SLE, as previously described, is applied to the results presented in Table 3 for the 200 year return period case, the DLE ground motions for Arup and Fugro models would be 0.54 g and 0.76 g respectively. These ground motions equate to return periods of about 630 years and 1600 years for Arup and Fugro models respectively, which are lower than the 2000 year specified for the facilities design. These differences not only point out the poor practice of scaling DLE from SLE, but also highlights the dependence of PSHA results on the seismic model used. 5.2

ISO defines the ELE and ALE return periods. The key input parameters are the slope of the seismic hazard curve (aR ) which best represents the principle vibration mode of the platform being assessed and the probability of failure (Pf ). However, since the vibration mode may not be available at the time when the selection of ALE and ELE levels is carried out, ISO recommends the use of the 1.0 second response-period hazard curve as a default. Even though no specific reason is given for the selection of 1.0 second period for analysis, this may be related to the fact that most offshore structures have a mode of vibration greater than 1.0 second, and 1.0 second is usually corresponds to to the constant velocity portion of the response spectrum. This seems reasonable since peak ground velocity (PGV) has been shown to be better correlated with damage than PGA (Bommer & Alarcon, 2006). Table 4 shows the resulting ELE and ALE return periods, and their corresponding ground motions for the Arup and Fugro models. From these results it is observed that the influence of the seismic hazard model on the PGA values is not as significant as in the case of the API procedure. This effect seems to be particularly true for the lower return periods (i.e., ELE vs SLE levels) in which the ISO results show a minimum ratio of 0.91 whilst the API results show a minimum ratio of 0.42. For the longer return periods (ALE vs DLE levels), there does not seem to be a significant difference in the ISO and API results.

produce identical results. However, the significant differences observed at the 2000 years return period may be unusual. These differences come from the source zoning, earthquake data processing and predictive equations selected in each model. Considering that each model has gone through a comprehensive review, the observed differences highlight the large uncertainties enveloped in any PSHA. 5 API AND ISO DESIGN GUIDELINES TheAPI RP-2A and ISO 19901-2 design guidelines are implemented using the seismic hazard results from the Arup and Fugro models. These results and sensitivity analyses carried out are presented herein.

5.1 API RP-2A As introduced above API defines the SLE and DLE ground motion levels for design. API does not define concrete return periods, but the values of 200 and 2000 years have become accepted by industry. Table 3 presents the PGA values calculated from the Arup and Fugro models for these two levels. Additionally, and for sensitivity analysis purposes, other return periods have also been included for the following return periods: 100 and 150 years for the SLE and 1000 and 1500 years for the DLE. These SLE and DLE values have been selected arbitrarily, but 50 years was selected for compatibility with the minimum value recommended in ISO 19901-2. © 2011 by Taylor & Francis Group, LLC

ISO 19901-02

Sensitivity assessment considering the response period hazard curve As previously introduced, ISO recommends the use of the seismic hazard curve at 1.0 seconds response period, if project specific data are not available. Since the offshore facility considered has a natural period of vibration of about 2.5 seconds, a sensitivity assessment

912

5.2.1

Table 4. Comparison of Fugro and Arup return periods and PGA values (in g) obtained when applying ISO 19901-2. The ratio on PGA between the Arup and Fugro models is given in the last row. Exposure & L1 Design Levels ELE Arup Fugro Rat. A/F

475 0.45 g 350 0.46 g 0.98

L2

L3

ALE

ELE

ALE

ELE

ALE

3125 1.04 g 3620 0.94 g 1.11

270 0.32 g 157 0.35 g 0.91

1175 0.70 g 1260 0.71 g 0.99

125 0.20 g 60 0.22 g 0.91

450 0.44 g 475 0.51 g 0.86

Table 5. Fugro ELE and ALE return periods (in years) calculated from hazard curves at different response periods. Exposure & Design Levels

L1

L2

L3

ELE

ALE

ELE

ALE

ELE

ALE

1.0 s 2.0 s 3.0 s

350 470 560

3620 3050 3000

157 212 265

1260 1180 1150

60 80 101

475 470 465

Figure 3. Derivation of the slope aR of the seismic hazard curve for a given response period. Table 7. Fugro ELE and ALE return periods (in years) calculated from different aR values.

Table 6. Arup ELE and ALE return periods (in years) calculated from hazard curves at different response periods. Exposure & Design Levels

L1

L2

L3

ELE

ALE

ELE

ALE

ELE

ALE

1.0 s 2.0 s

475 490

3125 3220

270 280

1175 1125

125 135

450 450

on the influence of the response period hazard curve used in the ALE and ELE evaluation is carried out. For this comparison hazard curves for 2.0 and 3.0 second response periods are used to evaluate the variability of the results. Table 5 and Table 6 present the results of this analysis. The results from the Fugro model show a degree of dependence of the ALE and ELE levels on the response period employed in the calculation, whilst the results from the Arup model do not seem to be significantly affected by it. The results presented in Tables 4 and 5 require further exploration before reaching a more definitive conclusion, since basic factors such as the GMPEs employed in each model may have had a significant influence on the results. 5.2.2 Sensitivity assessment considering the slope aR of the hazard curve A sensitivity assessment on the methodology used to calculate the hazard slope aR was carried out.The slope aR is defined as the ratio of the spectral accelerations corresponding to two probability values, one at either side of Pf , where Pf is the “Target annual probability of failure” as given in ISO 19901-2. ISO recommends © 2011 by Taylor & Francis Group, LLC

Exposure & Design Levels

L1

L2

L3

ELE

ALE

ELE

ALE

ELE

ALE

50 – 50 40 – 60 30 – 70

350 350 340

3620 3620 3600

157 155 155

1260 1250 1250

60 60 60

475 475 475

that the two probability values used to calculate aR (P1 and P2 ) should be one order of magnitude apart, with P1 being preferably close to Pf , as shown in Figure 3. The ALE and ELE variability on the selected distribution of the P1 and P2 values was examined. Three distributions were considered: when P1 and P2 are at one order of magnitude apart but each is at equal distance from Pf , this case is named as the 50-50 option. When P1 is moved closer to Pf in a condition where 40% of the “one order of magnitude apart” distance corresponds to the Pf to P1 separation, and thus the Pf to P2 distance is 60% of the total separation, then this case is named as the 40–60 option. The same principle applies for the 30–70 option. The resulting ALE and ELE return periods for the three conditions are presented in Table 7 and Table 8 for the Fugro and Arup models respectively. Negligible variations are observed in the ELE and ALE results for the three P1 - P2 distributions used to define aR . These results may be explained by the fact that normally a seismic hazard curve is plotted on a Log-Log scale which shows a linear trend from medium to low probabilities of exceedances (i.e. from intermediate to large return periods). It can therefore be concluded that for this case study the ISO methodolgy represents a consistent procedure that does not depend on the segment of the hazard curve used to define aR .

913

Table 8. Arup ELE and ALE return periods (in years) calculated from different aR values. Exposure & Design Levels

L1

L2

ELE

ALE

ELE

ALE

ELE

ALE

50 – 50 40 – 60 30 – 70

475 475 470

3125 3140 3180

270 273 273

1190 1190 1190

130 129 129

450 450 450

levels in the world. The conclusions reached herein may point to recommending the use of ISO 19901-2 over API RP-2A as a more robust design guideline. A review on the extensive practice of scaling the API DLE ground motions as twice the SLE level, shows that unconservative ground motion criteria may be obtained for design. This practice appears to be the result of a misinterpretation of API guidelines.

L3

REFERENCES Table 9. Design PGA from the combination of seismic models and design guidelines. Values in brackets present the ground motions return periods in years. API Guideline & Design Level Arup Fugro

6

ISO

SLE

DLE

ELE

ALE

0.27 g (200) 0.38 g (200)

0.88 g (2000) 0.81 g (2000)

0.32 g (270) 0.35 g (157)

0.70 g (1175) 0.71 g (1260)

COMPARISON OF DESIGN LEVELS

A brief comparison of design PGA levels between Arup and Fugro with API and ISO is presented in this section. The design levels from the possible combination of seismic models and design guidelines are presented in Table 9. From these results it is observed that ISO procedure produces smaller ALE PGA values than API DLE. For the API results the Arup model gives higher DLE PGA values. The Arup results using the ISO procedure are very similar.

7

CONCLUSIONS

The variability of design levels from the application of API RP-2A and ISO 19901-2 guidelines have been reviewed in the light of the offshore project currently under development by the Block 1 Operator in the Caspian Sea, Turkmenistan. Two independent albeit robust seismic models, developed by Fugro and Arup, were used in order to determine the sensitivity of the design guidelines with respect to the different seismic hazard models. From the analysis carried out, it may be concluded that for the particular case under study the API RP-2A approach presents a significant degree of dependence in the seismic model employed when compared to the ISO 19901-2 results. Therefore for this site, ISO 19901 represents a more robust procedure in terms of the definition of the ground motions for design. The extension of the findings presented in this study to a more generalised conclusion may be obtained by carrying out similar studies for regions with different seismicity © 2011 by Taylor & Francis Group, LLC

Arup (2007). Turkmenistan Block 1 Gas Development Project – Probabilistic Seismic Hazard Assessment. A report for Petronas Carigali (Turkmenistan) Sdn. BhD. Abrahamson, N.A. & W.J. Silva (1997). Empirical response spectra attenuation relations for shallow crustal earthquakes. Seismological Research Letters 68(1), 94–127. Ambraseys, N.N., K.A. Simpson & J.J. Bommer (1996). Prediction of horizontal response spectra in Europe. Earthquake Engng and Structural Dynamics 25, 371–400. API RP-2A-WSD (2000). Recommended practice of planning, designing and constructing fixed offshore platform – Working stress design. 21st edition. American Petroleum Institute. Bommer, J.J. & J.E. Alarcon (2006). The prediction and use of peak ground velocity. Journal of Earthquake Engineering 10(1), 1–31. Cornell, C.A. (1964). Stochastic processes in civil engineering. Ph.D. Thesis, Department of Civil Engineering, Stanford University. Cornell C.A. (1968). Engineering seismic risk analysis. Bulletin of the Seismological Society of America 58(5), 1583–1606. Crouse, C.B. (1991). Ground motion attenuation equations for earthquakes on the Cascadia subduction zone, Earthquake Spectra 7, 201–236. Esteva L. (1967). Criterios para la construcción de espectros para diseño sísmico. Proceedings of the XII Jornadas Sudamericanas de Ingeniería Estructural y III Simposio Panamericano de Estructuras, Caracas. Fugro West Inc (2006). Design Ground Motions for Platform and GBS Locations, Turkmenistan, Caspian Sea. A report for Petronas Carigali (Turkmenistan) Sdn Bhd. Gardner, J.K. and Knopoff, L. (1974). Is the sequence of Earthquakes in Southern California, with Aftershocks Removed, Poissonian?, Bulletin of the Seismological Society of America, 64(5), 1363–1367. ISO 19901-2 (2004). International Standard ISO 19901-2, Petroleum and natural gas industries – Specific requirements for offshore structures- Part 2: /seismic design procedures and criteria. Jackson, J., Priestley, K., Allen, M. and Berberian, M. (2002). Active Tectonics of the South Caspian Basin. Geophysical Journal International, 148, 214–245. Knapp, C.C., Knapp, J.H., and Connor, J.A. (2004). CrustalScale Structure of the South Caspian Basin Revealed by Deep Seismic Reflection Profiling. Marine and Petroleum Geology, 21, 1073–1081. Sadigh, K., C.Y. Chang, J.A. Egan, F. Makdisi & R.R. Youngs (1997). Attenuation relationships for shallow crustal earthquakes based on California strong motion data. Seismological Research Letters 68(1), 180–189. Youngs, R.R., S.J. Chiou, W.J. Silva & J.R. Humphrey (1997). Strong ground motion attenuation for subduction zone earthquakes. Seismological Research Letters 68(1), 58–73.

914

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Offshore geotechnics – safe and sustainable J. Peuchen & J. Haas Fugro, Leidschendam, The Netherlands

ABSTRACT: Offshore Geotechnics is a service industry supporting the installation, operation and decommissioning of offshore facilities. Client industries, particularly those that extract energy from offshore regions, regard offshore geotechnics as providing “added value”. What specifically comprises “added value” changes over time, as industry and societal values change. The growing awareness of the importance of safety and sustainability has affected offshore geotechnics in the past and will continue to offer challenges in the future. This paper describes trends in thought about safety and sustainability in offshore geotechnics within a pragmatic context of adding value to the industries it serves. Future directions for the industry are suggested.

1

INTRODUCTION

Offshore Geotechnics fits within a group of service industries supporting the installation, operation and decommissioning of offshore facilities. Client industries, particularly those that extract energy resources from offshore regions, regard offshore geotechnics as providing “added value”. What specifically comprises “added value” changes over time, as industry and societal knowledge bases and values change. Historically, the start of offshore geotechnics as an industry may be traced to the first offshore sample borings in the Gulf of Mexico. The acquired information and clever engineering calculations helped development of a significant offshore hydrocarbon industry. Traditionally, economy, and its derivative “technology”, were the key drivers. Particularly, offshore geotechnics achieved added value by adopting or adapting technologies and practices from diverse sources: ocean/earth research, oil and gas exploration and production organisations, resource mining operations, the military, and traditional onshore geotechnics. Lord Cullen’s report (1990) on the Piper Alpha disaster of 6 July 1988 changed perceptions about offshore operations. As a result, another driver was progressively and successfully adopted for offshore geotechnics: safety. At about the same time, the Brundtland team (1987) published on global sustainability, highlighting social responsibility.They defined sustainability as “economic development that meets the needs of the current generation without endangering the opportunities for meeting the needs of future generations”. The sustainability concept is broad and complex. Particularly, it is difficult to agree performance indicators that are consistent and transparent (Boyle & Depraz 2006). Some industry segments recognised this difficulty and developed guidelines on voluntary © 2011 by Taylor & Francis Group, LLC

915

corporate reporting on sustainability, for example IPIECA and API (2003). A more general approach is offered by the draft version of ISO 26000 (ISO 2009a). This standard provides guidance on ways to integrate and manage socially responsible behaviour into existing organisational strategies, systems, practices and processes. For service industries the concept of sustainability is slowly emerging and only now (2010) increasingly mandatory for individual service organisations (for example SenterNovem 2008). Yet, sustainability issues for an industry are more complicated than simple “stacking” of sustainability results achieved by individual organisations contributing to the industry. Voluntary collaboration and global and regional incentives are believed necessary for genuine advancement at industry level. 2 TERMINOLOGY It would seem that much still has to be developed to effectively implement sustainable practice in offshore geotechnics. For offshore geotechnics, the denotation given in Figure 1 offers some pragmatic guidance. It is clear from Figure 1 that sustainability encompasses safety and economy values. This should be no surprise as sustainability ultimately converges to some fundamental attitudes of individuals, organisations and societies toward man, things and nature. From this point, any usage of “sustainability” will include “safety”. A further split into “service” and “facility” issues is convenient. “Service” refers to activities in offshore geotechnics, including any on-site geotechnical operations. “Facility” refers to Client facilities and operations affected by information and advice produced by the offshore geotechnics industry i.e. the service product. The sustainability of offshore geotechnics must consider the interaction between service and facility

Figure 3. Illustrative example approaches for safe and sustainable practice: (E) modify and/or re-use, (I) improve, (A) use alternative.

2002). The scorecard measures performance in a number of important dimensions or indicators. The overall score is less important than the score compared to a reference score. The reference score is often indicated by a “footprint” in a radar diagram. The scorecard appears to suit offshore geotechnics, because offshore geotechnics adds value primarily in the early phases of the life cycle of a facility: site selection, development of facility options and detailed design. These early phases are especially influential for the sustainability of the facility (SISG 1993). The tools for offshore geotechnics typically comprise: (1) desk study and re-processing of available data, (2) geophysical survey, (3) in-situ testing, (4) sampling and laboratory testing, (5) empirical estimation, geotechnical modelling and risk assessment (ISO 2003, ISO 2009b, Van Staveren 2006). For the service scorecard, the dominating activities are those that involve offshore crews, vessels and equipment, i.e. (2), (3) and (4). Figure 4 shows proposed scorecards for assessing the sustainability of service and facility with respect to offshore geotechnics. Issues of particular importance to offshore geotechnics are indicated in italics. The principal topics are the same for both scorecards, the subcategories differ.

Figure 1. Sustainable practice according to IPIECA andAPI (2003).

Figure 2. Interaction between Service (S) and Facility (F): the total surface of both S and F relates to the overall improvement of sustainability.

aspects, as illustrated in Figure 2. Focus on service only may be at the expense of sustainability for the facility, resulting in net loss to society. Generic approaches for improving safe and sustainable activities for a facility or service are: (E) modify/re-use an existing facility or existing service data, (I) improve on existing concepts, (A) use alternative/new concepts. Figure 3 gives illustrative examples of the three approaches. In some cases it may be possible to evade the facility altogether, a special case ofApproach (E).Technology is often an important driver for improvement.

3

SCORECARDS: ASSESSING SUSTAINABILITY

4

The Infield Report (2009) forecasts 81,300 km of offshore pipeline to be installed globally over the 2009 to 2013 period. These pipelines may be placed in the following groups: in-field flowlines, tieback pipelines and grid interconnectors. As explained in Section 2 and Figure 3, three generic approaches are proposed (E), (I) and (A) for improving safe and sustainable practice of service and facility. The (E) approach is a difficult but obvious possibility for adding facility value to tieback pipelines and grid interconnectors. Transport of gas/fluid/solids by vessel instead of a pipeline would be an (A) example. Examples for improving (I) an existing facility concept could include:

Any assessment of alternatives requires benchmarking. Common tools for assessment are (a) scorecards, (b) life cycle techniques and (c) full valuations (Fisk © 2011 by Taylor & Francis Group, LLC

PIPELINES EXAMPLE

916

– Improved routing to suit seabed conditions and geohazards. – Improved site-specific design resulting in reduced use of pipeline materials.

Figure 4. Proposed score card for assessing the sustainability of service (a) and facility (b).

Figure 5. Example of using score cards to assess the safety and sustainability for both service and facility aspects of routing.

The following subsections clarify these examples, including the use of the proposed scorecards and the interaction between service and facility.

4.1 Improved routing Important trends observed for improving the routing of pipelines are (a) the role of Autonomous Underwater Vehicles (AUVs) for deepwater site characterization by geophysical techniques (e.g. Campbell & Burrell 2003) and (b) the increased use of knowledge repositories or databases. Figure 5 shows the resulting score © 2011 by Taylor & Francis Group, LLC

cards, comparing conventional techniques (dashed red line) and improved techniques (solid blue line) for (a) and (b) combined. In comparison with conventional towed systems, AUVs conduct geophysical survey close to the seafloor, improving data resolution and accuracy.AUV operations are not constrained to long tow cable, making curved survey lines easier to execute. An AUV survey is typically faster and results are more accurate. Service sustainability improves for the categories “Economy”, “Material Use, Waste and Recycling” and “Energy Efficiency and Emissions”. Facility sustainability improves for the same categories, primarily as a result of the higher data resolution allowing optimisation of facility design and risk management. Increasing use is made of knowledge repositories held by national geological surveys (public domain) and the regional geoscience industry (private domain) for routing of pipelines (Monteiro da Costa et al. 2009, Tervoort & Peuchen 2010). For offshore geotechnics as service, this leads to more efficient feasibility and planning of offshore surveys, predominantly adding value for the category “Economy” but also favourably affecting most of the other categories including safety. The use of knowledge repositories improves facility sustainability primarily for the design phase. The categories “Economy”, “Material Use, Waste and Recycling” and “Energy Efficiency and Emissions” have potential for significant improvement.

917

Figure 7. Example of using score cards to assess the safety and sustainability for both service and facility aspects of site-specific design.

Figure 6. SMARTSURF® operated in combination with a LARS.

It is suggested that “sparse data” technologies developed for oil and gas reservoir characterisation can add further value in the future. Surprisingly, some of these technologies rely on shallow CPT results (Geel & Donselaar 2007).

4.2

Improved site-specific design

The (I) approach for improved site-specific design is effective once a pipeline route has been selected. The observed trend for added value is one of integrated use of more advanced pipeline-soil models and specific geotechnical tools for on-site acquisition of important parameter values required for these geotechnical models. Examples of more advanced geotechnical models for surface-laid pipelines are given by White & Randolph (2007) and others. Examples of specific geotechnical tools are the FUGRO SMARTPIPE® (Hill & Wintgens 2009) for in-situ pipeline section testing and the companion SMARTSURF® for rapid in-situ testing and sampling of shallow seabed. The FUGRO SMARTPIPE® has been operational since 2008. It deploys an instrumented pipeline section, subjected to static and cyclic soil/pipe interaction forces in 3 dimensions (vertical, axial and lateral) in water depths up to 2,500 m. The equipment also includes supplementary cyclic mini T-bar and/or mini ball testing for characterising the upper portion of the seabed. The SMARTSURF® (Figure 6) allows efficient extrapolation of data acquired by pipeline section testing. This lightweight tool has been in use since 2009. It is equipped with retractable mudmats and © 2011 by Taylor & Francis Group, LLC

seafloor detection tools for operation on extremely soft seabed. Geotechnical tools include (a) 3 m push system for conventional CPT/T-bar/ball penetrometers, (b) a 1 m push system for pressure-compensated, cyclic mini T-bar or ball penetrometers and (c) 2 m PISTON SAMPLER. A single deployment permits use of each of these tools. Fugro’s LARS (Launch And Recovery System, Figure 6) suits the FUGRO SMARTPIPE® and the SMARTSURF® . It provides a step-change in service safety for deployments by A-frame. This is because of full remote control of back-deck operations. The scorecard of Figure 7 illustrates the interaction between the service and facility aspects according to Figure 2a. It can be inferred that improved sitespecific design increases on-site (vessel) operations. This has an adverse effect on service sustainability scores in the categories “Economy”, “Material Use, Waste and Recycling” and “Energy Efficiency and Emissions”. However, the facility score improves in the same categories, leading to significant overall benefits. It is suggested that a future scenario could possibly include regional “soil-pipeline interaction catalogues” which add strategic value to on-site testing.

5 WIND FARMS EXAMPLE The offshore geotechnics industry in Europe is showing rapid growth in line with rapid growth of offshore renewable energy, notably many wind farms. A single wind farm may comprise up to hundreds of identical bottom-founded offshore structures, spaced at about 500 m. Typically, the cost of offshore substructures, including foundation elements, accounts for about 15% to 25% of the total initial wind farm investment (Junginger et al. 2004). This unprecedented scenario is generating service and facility concepts for each of the categories (E), (I) and (A), including the topics of pile driving noise for bio-diversity and modular structure build-ups for social investments (Carbon Trust 2009). Up to about 2007, offshore geotechnics for wind energy converters saw approaches applied in the offshore oil and gas industry competing with traditional onshore geotechnics. The following (I) trends can

918

Offshore geotechnics offers added value by service. It is therefore essential to consider the interaction between service and facility aspects, i.e. the Client facilities and operations affected by information and advice produced by the offshore geotechnics industry. Focus on service only may be at the expense of sustainability for the facility, resulting in net loss to society. Scorecards appear practical. Specific criteria for scoring are proposed for offshore geotechnics as service and for facility sustainability. A brief examination of offshore geotechnics for pipelines and wind farms shows ongoing improvements and future opportunities for added value. These include: (a) enhanced integration of information from knowledge repositories, soil catalogues and advanced on-site testing; (b) implementation of “sparse data” technologies developed for oil and gas and gas reservoir characterisation; (c) introduction of rapid/robust probing techniques in combination with enhanced integration; and (d) development of alternative foundation concepts.

now (2010) be observed for tools used in offshore geotechnics: – As for pipelines, increasing use is made of knowledge repositories for characterising a wind farm site. Effective exploitation of this desk study tool greatly improves most factors of the service scorecard. – Most jack-up and one-off operations for in-situ testing and sampling are now replaced by specialist vessels, safe and with good weather endurance, with continuous operations only interrupted by periodic port calls and dry-docking. This “industrialisation” significantly improves service safety, but also improves scores on economy, energy efficiency and material use. – As evident elsewhere in offshore geotechnics, parameter values for geotechnical models are increasingly inferred from in-situ tests correlated to published, well-defined “soil catalogues”, for example based on results from over 60 International Geotechnical Experimentation Sites (Mayne et al. 2009). In practice, reliance on site-specific supplementary laboratory test results is limited. Insitu tests such as the Cone Penetration Test (CPT) provide results with reasonably well-defined accuracy (ISO 2010) compared to the usually overrated (Barends 2005) and lower accuracy achieved by a process of sampling, sample handling, laboratory testing (CUR 2002). This trend for in-situ testing adds significant value in terms of time schedule for data delivery and usually provides improvement on most other points of the service scorecard. It is suggested that extremely rapid/robust probing techniques can provide future potential. This should be in combination with knowledge repositories, advanced geophysical survey interpretation and regional soil catalogues. – Geotechnical modelling increasingly focuses on fine-tuning of soil-structure interaction. A brief literature survey indicated no less than 63 publications in three years on optimising the PY model for monopile design. However, limited future potential is foreseen here for further increase of safe and sustainable practice. The most significant offshore geotechnics service values for facility sustainability should probably be sought in helping to develop alternative (A) concepts for (a) offshore structures that can carry wind energy converters (Carbon Trust 2009) and (b) offshore structures for transport and storage of the generated energy (Carbon Trust 2009, Diepeveen 2009).

6

CONCLUSION AND DISCUSSION

Safe and sustainable practice is broad and complex. It is difficult to agree performance indicators that are pragmatic, consistent and transparent. For service industries the concept of sustainability is slowly emerging and only now (2010) increasingly mandatory for individual service organisations. © 2011 by Taylor & Francis Group, LLC

ACKNOWLEDGEMENTS The authors gratefully acknowledge Fugro’s commitment and support to improving safe and sustainable practice. The opinions expressed in this paper are those of the authors. They are not necessarily shared by Fugro. REFERENCES Barends, F.B.J. 2005.Terzaghi Oration 2005;Associating with advancing insight. In Proceedings of the 16th International Conference on Soil Mechanics and Geotechnical Engineering; 16ICSMGE, 2005, Osaka, Vol. 1:217–248. Rotterdam: Millpress. Boyle, B. & Depraz, S. 2006. Oil and gas industry guidance on voluntary sustainability reporting. In SPE international conference on health, safety, and environment in oil and gas exploration and production held in Abu Dhabi, U.A.E., 2–4 April 2006. Paper no. SPE 98585. Brundtland, G.H. (chairman), World Commission on Environment and Development 1987. Our common future. Oxford: Oxford University Press. Campbell, K.J. & Burrell, R. 2003. Deepwater development fast-tracking: the critical role AUV surveys play in integrated site investigation and geohazard assessment. In 7th Annual Offshore West Africa Conference and Exhibition, 11–13 March 2003, Windhoek, Namibia. Carbon Trust. 2009. Carbon Trust and Ed Miliband launch global competition to cut the cost of offshore wind energy. http://www.carbontrust.co.uk/news/news/presscentre/2009/pages/2009.aspx (accessed January 18th, 2010). Cullen, W.D. (Lord) 1990. The public inquiry into the Piper Alpha disaster. London: HMSO. CUR Civieltechnisch Centrum Uitvoering Research en Regelgeving. 2002. Research of compression tests. CURrapport 2002-2. Gouda: CUR. (in Dutch).

919

Diepeveen, N.F.B. 2009. Closed-loop fluid pumping as a means to transfer wind energy. In European Offshore Wind Conference & Exhibition Stockholm, Sweden, 14–16 September 2009. Fisk, D. 2002. Sustainability & diversity – after the World Summit. In M. Cook, M. et al. (eds.), Offshore site investigation and geotechnics; ’Diversity and sustainability; Proceedings of an international conference held in London, UK, 26–28 November 2002: 9–12. London: Society for Underwater Technology. Geel, C.R. & Donselaar, M.E. 2007. Reservoir modelling of heterolithic tidal deposits; sensitivity analysis of an object-based stochastic model. Netherlands Journal of Geosciences – Geologie en Mijnbouw 86(4): 403–411. Hill, A.J. & Wintgens, J.F. 2009. In-situ measurement of pipesoil interaction in deep water – results of a successful offshore campaign. In Proceedings of the SUT Annual Conference 2009, Perth, Western Australia. Infield Systems Limited. 2009. Offshore pipelines & control lines market update report 2009/13. London: Infield Systems Limited. IPIECA International Petroleum Industry Environmental Conservation Association & API American Petroleum Institute. 2003. Compendium of sustainability reporting practices and trends for the oil and gas industry: joint corporate reporting project. London: IPIECA, & Washington: API. ISO International Organization for Standardization. 2003. Petroleum and natural gas industries: specific requirements for offshore structures – Part 4: geotechnical and foundation design considerations. International Standard ISO 19901-4. Geneva: ISO. ISO International Organization for Standardization. 2009a. Guidance on social responsibility. Draft International Standard ISO/DIS 26000. Geneva: ISO. ISO International Organization for Standardization. 2009b. Risk management – Principles and guidelines. International Standard ISO 31000. Geneva: ISO.

© 2011 by Taylor & Francis Group, LLC

ISO International Organization for Standardization. 2010. Geotechnical investigation and testing – Field testing – Part 1: Electrical cone and piezocone penetration tests. ISO/NP 22476-1. Junginger, M., Faaij, A. & Turkenburg, W.C. 2004. Cost reduction prospects for offshore wind farms. Wind Engineering 28(1): 97–118. Mayne, P.W., Coop, M.R., Springman, S.M., Huang, A.B. & Zornberg, J.G. 2009. Geomaterial behavior and testing. In M. Hamza, M. Shahien, & Y. El-Mossallamy (eds.), Proceedings of the 17th International Conference on Soil Mechanics and Geotechnical Engineering, Alexandria, Egypt, 5–9 October 2009: 2777-2872. Rotterdam: Millpress. Monteiro da Costa, R., Lopes Teixeira, J., Des Déserts, L., Harris, R.J.S. & Burban, P.Y. 2009. The intentions and achievements of the Angola Deepwater Consortium. In OTC.09: Proceedings 2009 Offshore Technology Conference, 4–7 May, Houston, Texas, USA. OTC Paper 19940. SenterNovem. 2008. How to purchase sustainable external consultancy. The Hague: SenterNovem. (in Dutch). SISG Site Investigations Steering Group. 1993. Site investigation in construction. 4 Vols. London: Thomas Telford. Tervoort, E. & Peuchen, J. 2010. Exploring the shallow subsurface of the North Sea. In First International Conference on Frontiers in Shallow Subsurface Technology, 20–22 January, 2010, Delft, The Netherlands. Van Staveren, M.Th. 2006. Uncertainty and ground conditions: a risk management approach. Amsterdam: Elsevier. White, D.J. & Randolph, M.F. 2007. Seabed characterisation and models for pipeline-soil interaction. In The proceedings of the Seventeenth (2007) International Offshore and Polar Engineering Conference ISOPE, Lisbon, Portugal, July 1–6, 2007, Vol. 2: 758–769. Cupertino: International Society of Offshore and Polar Engineers.

920